Steel Construction 2017 01 free sample copy

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Steel Construction

Volume 10 February 2017 ISSN 1867-0520

Design and Research

– Method for assessing fire safety of bridges – Composite columns with non-Eurocode cross-sections – Optimized preliminary structural design of composite buildings – Design of composite bridges with integral abutments – Remote monitoring of structural health in composites – Vehicle dynamic effects in fatigue design of short bridges – Early-age shrinkage in beam-like composite structures – Effectiveness of high-frequency peening – Evaluation of cracks in an offshore crane runway girder – Steelworks for Amager Bakke waste-to-energy plant

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References: • Bahia de Cadiz, Spain • Hochmoselübergang, Germany • Izmit Bay Bridge, Izmit, Turkey • Mainbrücke Randersacker, Germany • Millau Viaduct, France • Rheinbrücke Schierstein, Germany • Rion Antirion, Greece • Russky Island Brigde, Vladivostok, Russia • Tsing Ma, China

forces in motion

Content

The viaduct over the water route Teltowkanal in Berlin/Germany is the first railway bridge implementing the VFT construction method with integral abutments. By now this construction method established by SSF Ingenieure became the state-of-the-art in the design of slender bridges crossing existing traffic routes (see pp. 23–30).

(SSF Ingenieure AG + Florian Schreiber Fotografie, Munich)

Steel Construction 1 Volume 10 February 2017, No. 1 ISSN 1867-0520 (print) ISSN 1867-0539 (online)

Editorial 1

Marcus P. Rutner Articles

2

Norifumi Yanagisawa, Yusuke Imagawa, Osamu Ohyama, Marcus P. Rutner, Akimitsu Kurita Fire safety of bridges – methodology supporting design and forensic evaluation

10

Gerhard Hanswille, Marco Bergmann, Reinhard Bergmann Design of composite columns with cross-sections not covered by Eurocode 4

17

Martin Mensinger, Li Huang Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer

23

Daniel Pak, Hetty Bigelow, Markus Feldmann Design of composite bridges with integral abutments

31

Behnoush Golchinfar, Dimitri Donskoy, Julius Pavlov, Marcus Rutner Remote monitoring of structural health in composites

37

Wei Zhang, Mengxue Wu, Jin Zhu Evaluation of vehicular dynamic effects for the life cycle fatigue design of shortspan bridges

47

Xi Li, Branko Glisic Evaluating early-age shrinkage effects in steel-concrete composite beam-like structures

54

Ingbert Mangerig, Robert Kroyer, Matthias Koller Experimental and numerical analyses of the effectiveness of high-frequency peening processes

67

Jaap Wardenier, Peter de Vries, Gerrit Timmerman Evaluation of cracks in an offshore crane runway girder Reports

72 http://wileyonlinelibrary.com/journal/stco

Johannes Hauptenbuchner Design and construction of the complex steel structure for the Amager Bakke waste-to-energy plant Regular Features

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Erratum Announcements People ECCS news Book reviews Discussion

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Products & Projects

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Well parked rather than badly driven The new storage and handling hall of Norline AG in Rümlang (Canton Zurich) offers plenty of space. It is used for the storage and loading of construction machinery and building materials for dispatch. The large vehicles required for this purpose are on site: Five impressive parking levels stretching along the entire hall make it possible. Doing without assembly supports enabled the project to impress by its fast, rational building method. Montana Building Systems Ltd. from Villmergen (Canton Aargau) provided support in the implementation of the car park project. Its composite steel systems enabled the particularly thin design of the levels, thus forming the basis for a highly economic overall solution.

Two in one: Norline AG, with its headquarters in Neuhausen am Rheinfall (Canton Schaffhausen) accommodates the logistics and construction divisions under one roof. In 2016, the company enlarged its storage and floor space in the Zurich area in order to be able to continue to offer its customers flexible logistics and a diverse selection of construction machines and products in future.

Fig. 2. The trapezoidal form of the SUPERHOLORIB® elements enabled the curvature in conjunction with the sloping gradients.

The right building material for plenty of space A sufficient amount of space is required to enable a construction company to comfortably organise its logistics services.

Norline AG now has this space in its new storage and handling hall in Rümlang not far from the airport in Zurich Kloten. The special feature of the building are the five parking levels provided above the hall. Due to the immense span of the installed steel girders, the parking areas could be realised with few supporting pillars. Thus, a spacious car park came into being, suitable for the accommodation of many – even large – vehicles. The view at the official opening ceremony was particularly impressive: The building owner had an original Formula BMW racing car whizzing around the new parking levels. The video of the event impressively shows how much space the individual parking levels and the generous driveway have to offer.

Norline AG looked for a particularly viable and economic solution for the implementation of the vast car park. The SUPERHOLORIB® composite steel flooring from Montana AG provided just the right static properties here. This composite profile, consisting of high-quality, corrosion-protected sheet steel, forms the basis for composite floor decks that optimally combine the

Fig. 3. The composite profile SUPERHOLORIB®, consisting of high-quality, corrosion-protected sheet steel, enabled a fast, rational and economical building method. (© montana)

building structure properties of steel and concrete. While the steel inserts are embedded in the concrete layer in classic reinforced or pre-stressed concrete and only serve there as reinforcing elements for the concrete, the steel profiles and reinforced concrete parts are connected in steel composite construction by means of the undercut roll tool. This enables the strengths of both building materials to be optimally exploited, because: Here, the steel is responsible for accommodating the tensile forces, and the concrete for eliminating the compressive forces.

Slim, strong and economical

Fig. 1. The perforated SWISS PANEL® facade profiles contribute to a light atmosphere and good visibility in the car park.

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Steel Construction 10 (2017), No. 1

The successful combination of steel and concrete enables exactly what Norline AG wanted for its new car park – a fast, rational and economical building method. But this was not the only decisive aspect: Due to the optimal discharge of the exerted forces, material – and hence dead weight – can be saved. Therefore, considerably slimmer constructions can be achieved using composite steel systems than using conventional concrete building methods. In addition, composite structures offer a higher fire resistance and sound insulation as well as a higher thermal storage capacity compared with steel structures. Due to the material saved, the composite systems of Montana Building Systems Ltd. are also considerably cheaper than solid concrete floors or a steel construction. In addition, the SUPERHOLORIB® composite profile is an industrially pre-fabri-

Safety in icy conditions Finding a solution to the problem of enabling safe use of the ascents to the parking levels throughout the winter posed a further challenge. “It was of great advantage here that the composite steel floor could be combined with thermal activation”, emphasises Schweizer. The socalled thermal component activation enables the use of the building elements – meaning walls or floors – for room heating or cooling. For this purpose, pipes are installed in the corresponding building components, through which heating or cooling water flows. The component then discharges the temperature of the water over its entire surface. Icing of the ramp in winter can thus be pre-

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Ovako launches Steel Navigator

Cross-Sections

Bridge Construction

Ovako recently launched Steel Navigator, a digital platform that allows customers to search for and identify the best steel for their applications. The investment in this tool is part of Ovako’s global digitalization strategy and underscores the group’s objective to provide its customers with the best service in the industry.

“The steel industry is facing a major transformation with the ongoing wave of digitalization, and we want to be a pioneer in this work,” says Göran Nyström, EVP Group Marketing & Technology.

“The launch of Steel Navigator enhances our service offering to the market for engineering steel and will help our customers to quickly identify the right steel for their application, making their processes more efficient.”

Steel Navigator allows customers for search for specific steel grades by group, quality, type of process, product and chemical composition, not just among Ovako’s steels but also among steels supplied by its competitors.

Steel Navigator is being launched in conjunction with the Elmia Subcontractor trade fair in Jönköping and is available globally at “steelnavigator.ovako.com.” www.ovako.com

© www.novumstructures.com

3D Frameworks

Stability and Dynamics

Apart from the vast parking levels, there was another element that saw planners and assembly team faced with a special task: The ascent to the five parking levels on the upper floors. This was designed as a roundabout system and offered various gradients that had to be planned and subsequently put into practice. The on-site installation of the profiles was easy thanks to excellent advance planning.

The SUPERHOLORIB® elements for the ascent were created in small sections to enable the curvature in conjunction with the sloping gradients. Each of these being trapezoidal, enabling the components to finally form an overall rising polygon.

This optimally fits the round shape of the ascent driveway. “All in all we were able to achieve a final floor thickness – for the parking levels and ramp – of only 150 millimetres by means of the SUPERHOLORIB® profile” says Reto Schweizer, Project Manager of the general contractor SM Bau AG. “That would not have been possible with a solid concrete floor.” A total area of 7,200 square metres was equipped with the composite steel floor in Rümlang.

As the composite profiles were delivered packaged in compact bundles ex works, they only required a small storage space on site, thus causing only low traffic on the construction site. The storey-by-storey installation of the steel sheets, which took place directly after the assembly of the load-bearing structure, also increased on-site safety. The elements were quickly and easily installed thanks to the exact pre-fabrication.

Structural Analysis and Design

© www.wirth-baustatik.de

Strong compound – strong hold

vented. Due to the comparatively large transfer surface, the heating water must not be heated to the same extent that is necessary for central heating in residential buildings, for example. This makes thermal component activation a particularly energy-efficient heating method. The ideal solution for a car park is use all year round.

And Montana AG was also able to help again when it came to the aesthetic design of the car park. The parking levels and the ascent roundabout were partly clad with its SWISS PANEL® facade profiles. The top-quality aluminium elements provide a visual boundary on the outside of the car park. Due to their perforation in a fine hole pattern they are translucent and contribute to a light atmosphere and good visibility in the car park. The special light pattern arising from these elements is another attractive aspect. This way, a highly-modern car park was created, not only in terms of design, but also of functionality, which enables the complete and extremely flexible use of its overall area of 10,000 m2.

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cated component, which considerably reduces construction time. This also applies to the Norline car park in Rümlang: Completion of the project was achieved in a total construction period of only 6 months, which was greatly appreciated by the building owner. And this although the building posed several challenges.

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Refurbishment of the Silesian Stadium Roof The Silesian Stadium in Chorzow near Katowice has a long history. First built in the 1950s it has been operating as a venue for football, motor sports and concerts. Since 1956 it has been declared a National Stadium of Poland.

A complex refurbishment process started in stages going back to 1996, with the aim to reach standards for the EURO 2012. Facilities and stands have been updated over continuing years. By 2007 it had been planned to top the stadium up with a newly designed roof.

But the building process for the bowl had been delayed. In 2008 the new Warsaw stadium was declared National Stadium instead of this one in Silesia. Nevertheless construction of a new roof in Chorzow was started to achieve training status for the EURO 2012. Then a failure during the erection process of the roof brought the works to a standstill that took about 4 years.

A quite special renovation job In 2013 the Silesian government (the owner) finally managed to tender out the refurbishment of the roof structure under a new contract. This actually meant a quite special renovation job. The task was to take over an existing but never finished design, to examine and hopefully accept already existing supplies for the roof, to combine these with the balance of other parts and materials yet to be determined on a basis of an amended design.

With all its company and personal experiences from refurbishment projects like Olympic Stadium Berlin, London Olympic Stadium (Legacy mode) and Khalifa Stadium in Qatar, PFEIFER was qualified and has tendered for the complete job.

Soon after the company was appointed by the Silesian Govern-

Do you require special prints of selected articles from the journals at Ernst & Sohn?

Please contact: Janette Seifert Verlag Ernst & Sohn Rotherstraße 21, 10245 Berlin Tel +49(0)30 47031-292 Fax +49(0)30 47031-230 E-Mail Janette.Seifert@wiley.com

The Silesian Stadium is again the pride of the people of Silesia and will serve as a venue for local football clubs, motorcycle races, concerts, international matches, and other events. (© Pfeifer)

ment directly as a specialty roof contractor, with responsibility for and coordination of cable net works, repair works to the steel structure, and the polycarbonate roofing works.

In a carefully fine-tuned process, the investigation and the verification started for the cable net that had been lying on the ground for quite a long period of time. Decision was made to introduce new node points to the net since updated standards required a revised analysis anyway. The cables were inspected on site, a portion of them coiled, sent to the company’s plant in Memmingen, Germany and tested there. Only a few replacement cables were required. The cable net, combined from inspected and accepted old parts and several new elements was now ready for the new works on site. A reliable erection process was proposed by the PFEIFER team, analyzed by the engineers and verified by the Silesian government. This way any interruption or failure could be eliminated and the entire cable net went up in a big lift at the end of 2015.

During the erection process of the structure a different verification process happened for the existing supply parts of the future polycarbonate roofing and its substructure. Each and every part of a number of about 5,000 pieces was examined and either accepted or replaced. New parts according to the amended design for construction were purchased. An erection process was produced to facilitate as much pre-assembly work as possible to be done on the ground, for both, the substructure and the polycarbonate roofing.

Again the pride of the people of Silesia The works are completed with a monitoring system for the roof that measures the environmental conditions, scans the loads within selected cable members and, based on the initial structural analysis, evaluates from these sources the load history of the whole roof.

The Silesian Stadium has been finalized in 2016. It is again the pride of the people of Silesia and will serve as a venue for local football clubs, motorcycle races, concerts, international matches, and other events .

For PFEIFER, this project has proven again the outstanding ability to evaluate and manage risks properly, to provide safe methods of erection within a situation of existing building and to find solutions where others failed. In a totally different environment PFEIFER is about to tackle its next refurbishment project very soon: replacing the roof on top of the Kuala Lumpur Olympic Swimming Pool. www.pfeifer.de

www.ernst-und-sohn.de/en/reprints 1009286_dp

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Steel Construction 10 (2017), No. 1

Editorial

This issue is a collection of scientific contributions from three countries, Germany, Japan and the USA, and is dedicated to Prof. Dr. Eng. Akimitsu Kurita, Osaka Institute of Technology, for his excellence in scientific work and his dedication in mentoring and teaching.

Eight scientific articles focus on topics with current research demand, addressing innovative design, life cycle assessment and management of steel and steel-concrete composites structures.

The first paper deals with an unanswered question in fire safety design, i.e. the ability to identify efficiently the governing failure mode of a bridge subjected to severe fire loading and ranking the regions of the structure with the greatest fire exposure risk.

The second paper fills a gap in the current Eurocode 4. For specific composite column types, the general design method, as it is proposed, gives no guidance on the verification of the required level of safety. The study introduced here provides clarity and a solution in this respect.

The third paper links sustainability and structural design undertaken in the preliminary design phase in the form of an efficient automated methodology that targets optimization in the preliminary structural design of steel-concrete composite buildings.

The fourth paper provides an overview of the design of composite bridges with integral abutments. This type of

bridge design enables slender superstructures and combines advantages with respect to production, maintenance and economic and socio-economic costs.

The fifth paper introduces a new method for detecting small-scale internal defects in composites. The method allows the level of integrity of structural components with complicated geometry to be quantified within seconds of response time.

The sixth paper introduces a new dynamic amplification factor for evaluating vehicle dynamic effects for the life cycle fatigue design of short-span bridges taking into account environmental effects.

The seventh paper focuses on the phenomenon of concrete early-age shrinkage effects in steel-concrete composite structures. The objective was to develop an efficient analysis method coupling physical monitoring with numerical simulation.

The eighth paper describes a method that increases the longevity of metal components. Experimental tests and numerical analyses were conducted in order to shed light on the effectiveness of high-frequency peening processes.

I hope you enjoy reading the contributions as much as I did.

Dr.-Ing. habil. Marcus P. Rutner, Associate Professor Stevens Institute of Technology, USA

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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Articles Norifumi Yanagisawa Yusuke Imagawa Osamu Ohyama* Marcus P. Rutner Akimitsu Kurita

DOI: 10.1002/stco.201710002

Fire safety of bridges – methodology supporting design and forensic evaluation Dedicated to Prof. Dr. Eng. Akimitsu Kurita on his 70th birthday, in honour of his scientific achievements, guidance and the education of his students.

Fire safety in bridge design is not as developed as fire safety in building design, even though a bridge failure can cause significant economic damage impacting on an area. This paper addresses an unanswered question with regard to fire safety, i.e. the capability to identify the governing failure mode of a bridge subjected to severe fire loading and ranking the regions of greatest fire exposure risk. Hence, this proposed methodology is also expected to support forensic work identifying the failure mode where a bridge has failed due to a severe fire, as will be shown using the 9-Mile Road Overpass collapse as an example. In an effort to mitigate fire damage, the fire protection panel (FFP) is introduced, which is part of a sacrificial structure shielding the bridge superstructure from exposure to fire from underneath.

1 Introduction Fire counts among the most severe environmental hazards for transportation infrastructures [1]. However, fire safety is the least developed area within fire science [2]. Bridges are part of the transportation infrastructure and since they often constitute “bottlenecks”, their loss or closure due to fire would cause significant economic impact on an area [3]. Bridge design codes do not account for fire safety [2].

In the US the National Fire Protection Association developed provisions in the NFPA 502 standard [4] which pertain to the fire safety design of bridges. However, the information is very limited, comprising the following general statement: “Critical structural members shall be protected from … high temperature exposure that can result in dangerous weakening or complete collapse of the bridge or elevated highway” [4]. Most bridges in the US are not capable of resisting severe fire conditions [2], [5]. In an effort to define bridge fire hazard levels, a National Cooperative Highway Research Project (NCHRP) elaborated on highway bridge fire hazard assessment, as published in [6]. The primary objectives of this study were: 1) the determination of the susceptibility of highway bridges to fire damage, 2) the development of damage assessment and repair techniques and their manifestation in guidelines, and 3) the development of guidelines that facilitate reductions in fire damage risk.

* Corresponding author: osamu.oyama@oit.ac.jp

2

Over the last two decades, various severe bridge fires have occurred which resulted in partial collapse of the bridge structure. Subsequent forensic work, e.g. [7], [8], led to suggestions for changes in design to account for severe fire loading. Examples of fire incidents in the US in which bridge infrastructure was severely damaged are the MacArthur Maze bridge fire in California in 2007 and the 9-Mile Road Overpass bridge fire near Detroit in 2009. A post-fire photo of the 9-Mile Overpass, which was a steel-concrete composite girder bridge, is shown in Fig. 1a [9]. Examples in Japan are the bridge damage on the Ikebukuro line along the Metropolitan Expressway in 2008, as shown in Fig. 1b [10]. The steel girder deformed due to shear buckling in the region of the support. As a consequence, this urban expressway was closed for about six weeks for repairs. According to the Metropolitan Expressway Company Ltd., the economic loss following this accident was approx. € 38 million.

The proposed methodology introduced here addresses all three objectives as mentioned in the NCHRP report.

The methodology quantifies the susceptibility of a specific bridge to specific fire damage. Further, the methodology allows quick damage assessment, provided the fire signature, fire pool size and location are known beforehand.

The proposed methodology can be easily implemented in guidelines.

The methodology is introduced in section 2 and applied using the example of the 9-Mile Road Overpass in sections 2–5. The temperature-dependent material properties assumed are discussed in section 3. Assumptions for the analysis of the reduced load-carrying capacity representing specific failure mechanisms of the composite bridge are introduced in section 4 and the load-carrying capacities and heat transfer analyses are investigated. The time taken to failure is discussed in section 5. Section 6 looks at future topics for discussion and the conclusions follow in section 7.

2 Analytical model Fig. 2 shows the system of and a section through the 9-Mile Road Overpass. A reinforced concrete slab is supported by 10 steel W36w135 girders to create a steel-concrete composite bridge. Fig. 2 also shows the location of the fire accident.

Table 1 lists the dimensions and material properties of the structural members making up the overpass.

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

Fig. 1. Examples of fire accidents

Fig. 2. General view of 9-Mile Overpass (units: mm)

A critical shortcoming in the design of steel girder bridges for fire safety is the fact that only forensic findings are available, as pointed out in [11], a study motivated by the US Department of Homeland Security (DHS). The study comes to the conclusion that there is a significant lack of information with regard to 1) the regions of the overpass which face the greatest fire exposure risk and 2) the fire signature or maximum temperatures reached.

This lack of information is addressed in this research study by searching for possible answers through a thorough structural investigation. The scope of work in this research is: 1) to define potential failure mechanisms of the bridge structure when subjected to a temperature increase and identifying which structural members or cross-sections are at risk and trigger collapse of the bridge, 2) to calculate the stress resultant due to dead load acting in each structural member or cross-section identified,

3) to calculate – for each of the structural members or cross-sections identified – the fire duration required to reach the temperature such that the decreased load-carrying capacity of the structural members identified equals the stress resultant due to dead load as calculated in (2), thus triggering the respective failure mechanisms, and 4) to order the failure mechanisms with respect to criticality according to the fire exposure durations computed.

The study of the structural system of the overpass, as shown in Fig. 2a, reveals four potential failure mechanisms and assigns a specific structural member or cross-section to each failure mode, as listed below: – Failure of suspender due to tensile axial force – Shear failure at the intermediate support – Positive bending moment failure at the centre of the main girder

Steel Construction 10 (2017), No. 1

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N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

Table 1. Dimensions and material properties

Main girder

Width of flange: 303.4 mm Thickness of flange: 20.2 mm Thickness of web: 15.2 mm Depth of girder: 903.0 mm Yield strength: 248 N/mm2

Side girder

Width of flange: 304.8 mm Thickness of flange: 25.9 mm Thickness of web: 16.6 mm Depth of girder: 914.4 mm Yield strength: 248 N/mm2

Positive moment area

Thickness of slab: 203 mm Effective width: 2080 mm Strength of concrete: 30 N/mm2

Negative moment area

Thickness of slab: 203 mm Effective width: 912 mm Strength of concrete: 30 N/mm2

Steel girders

Concrete slab

Reinforcing bars

Area: 285.2 mm2 Yield strength: 400 N/mm2

Suspenders

Width of plate: 190.5 mm Thickness of plate: 19.1 mm Yield strength: 248 N/mm2

– Negative bending moment failure at the intermediate support of the side girder The following analysis is intended to clarify which of these failure modes is the most critical one and most likely triggered the collapse of the 9-Mile Road Overpass. The stress resultant due to gravity loading for each critical structural member or cross-section mentioned above is listed in Table 2.

3 Yield strength of carbon steel and compressive strength of normal concrete at high temperatures One of the main questions to be answered regards the area of fire exposure and temperature distribution along the bridge span. According to photos (e.g. Fig. 1a) and reports [11], the entire span was engulfed in flames. Hence, a uniform temperature profile along the whole bridge span can be assumed. This statement is supported by a photo of the post-fire damage state of the bridge span, which collapsed along the entire bridge span, landing on the roadway [2].

The Eurocodes provide reduction factors for the yield strength of carbon steel [12] and compressive strength of normal concrete [13], as shown in Figs. 3a and 3b respectively.

Fig. 3a reveals that carbon steel begins to lose yield strength when temperatures exceed 400 °C. The decrease rate of compressive strength of normal concrete (Fig. 3b) is less pronounced, but already starts at a temperature of about 100 °C. The temperature represented by the x axis in the diagrams in Fig. 3 rises to 1200 °C. It is assumed that the uniformly distributed fire had a maximum temperature of 1100 °C. Please note that forensic studies of the MacArthur Maze bridge collapse assume similar maximum temperatures, specifically 1100 °C during a pre-collapse phase lasting 37 min and 890 °C during a post-collapse phase lasting 70 min [14].

4 Load-carrying capacities of structural members identified which triggered the failure modes of the 9-Mile Overpass at high temperature The load-carrying capacities of the structural members identified which triggered a specific failure mode of the bridge, as discussed in section 2, are calculated next. The

Table 2. Stress resultants due to dead load Stress resultant due to dead load 1) Tensile normal force in suspender

183 kN

2) Shearing force at intermediate support

210 kN

3) Positive bending moment at centre of main girder

1017 kNm

4) Negative bending moment at intermediate support

330 kNm

Fig. 3 Reduction factors for yield strength of carbon steel and compressive strength of normal concrete

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Steel Construction 10 (2017), No. 1

N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

Fig. 4. Full plastic tensile force depending on temperature

Fig. 5. Shear resistance depending on temperature

following assumptions were made for setting up the equations: 1) There is no slip between steel beam and concrete slab.

2) The cross-section can reach the full plastic condition without prior failure through local buckling.

3) The stress-strain relationship for steel, in both tension and compression, is described by perfectly elastic–plastic behaviour.

4) The stress distribution in the concrete between upper concrete surface and plastic neutral axis is assumed to be constant and 0,85Xck, where Xck is the characteristic concrete compressive strength.

5) The shearing force resistance of the composite girder can be evaluated by considering only the contribution of the shear capacity from the steel web.

where Aw is the sectional area of the steel web (mm2).

Fig. 5 shows the relationship between the shearing force capacity and the temperature. By interpolation, the critical temperature is found to be 820 °C. At that temperature the shearing force capacity equals the shear force due to dead load.

1) Failure of suspender due to tensile axial force The resistance NθpC of a tension member at a uniform temperature can be expressed as follows: NθpC

= κ y,θσ y A k

(1)

where: ky,V reduction factor for yield strength of carbon steel (see Fig. 3a) Xy yield strength of steel (N/mm2) Ak sectional area of steel (mm2) Fig. 4 shows the relationship between temperature and full plastic normal force. Fig. 4 readily shows that the tensile force capacity of the suspender equals the tensile force due to dead load at 830 °C. 2) Shear failure at intermediate support The shear resistance QθpC at the intermediate support is given by the following equation (considering a uniform temperature distribution): QθpC = κ y,θ

σy 3

Aw

(2)

3) Positive bending moment failure at centre of main girder Fig. 6 shows the assumed temperature distribution and positive bending moment and shearing force interaction over the depth of the composite girder.

Referring to Fig. 6, when the positive bending moment and shearing force are acting simultaneously on the composite girder, the plastic neutral axis is located within the concrete slab and the distance of the plastic neutral axis from the upper surface of the concrete slab x is obtained from Eq. (3): x=

(

)

σ y κ yu,θ A u + κ yC,θ A C + κ yw,θ σ y,Q A w u

C

w

0.85σ ck b

(3)

where: κ yu,θ reduction factor of the upper steel flange for yield u strength κ yw,θ reduction factor of the web plate for yield strength w κ yC,θ reduction factor of the lower steel flange for yield C strength sectional area of upper steel flange (mm2) Au sectional area of web plate (mm2) Aw sectional area of lower steel flange (mm2) AC compressive strength of concrete (N/mm2) Xck b effective width of concrete slab (mm) ⎛ Q ⎞ and σ y,Q = σ y 1 − ⎜ θ ⎟ ⎝ Q pC ⎠ As a consequence, the full bending plastic moment MpC,Q θ+ taking into account the shearing force subjected to a non-uniform temperature distribution can be expressed as follows:

Steel Construction 10 (2017), No. 1

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N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

Fig. 6. Positive bending moment and shearing force interaction M-Q and temperature distribution

(

)

MpC,Q θ+ = σ y κ yu,θ A ud1 + κ yC,θ A Cd 2 + κ yw,θ σ y,Q A wd 3 (4) u

C

w

where: d1 " hc – x/2 tu/2 (mm) d2 " hc – x/2 hs – tC/2 (mm) d3 " hc – x/2 hw/2 tu (mm) Fig. 7 shows how the positive bending moment capacity depends on the temperature. Fig. 7 reveals that 81 % of the total positive bending moment capacity at normal temperature remains at 500 °C, which decreases to 26 % at 700 °C. Interpolation shows that the positive bending moment capacity equals the bending moment due to dead load at a temperature of 720 °C. 4) Negative bending moment failure at intermediate support of side girder Accounting for negative bending moment and shearing force interaction in the composite girder, the distance of the plastic neutral axis from the upper flange x is obtained from Eq. (5):

x=

(

NθC − Nθu + Nθw − σ ry A ru + A rC 2κ yw,θ σ y,Q t w

)

(5)

w

where: Nθu = κ yu,θ σ y A u u

θ

N w = κ yw ,θw σ y,Qh w t w NθC = κ yC,θ σ y A C C

Xry Aru ArC hw tw

yield strength of reinforcing bar sectional area of upper reinforcing bar sectional area of lower reinforcing bar depth of steel web plate thickness of steel web plate

The full bending plastic moment MpC,Q θ− taking into account the shearing force can be expressed as follows:

(

)

MpC,Q θ− = σ ry A ruu1 + A rCu 2 + Nθuu 3 + NθC C1

{

(

)

(6)

}

+κ yw,θ σ y,Q t w xu 4 + h w − x t wC 2 w

where: u1 " hc – c tu x (mm) u2 " c tu x (mm) u3 " tu / 2 x (mm) u4 " x / 2 (mm) C1 " hw – x tC / 2 (mm) C2 " (hw – x) / 2 Fig. 9 plots the negative bending moment capacity over the temperature and provides the critical temperature initiating the failure mode at 800 °C.

5 Time to failure

Fig. 7. Positive bending moment capacity depending on temperature

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Steel Construction 10 (2017), No. 1

Four failure modes have been identified for the overpass and the remaining task is to find out which of the failure modes is most critical and most likely caused the failure of the 9-Mile Overpass. The approach we took to solve this was to calculate the duration of fire exposure required to

N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

Fig. 8. Negative bending moment and shearing force interaction M-Q and temperature distribution over depth of composite girder

Fig. 9. Negative bending moment capacity depending on temperature

Fig. 10. Fire curves [16]

reach the critical temperatures in the respective structural member, as found in section 4.

The authors carried out the heat transfer analysis by using the SOFiSTiK finite element software [15]. The thermal properties of the carbon steel and normal concrete are based on Eurocodes 2 and 3 [12], [13]. The temperature– time curves provided by the Eurocode are standard temperature–time curve (ISO), external fire curve (EX) and hydrocarbon curve (HC), as cross-plotted in Fig. 10 [16].

Whereas the standard temperature–time curve (ISO) is used for the fire analysis of buildings, the external fire curve (EX) is used for the fire response calculation of members exposed to fire through windows. The hydrocarbon curve (HC) is used to represent the combustion of gasoline resulting in temperatures of up to 1100 °C, which corresponds well to the forensic reports of bridge failures, e.g. [2]. Hence, in this analysis, the HC curve was used.

Table 3 shows the duration of fire exposure required to reach the respective temperatures and initiating respective

failure modes. Column 1 in Table 3 lists the four failure modes discussed, column 2 shows the critical temperature in °C and column 3 lists the duration of fire exposure up to failure in minutes.

Referring to Table 3, the 9-Mile Road Overpass failed due to shearing failure at the intermediate support. This result is supported by a photo of the post-fire condition of the bridge, see Fig. 11.

6 Future topics for discussion Fire protection mitigation measures have been developed, especially for buildings, but not for bridges [18]. Most bridges do not have passive fire protection materials that can withstand the elevated temperatures exceeding 1100 °C accompanying a fire pool. Governmental agencies around the world, such as the US Department of Homeland Security, are interested in mitigation measures for fire damage and loss of structural integrity of bridges when subjected to

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N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

Table 3. Time of fire exposure up to failure Collapse temperature [°C]

Time to failure [min]

1) Tensile normal force of suspender

830

20.4

2) Shearing force at intermediate support

820

9.5

3) Positive bending moment at centre of main girder

720

16.2

4) Negative bending moment at intermediate support

800

17.7

Fig. 12. Structural detailing of fire protection panel (FFP) connected to I-girder bridge Fig. 11. After the fire accident of the 9-Mile Road Overpass [17]

a severe fire [11]. In order to avoid serious damage to a bridge due to fire attack, we are proposing that fire protection panels (FPP) be added to protect viaducts or bridges at locations where there is a high probability of fire exposure following an accident [19], [20].

FFPs cover a steel frame structure that is connected to the bottom flanges of the I-girders, providing an additional vertical space of 600 mm, as shown in Fig. 12. Please note that the air volume within the additional 600 mm depth provides fire insulation additional to the FFP. The weight of the FFP mitigation construction is about 60 kg/m2.

7 Conclusions Fire safety in bridge design is not as developed as fire safety in building design, even though a bridge failure can cause significant economic damage and impact on an area. This research introduces a methodology capable of identifying the governing failure mode of a bridge structure exposed to severe fire loading and defining the regions of the bridge facing the greatest fire exposure risk. Hence, this proposed methodology is also able to support forensic work identifying the failure mode of a bridge that has failed due to a severe fire.

The proposed methodology is a multi-step procedure.

Firstly, the critical structural members or cross-sections of the bridge structure are identified. Failure of these critical structural members or cross-sections would trigger a specific failure mode, e.g. shear or bending failure mode. The stress resultant due to dead load acting on the critical structural member or cross-section identified is then calculated. Further, the load-carrying capacity of the critical member or cross-section identified is calculated and the

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Steel Construction 10 (2017), No. 1

decay of this capacity taking into account the temperature-dependent reduction in strength is determined. The critical temperature at which the capacity becomes equal to the stress resultant due to dead load is found by interpolation. In a final step, the duration of fire exposure of the respective critical member or cross-section taken to reach that critical temperature becomes the parameter that enables the criticality of members and cross-sections of the bridge to be identified and ranked. It is emphasized that the fire signature, pool fire location and size affect the identification of critical structural members and cross-sections and should be known beforehand. In this study, which investigated the collapse of the 9-Mile Road Overpass, the fire loading was assumed to be uniformly distributed along the whole bridge span, which was confirmed by photographs taken during the fire incident and by reports.

In an effort to mitigate fire damage, the fire protection panel (FFP) is introduced, which is part of a sacrificial structure shielding the bridge superstructure from exposure to fire from underneath.

References [1] Ham, D. B.; Lockwood, S.: National Needs Assessment for Ensuring Transportation Infrastructure Security. American Association of State Highway & Transportation Officials (AASHTO), Washington, D.C., 2002.

[2] Kodur, V. K.; Gu, L.; Garlock, M. E.: Review and Assessment of Fire Hazard in Bridges. Transportation Research Record 2172, TRB, Washington, D.C., 2010.

[3] Roberts, J. E.; Kulicki, J. M.; Beranek, D. A.: Recommendations for Bridge and Tunnel Security. Report FHWA-IF-03-036 (FHWA)/AASHTO Blue Ribbon Panel, 2003.

[4] National Fire Protection Association (NFPA): NFPA 502 – Standard for Road Tunnels, Bridges, and Other Limited Access Highways, 2011 ed., Quincy, MA.

N. Yanagisawa/Y. Imagawa/O. Ohyama/M. P. Rutner/A. Kurita · Fire safety of bridges – methodology supporting design and forensic evaluation

[5] Beard, A.; Carvel, R.: The Handbook of Tunnel Fire Safety, Thomas Telford Press, London, 2005.

[6] National Cooperative Highway Research Program (NCHRP): Design Fires in Road Tunnels. NCHRP Synthesis 415, Washington, D.C., 2011.

[7] Astaneh-Asl, A.; Noble, C. R.; Son, J.; Wemhoff, A. P.; Thomas, M. P.; McMichael, L. D.: Fire Protection of Steel Bridges and the Case of the MacArthur Maze Fire Collapse.

TCLEE 2009: Lifeline Earthquake Engineering in a MultiHazard Environment, Oakland, CA, 28 June – 1 July 2009.

[8] Dunn, D. S.; Chowdhury, A. H.: Analysis of Structural Materials Exposed to a Severe Fire Environment. United States Nuclear Regulatory Commission (USNRC), Report NUREG/ CR-6987, Washington, D.C., 2009.

[9] Hedden, J.; Quagliata, M., Wandzilak, T.: Emergency Renovation, Steel Bridge NEWS. Modern Steel Construction, pp.

36–39, Modern Steel Bridge Alliance, Sept 2010.

[10] Kuwano, T.; Matsui, T.; Suzuki, H.; Yoda, K.: Restoration of Urban Expressway Viaducts Damaged by Severe Fire Accident. Bridge and Foundation Engineering, vol. 43, No. 4, pp.

13–18, Apr 2009 (in Japanese).

[11] Davidson, M.: Assessment of Passive Fire Protection on Steel-Girder Bridges, Western Kentucky University, Masters Theses and Specialist Projects, Paper 1213, 2012.

[12] CEN: Eurocode 3 – Design of steel structures – Part 1-2: General rules-Structural fire design, prEN 1993-1-2, 2003.

[13] CEN: Eurocode 2 – Design of concrete structures – Part 1-2: General rules-Structural fire design, prEN 1992-1-2, 2002.

[14] Bajwa, C. S.; Easton, E. R.; Adkins, H.; Cuta, J.; Klymyshyn, N.; Suffield, S.: The MacArthur Maze Fire and Roadway Collapse: Consequences for SNF Transportation, WM2012 Conference, Phoenix, AZ, 26 Feb – 1 Mar 2012.

[15] SOFISTIK Finite Elemente Software, http://www.sofistik. com/en/ [16] CEN: Eurocode 1 – Actions on structures – Part 1-2: General actions – Actions on structures exposed to fire, EN 19911-2, 2002.

[17] Photos of I-75 and Nine Mile Overpass explosion taken from the following website: http://www.gussysews. com/2009/07/i-75-and-nine-mile-overpass-explosion/ [18] Payá-Zaforteza, I.; Garlock, M. E.: A 3D Numerical Analysis of a Typical Steel Highway Overpass Bridge Under a Hydrocarbon Fire. Structures in Fire, Proc. of 6th Intl. Conf., 2010.

[19] Imagawa, Y.; Ohyama, O.; Kurita, A.: Design of Fire Protection for Steel Girder Bridges, IABSE Symposium BANGKOK 2009.

[20] Yanagisawa, N.; Echigo, S.; Ohyama, O.; Kurita, A.: Fire Protection Panel for Bridges. Proc. of 18th Congress of IABSE, Seoul, Korea, 19–21 Sept 2012, 7A-7, pp. 1–8.

Keywords: fire safety; design; failure mechanism; forensics; progressive collapse; mitigation

Authors: Dr. Eng. Norifumi Yanagisawa Kawada Construction Co., Ltd. 6-3-1 AK Building, Takinokawa, Kita-ku Tokyo, 114-8505 Japan Dr. Eng. Yusuke Imagawa Fuji Engineering Co., Ltd. 5-5-28, Higashimikuni, Yodogawa-ku Osaka, 532-0002 Japan Prof. Dr. Eng. Osamu Ohyama Osaka Institute of Technology Department of Civil Engineering and Urban Design 5-16-1, Ohmiya, Asahi-ku Osaka, 535-8585 Japan Dr.-Ing. habil. Marcus Rutner, Associate Professor Stevens Institute of Technology Schaefer School of Engineering & Science Department of Civil, Environmental and Ocean Engineering 1 Castle Point on Hudson Hoboken, NJ 07030-5991 USA Prof. Dr. Eng. Akimitsu Kurita Yawata Engineering Laboratory Osaka Institute of Technology Ichinotani 4, Minoyama, Yawatashi Kyoto, 614-8289 Japan

Erratum Lindner, J.; Kuhlmann, U.; Just, A.: Verification of flexural buckling according to Eurocode 3 part 1-1 using bow imperfections. Steel Construction – Design and Research 9 (2016), No. 4, pp. 349– 362. DOI: 10.1002/stco.201600004 In the aforementioned article, Fig. 4 unfortunately was published incorrectly.

This is the correct version of Fig. 4: http://onlinelibrary.wiley.com/doi/ 10.1002/stco.201600004/epdf

Steel Construction 10 (2017), No. 1

9

Articles Gerhard Hanswille* Marco Bergmann Reinhard Bergmann

DOI: 10.1002/stco.201710004

Design of composite columns with cross-sections not covered by Eurocode 4 Dedicated to Prof. Dr. Akimitsu Kurita on his 70th birthday

The use of composite columns with special types of sections e.g. concrete filled tubes with massive inner cores became very popular in the last years. These sections are not in the scope of the simplified design method of Eurocode 4. The paper deals with a design proposal for columns with those sections based on the general design method and it shows how this method can also be used extending the scope of the simplified design method in Eurocode 4.

1 Introduction Eurocode 4 for the design of composite structures of steel and concrete [1] gives two methods for the design of composite columns: a simplified design method and a more sophisticated general design method. The simplified method works for the design of composite columns and composite compression members with concrete-encased sections, partially encased sections and concrete-filled rectangular and circular tubes (see Fig. 1) and is based on the full plastic resistance of the cross-section.

In practice, many cross-sections often do not fulfil the conditions for the use of the simplified design method because they have no doubly symmetrical cross-sections or the structural steel components are unconnected. But even in the case of doubly symmetrical cross-sections, the simplified method often cannot be used because the cross-sections have an unfavourable geometry that needs very high strains to develop the full plastic resistance, or they have steel sections with high residual stresses. For these cross-sections, the design has to be performed on the basis of the general method in Eurocode 4 [1]. This method considers a physically and geometrically non-linear behaviour of the compression member. In the present version of Eurocode 4, no clear guidance is given for this method regarding the verification of sufficient safety, because the partial factor method cannot be used in combination with a non-linear calculation.

This paper gives some background information to the application of the general method in Eurocode 4 based on a method given in the German National Application Document for Eurocode 4 [2]. There is also some more guidance on the design of composite columns that cannot be

* Corresponding author: hanswill@uni-wuppertal.de

10

designed following the simplified method even if the cross-section dimensions comply with the design method.

2 Design methods for composite columns in Eurocode 4 2.1 General method The general design method can be used for all types of cross-section without any limitations regarding the shape of the steel or concrete section. This design method takes account of second-order effects, including residual stresses, geometrical imperfections, local instability, cracking of concrete, creep and shrinkage of concrete and yielding of structural steel and reinforcement. It has to be verified that instability does not occur for the most unfavourable combination of actions at the ultimate limit state and that the resistance of individual cross-sections subjected to bending, longitudinal force and shear is not exceeded. The internal forces should be determined by elasto-plastic analysis. Shear lag effects may be neglected and full composite action up to failure may be assumed between the steel and concrete components of the member.

Eurocode 4 gives no detailed rules for the verification of the column in the case of a non-linear calculation. Refs.

[2] and [3] present a method for determining the design value of the resistance. According to [2] and [3], an overall safety LR has to be determined which depends on the relation between the normal force NEd and the bending moment acting MEd (Fig. 2).

The global safety factor LR can be determined using the full plastic cross-section interaction curve of the sec-

Fig. 1. Typical cross-sections for the design of composite columns according to the simplified method

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

G. Hanswille/M. Bergmann/R. Bergmann · Design of composite columns with cross-sections not covered by Eurocode 4

Fig. 2. Safety concept for non-linear design of composite columns [2], [3]

tion. As shown in Fig. 2, it is necessary to determine the interaction curves for Rpl,m using the mean fi,m or nominal value fi,R of the material strength as well as the curve for the design value Rpl,d using the design values according to Eurocode 4. The partial factors 1.5 for concrete, 1.1 for structural steel and 1.15 for reinforcement have to be used for the interaction curve based on design values. For a given combination of internal forces NEd and MEd, the safety factor LR is then given by the ratio of the vectors Rpl,m and Rpl,d.

In a second step, an incremental finite element analysis has to be performed using an initial geometrical bow imperfection and considering residual stresses. It has to be verified that the amplification factor Mu related to the design action effects is greater than the global factor LR according to Fig. 2. The verification has to be performed for the relevant critical cross-section. The stress-strain relation based on the nominal or mean strength values fc,R " fc,m and fs,R " fs,m for concrete and reinforcing steel respectively should be used in the analysis. When calculating LR according to Fig. 2, for simplicity, fc,R may be assumed as being equal to fck for concrete up to strength class C50/60.

For determining the mean yield strength of structural steel, fy,R and fs,R may be taken as equal to fyk and fsk respectively.

It should be mentioned that in Eurocode 4-1-1 the structural steel contribution ratio I has to be in the range between 0.2 and 0.9 for the general and the simplified methods. This limitation was originally introduced for the simplified method only in order to distinguish composite columns from steel and concrete columns. Using the general method, the I limitation is unnecessary.

2.2 Simplified design method The scope of the simplified method is limited to members with a doubly symmetrical and uniform cross-section over

the member length with rolled, cold-formed or welded steel sections. The method is not applicable if the structural steel component consists of two or more unconnected sections. Furthermore, the relative slenderness λ is limited to 2.0 and the structural steel contribution ratio I has to be in the range 0.2 f I f 0.9. In addition, the method applies to columns and compression members with steel grades S235 to S460 and normal-weight concrete of strength classes C20/25 to C50/60.

The resistance to bending can be determined on the basis of the full plastic interaction curve according to Fig. 3, where factor FM mainly takes into account the effects of the reduction in the full plastic resistance to bending due to the strain limitations for concrete.

The internal forces have to be determined based on second-order theory using an equivalent elastic flexural stiffness of the cross-section (see Fig. 3). The influence of geometrical and structural imperfections (residual stresses due to welding or rolling) is taken into account by equivalent geometrical bow imperfections according to Fig. 4.

Fig. 5 shows a comparison between the resistance determined by the general and the simplified methods for concrete-filled tubes with different relative slenderness values. The figure illustrates that the simplified method leads to results that are in very good agreement with the results of the general method.

3 Typical applications for general method 3.1 General The development of the simplified method in Eurocode 4 was based on columns with cross-sections according to Fig. 1. If the scope of the method is to be extended further to other types of doubly symmetrical cross-sections, some additional aspects should be considered. There are mainly two reasons why the simplified method – even in case of doubly symmetrical cross sections – is not always applica-

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G. Hanswille/M. Bergmann/R. Bergmann · Design of composite columns with cross-sections not covered by Eurocode 4

Fig. 3. Simplified design method in Eurocode 4 [1]

Fig. 4. Equivalent geometrical bow imperfections for composite columns

Fig. 5. Comparison of general and simplified design methods to Eurocode 4

ble. The first is a poor plastic shape factor of the cross-section and the second the high residual stresses due to rolling or welding which are not covered by the cross-sections given in Eurocode 4.

A typical example for a column with a very high plastic shape factor FM are the main columns for the arch structures above the east-west glass roof of the central railway station in Berlin, which form arches across the tracks.

The cross-section (see Fig. 7) consists of two welded box sections nested inside each other. The inner box is made of steel grade S355 and the outer box is of S235. In the case of bending, the inner box especially needs excessive strains in the extreme fibre of the outer box in order to develop the full plastic bending capacity. The reduction factor for the use of the simplified method is illustrated in Fig. 7. The determination is based on the general method, described in section 2.1. It can be seen that in this case the columns follow the condition of the simplified method in Eurocode 4 regarding the double symmetry, and normally the design could be performed based on buckling curve b for welded concrete-filled box sections. The comparison with the results based on the general method shows that buckling curve c has to be used in this special case.

Fig. 8 shows the interaction curves of some concrete-filled tubes with and without inner solid core sections for different concrete classes. The figure demonstrates that for normal-strength concrete, the difference

3.2 Columns with an unfavourable cross-section with a high plastic shape factor In the simplified method according to Fig. 3, the factor FM for the determination of MRd takes into account the difference in the resistance of the cross-section in the case of a full plastic calculation using stress blocks or a calculation with strain limitation using the stress-strain relation of the materials. The factors 0.9 for steel grades S235 and S355 and 0.8 for steel grade S460 were developed for the cross-sections according to Fig. 1. If the plastic shape factor is more unfavourable, different FM factors have to be used [4]. The problem is illustrated in Fig. 6 for a circular concrete section with an inner solid steel core. In the extreme fibre of the concrete section, excessive strains are necessary in order to develop the plastic resistance in the structural steel core.

The cross-section can be classified as a cross-section with a high plastic shape factor FM " Mpl,Rd/MRd.

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G. Hanswille/M. Bergmann/R. Bergmann · Design of composite columns with cross-sections not covered by Eurocode 4

Fig. 7. Columns in the arch structure at the central station in Berlin

Fig. 6. Full plastic bending resistance and bending resistance in the case of strain limitation

between the plastic interaction curve and the interaction relation regarding strain limitations of the materials is quite small and can be covered by the factor FM " 0.9 to Eurocode 4. Even for the cross-section with normal-grade concrete and an inner solid core section, the difference does not increase significantly. But when high-strength concrete is used, the difference becomes much less favourable. This is the main reason behind limiting the concrete strength to C50/60 in Eurocode 4.

Fig. 9 shows a cross-section consisting of a circular concrete section with an inner I-section. There is no ex-

plicit reference to this type of cross-section in Eurocode 4-1-1. Comparing the reduction factors determined with the general method with buckling curves shows that this type of cross-section can be classified under buckling curve b in the case of a centrically loaded column. This is the same classification as for concrete-filled tubes. Nevertheless, for combined compression and bending, the simplified method leads to unsafe results even when using buckling curve b in combination with factor FM " 0.9 according to Eurocode 4.

This is shown in Fig. 10, where a further reduction in the factor FM is necessary to obtain safe results for this cross-section if the design for columns in combined compression and uniaxial bending is based on buckling curve b for circular cross-sections.

3.3 Examples of columns with high residual stresses due to welding or cooling of solid sections In recent years, concrete-filled tubes with solid inner core sections or additional, very thick inner welded plates have very often been used for high-rise buildings. Typical examples are shown in Fig. 11.

Fig. 8. Examples of interaction curves for columns with concrete-filled tubes

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G. Hanswille/M. Bergmann/R. Bergmann · Design of composite columns with cross-sections not covered by Eurocode 4

Fig. 9. Circular concrete section with an inner steel section and centrically loaded

Although this type of cross-section formally complies with the conditions of the simplified method in Eurocode 4-1-1, such cross-sections cannot be designed on the basis of the simplified method. The simplified method was developed assuming a geometrical imperfection of L/1000 and typical distributions of residual stresses [5], [6], [7], [8] for welded and rolled I-sections according to Eurocode 3 for steel members in compression. Where the cross-sections used have a less favourable distribution of residual stresses, the simplified method can lead to unsafe results. This can be the case, for example, with cross-sections with concrete-filled tubes and solid inner core sections according to Fig. 11.

A design method for concrete-filled tubes with cylindrical core sections was developed in [9], [10]. The main problem was the realistic determination of the distribution of the residual stresses in the inner core sections. In this case, residual stresses arise during production of these sections due to the uneven temperature distribution while the steel bar is cooling. The surface areas cool and harden faster as cooling proceeds. Because the inner parts cool later, compression stresses occur in the already hardened

Fig. 10. Circular concrete section with an inner steel section in bending and compression

part near the surface and tension stresses in the inner parts of the cross-section.

In composite columns these residual stresses cause a reduction in flexural stiffness and a reduction in the ultimate load due to earlier yielding of the structural steel and increasing yielding zones in the structural steel section. In [9] it is shown that the residual stresses due to cooling of circular core sections can be determined with the approximation according to Fig. 12.

With the residual stresses according to Fig. 12 and a bow imperfection of L/1000, concrete-filled circular tubes with additional inner solid core sections were investigated [9] in order to prove the design concept according to section 2.1. In a first step, the results of 12 column tests were evaluated according to EN 1990, Annex D, Design assisted by testing [11] (red squares in Fig. 13). In a second step, the ultimate loads were determined with the general method in Eurocode 4-1-1 and the safety concept according to section 2.1 (blue circles in Fig. 13.). The comparison of the two results shows that the proposed safety concept [2], [3] leads to safe results.

Fig. 11. Examples of columns with solid welded or rolled inner core sections

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G. Hanswille/M. Bergmann/R. Bergmann · Design of composite columns with cross-sections not covered by Eurocode 4

Table 2. Geometrical bow imperfection for concrete-filled tubes with additional inner core sections, coefficients kd and kQ f 200 mm

kd = 1 +

dK [mm] 400

# 200 mm

kd = 2 −

dK [mm] 400

Diameter dk of core

Related slenderness

Fig. 12. Distribution of residual stresses due to cooling during production

λK =

Npl,k Ncr

λ K ≤ 0.5

kQ " 0.8

λ K > 0.5

k λ = 0.7 + 0.2 λ K

4 Summary In recent years, composite columns with cross-sections which are not covered by the simplified design method in Eurocode 4 have been especially popular for high-rise buildings. In this case the design has to be performed on the basis of the general design method in Eurocode 4.

When using this method, the present version of Eurocode 4 [1] gives no guidance on the verification of the required level of safety. Based on the evaluation of column tests in [2], a procedure is given which allows the verification of the sufficient safety of composite members when using a geometrical and physical non-linear calculation. It is shown by means of examples, however, that this method allows the scope of the simplified method in Eurocode 4 to be extended to doubly symmetrical cross-sections not covered by the Eurocode.

Fig. 13. Evaluation of column tests and comparison with the general method

References

It is shown in [9], [10] that, based on the results of the general method, the scope of the simplified method in Eurocode 4 can be extended. The research shows that the simplified method according to Figs. 3 and 4 can be used for concrete-filled tubes with solid inner core sections in S235 or S355 if the values for factor FM (see Table 1) and, in addition, the equivalent initial bow imperfection, are modified according to Eqs. (1) and (2), where factors FMo and FN as well as factors kd and kQ are given in Tables 1 and 2. In Table 1, for dk/d values between 0.0 and 0.75, the values FMo and FN may be determined by linear interpolation. α M = α M,o − α N

(1)

L 400

(2)

wo =

1 kd ⋅ k λ

[1] EN 1994-1-1 (Eurocode 4): Design of composite steel and concrete structures – Part 1-1: General rules and rules for buildings, 2010.

[2] DIN EN 1994–1-1/NA: National Annex – Nationally determined parameters – Eurocode 4: Design of composite steel and concrete structures – Part 1-1/NA: General rules and rules for buildings.

[3] Hanswille, G.: Die Bemessung von Stahlverbundstützen nach nationalen und EU-Regeln. Der Prüfingenieur 22 (2003).

[4] Hanswille, G.; Bergmann, R.: New design methods for composite columns including high strength steel. Engineering Foundation Conferences Composite Construction V, South Africa, July 2004.

[5] Roik, K.; Bergmann, R.: Composite Columns, Constructional Steel Design, An International Guide, Elsevier Science Publishers Ltd., London, 1990.

[6] Hanswille, G.; Bergmann, R.: Ermittlung geometrischer Ersatzimperfektionen für Verbundstützen mit hochfesten Stählen. Research project P3-5-17.10-99201, Deutsches Institut für Bautechnik, Berlin, Oct 2001.

Table 1. Values of FMo and FN Steel grade of core section S235 and S355

dK/d " 0

dK/d " 0.75

Concrete grade

FMo

FN

FM0

FN

C30/37

0.90

0.10

0.85

0.15

C50/60

0.90

0.20

0.80

0.15

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G. Hanswille/M. Bergmann/R. Bergmann · Design of composite columns with cross-sections not covered by Eurocode 4

[7] European Convention for constructional steelwork: 2nd Intl. Coll. on stability, Introductory Report, Tokyo/Liege/ Washington, 1977.

[8] Hawranek, R.; Petersen, C.: Sicherheit gedrückter Stahlstützen, Berichte zur Sicherheitstheorie der Bauwerke, Bo. 8/1975, Sonderforschungsbereich 96, Werner Verlag, Düsseldorf, 1975.

[9] Hanswille, G.; Lippes, M.: Design of composite columns made of concrete filled tubes with inner core profiles and high strength materials, Engineering Foundation Conferences, Composite Construction in Steel and Concrete VI, Colorado, USA, 2008.

[10] Hanswille, G.; Lippes, M.: Einsatz von hochfesten Stählen und Betonen bei Hohlprofil-Verbundstützen. Stahlbau 77 (2008), No. 4. DOI: 10.1002/stab.200810041 [11] EN 1990: Eurocode – Basis of structural design, 2002.

Keywords: composite columns; general design method; safety concept for non-linear design methods

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Steel Construction 10 (2017), No. 1

Authors Prof. Dr.-lng. Gerhard Hanswille University of Wuppertal Faculty for Architecture & Civil Engineering Institute for Steel & Composite Structures Pauluskirchstr. 11 42289 Wuppertal Germany Dr.-lng. Reinhard Bergmann Consulting Engineer Himmelohstr. 127 58454 Witten Germany Dr.-lng. Marco Bergmann HRA Consulting Engineers Kohlenstr. 38 44795 Bochum Germany

Articles Martin Mensinger* Li Huang

DOI: 10.1002/stco.201710005

Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer Comparing optimizations to satisfy different objectives Preliminary structural design is used for estimating and comparing design alternatives. It is mainly a manual procedure based on experience and documented in the form of design tables or charts. In an effort to automate this procedure for steel composite office buildings, the Sustainable Office Designer (SOD) has been developed as a SketchUp plug-in. It can generate optimized preliminary structural designs for given rectangular boundary shapes and employs a rapid approximate calculation instead of an expensive structural analysis. Steel columns are used and verified according to DIN EN 1993-1-1, while beams are designed as composite members and verified to DIN EN 1994-1-1. Pre-calculated design tables are used for slab systems. Optimization results are obtained and compared for different objectives, e.g. life cycle assessments (LCA) and costs.

2 Sustainable Office Designer (SOD) Fig. 1 shows the algorithm of SOD, which uses a parametric model and a genetic algorithm to automate the preliminary structural design procedure. The current implementation of SOD is able to generate structures for buildings with a rectangular shape. SOD optimizes both the geometry of the structure, such as the orientation of primary beams and the number of secondary beams, and the selection of structural elements. The whole building is represented by a set of parameters and optimized through an evolutionary approach. At the start, a large number of different parameter sets are generated randomly. Using the

1 Introduction In a building project preliminary structural design is used in the planning phase for estimating and comparing design alternatives [1]. It is mainly a manual procedure based on the engineer’s experience and makes use of available design tables and charts. Although it is a rapid procedure in comparison to detailed design, as approximate analyses and simplified methods are used, it is still not fast enough when many different design options are involved. It also has the potential to be automated. So SOD [2] has been developed which generates preliminary structural designs automatically for rectangular low-rise steel composite office buildings. Since sustainability has become one of the important objectives for a building project, it is considered in SOD, too.

Using SOD, it is possible to obtain a large number of preliminary structural design models for different design alternatives very quickly. Some examples with different geometries are performed to learn from the outcome, also with regard to different objectives, such as life cycle assessments (LCA) and the costs involved. Since LCA and costs are calculated mainly based on the consumption of materials, it can be expected that the two objectives lead to the same or a similar optimized structure. The result presented in Table 2 is consistent with our prediction, i.e. in most cases, exactly the same results are obtained for both objectives. * Corresponding author: mensinger@tum.de

Fig. 1. Optimization algorithm in SOD [3]

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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M. Mensinger/L. Huang · Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer

objective function and penalty value obtained from structural verification, these parameter sets can be compared with each other. Similarly to the natural evolution procedure, they go through a selection, crossover and mutation process, striving towards a better design from generation to generation.

2.1 Preliminary structural design In the optimization procedure a penalty value is applied to compare the structural performance. It is calculated from structural verification and indicates the validation of design rules, i.e. stress limits and deflections, and geometric requirements, e.g. maximum depth limits can be defined for the structural elements. As a preliminary design in our implementation, only vertical loads are considered, i.e. dead loads and live loads. The values are applied to office buildings according to DIN EN 1991-1-1 [4]. Owing to the complexity and only small influence on the final estimate, bracing is not considered. Structural elements, i.e. steel columns and composite beams, are designed and verified according to DIN EN 1993-1-1 [5] and DIN EN 1994-1-1 [6].

In the structural model, slabs are supported directly by secondary beams, which are then mounted on the primary beams at an even spacing. To meet construction practice requirements, all secondary beams of one rectangular floor layout as well as all primary beams use the same steel section. To be consistent, only one type and thickness of slab system is used over the whole building, which may be composed of multiple rectangles. For each floor a maximum of two different steel sections are used for the columns, i.e. one for external columns and one for internal columns. Office buildings with a more flexible layout, with a width in the range 12–18 m, may lead to a more sustainable design [3], [7]. To cover more design options in SOD the range 9–21 m is used, which is an expansion of the range 12–18 m. For a layout with a width of this range, 9–21 m, the maximal number of rows is set to two. The actual optimized number of rows, which can be zero, one or two, and positions are obtained using SOD. Primary beams are designed as continuous composite beams, while a plastic structural verification is performed on the primary beams considering cracking of the concrete at inner supports. Propped and pre-cambered construction is considered in all examples.

2.2 Objective functions In the optimization algorithm, an objective function is used to calculate and compare the actual performance of different structural solutions. Two objective functions Table 1. Options for structural elements Structural element

Options

Primary beam

IPE, HEA, HEB, HEM

18

Secondary beam

IPE, HEA

Column

HEA, HEB, HEM

Slab

Cofraplus 60, h " 120, 140, 160, 180 mm; t " 0.75, 0.88, 1.00, 1.25 mm

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are used, one for an environmental aim and the other for an economics aim. Data available from Ökobau.dat and several German environmental product declarations (EPDs) are used [7], [8], [9] to calculate the first objective value, which we call the EPD value. As shown in Eq. (1), it is a weighted summation of different environmental criteria. The detailed calculation is explained in [8], where the prices of different construction materials are also provided. The prices are used in Eq. (2) to calculate the cost.

J = 3 × GWP + 1 × ODP + 1 × POCP + 1 × AP + 1 × EP + 3 × PEne + 2 × PEe

(1)

where: GWP Global Warming Potential [kg CO2 eqv.]

ODP Ozone Depletion Potential [kg R11 eqv.]

POCP Photochemical Ozone Creation Potential [kg C2H4 eqv.]

AP Acidification Potential [kg SO2 eqv.]

EP Eutrophication Potential [kg PO43– eqv.]

PEne Primary Energy (non-renewable) [MJ] PEe Primary Energy (renewable) [MJ] J = Pc × Vc + Pr × Mr + Pp × Mp + Ps × Ms where: Vc Mr Mp Ms and

(2)

volume of concrete mass of reinforcement mass of profiled sheeting mass of section steel Pc, Pr, Pp and Ps are the corresponding prices

As given in [8], the prices are taken as Pc " 121 €/m3 for C30/37 concrete, Pr " 580 €/t, Pp " 1500 €/t and Ps " 945 €/t for S355 steel.

3 Parameters and examples To learn about the results of optimization with respect to different objectives, a series of examples is set up. To limit the possibilities, only composite slab Cofraplus 60 [10] is used with four different slab thicknesses, 120, 140, 160 and 180 mm, and four different sheeting thicknesses, 0.75, 0.88, 1.00 and 1.25 mm. Sections for secondary beams are selected from IPE and HEA, while HEA, HEB and HEM are used for the columns. For primary beams, four types of section can be chosen, IPE, HEA, HEB and HEM, as shown in Table 1. Concrete C30/C37 and steel grade S355 are selected as the only materials for the concrete and the steel sections.

Five floor layouts are selected for the examples as shown in Fig. 2. For each of the first three floor layouts, two examples are processed – one with two floors and the other with seven. For these three layouts with a 1.2 m façade width, the positions of inner columns are suggested in [7], i.e. 4.8 m to the outer boundary. So the allowable column zone [2] is defined as [4.4–5.2 m] to the boundary with 0.4 m variability to each side. For a layout with a 1.35 m façade width, the suggested inner column row is 4.05 m from the boundary. This applies to the last

M. Mensinger/L. Huang · Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer

Table 2. Optimization results Minimize EPD

Minimize cost

EPD

1.358478e 6

1.362039e 6

cost

4.072985e 4

3.955880e 4

36 w 9.6 m 7 floors (NO inner row)

EPD

5.006404e 6

5.066147e 6

cost

1.491675e 5

1.456780e 5

36 w 15.6 m 2 floors (1 inner row)

EPD

2.308033e 6

2.308033e 6

cost

6.721963e 4

6.721963e 4

36 w 15.6 m 7 floors (1 inner row)

EPD

8.597854e 6

8.597854e 6

cost

2.489224e 5

2.489224e 5

36 w 20.4 m 2 floors (2 inner rows)

EPD

2.531589e 6

2.531589e 6

cost

7.544848e 4

7.544848e 4

36 w 20.4 m 7 floors (2 inner rows)

EPD

9.346992e 6

9.346992e 6

cost

2.754771e 5

2.754771e 5

36 w 12 m 2 floors (NO inner row)

EPD

1.876847e 6

1.876847e 6

cost

5.429104e 4

5.429104e 4

40.5 w 12.15 m 2 floors (1 inner row)

EPD

1.711258e 6

1.711258e 6

cost

5.286920e 4

5.286920e 4

40.5 w 12.15 m 2 floors (NO inner row)

EPD

2.185108e 6

2.220618e 6

cost

6.528721e 4

6.401134e 4

Index

Layout

1

36 w 9.6 m 2 floors (NO inner row)

2

3

4

5

6

7

8

9

4 Results

Fig. 2. Floor layouts

layout, 40.5 w 12.15 m. Including variability, the range 3.65–4.45 m is used. Two-floor examples are set for the last two layouts, 36 w 12 m and 40.5 w 12.15 m, while the latter has two variants, with or without an inner column row. For the first four layouts, the column spacing can be chosen from 3 w 12 m, 5 w 7.2 m or 6 w 6 m, while for the fifth one the options are 3 w 13.5 m, 5 w 8.1 m or 6 w 6.75 m.

In total, nine examples are processed, while for each example and each objective function, eight optimization procedures are performed and the best of each set of eight is selected as the result of that respective optimization set, as listed in Table 2. Although only one objective value is used for each optimization, both the EPD value and the costs are recorded for analysis.

The best solution for each optimization is given in Table 2, the first one or two governing criteria are presented in Table 3 and the choice of structural members for the optimized structure is given in Table 4. Since the influence of and difference between columns are relatively small, only beams and slabs are included in Table 4. For the 36 w 15.6 m layout with one row of inner columns, 36 w 20.4 m with two inner rows, 36 w 12 m with no inner row and 40.5 w 12.5 m with one inner row, the results show that both objectives lead to the exact same optimized result. For the 36 w 9.6 m layout with no inner row and 40.5 w 12.5 m with no inner row, the results using two different objective functions show a slight difference. The optimized structural design is illustrated in Fig. 3, while the dimensions including the column spacing are specified. It shows that for office buildings, a more economical design is likely to be more environmentally friendly, except those with long secondary beams, and is also very likely governed by the deflection of the primary beam.

5 Conclusion SOD users are able to perform automated preliminary structural designs for composite structures for office buildings. In this paper, the optimization of a set of examples is performed on two different objectives. One is the EPD

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M. Mensinger/L. Huang · Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer

Table 3. Governing criteria; PB: primary beam, SB: secondary beam

36 w 9.6 m 2 floors (NO inner row)

Criterion PB: deflection PB: bending min moment, EPD construction state SB: bending min moment, final state cost PB: deflection PB: deflection

36 w 9.6 m 7 floors (NO inner row)

36 w 15.6 m 2 floors (1 inner row)

36 w 15.6 m 7 floors (1 inner row)

36 w 20.4 m 2 floors (2 inner rows)

36 w 20.4 m 7 floors (2 inner rows)

36 w 12 m 2 floors (NO inner row)

min EPD

min cost

Secondary beam

Slab

min EPD

IPE300

IPE240

140 mm 1 mm

min cost

IPE360

IPE270

120 mm 1 mm

min EPD

IPE300

IPE240

140 mm 1 mm

min cost

IPE330

IPE240

120 mm 1 mm

36 w 15.6 m 2 floors (1 inner row)

IPE400

IPE270

120 mm 1 mm

36 w 15.6 m 7 floors (1 inner row)

IPE400

IPE270

120 mm 1 mm

1.79e 8!#2.04e 8

36 w 20.4 m 2 floors (2 inner rows)

IPE270

IPE300

120 mm 1 mm

3.010e 8!#3.071e 8

36 w 20.4 m 7 floors (2 inner rows)

IPE220

IPE240

120 mm 1 mm

2.523e 8 !#2.562e 8

36 w 12 m 2 floors (NO inner row)

IPE400

IPE330

120 mm 1 mm

40.5 w 12.15 m 2 floors (1 inner row)

IPE450

IPE200

140 mm 1 mm

min EPD

IPE450

IPE330

140 mm 1 mm

min cost

IPE450

IPE300

120 mm 1 mm

1.40e 8!#1.48e 08

2.38e 8!#2.58e 8 15.024!#17.143 17.139 !# 17.143 1.40e 8!#1.48e 08

PB: deflection

15.155!#17.143

SB: bending moment, final state

PB: deflection 40.5 w 12.15 m SB: bending 2 floors moment, final (1 inner row) state SB: bending min moment, final state 40.5 w 12.15 m EPD PB: deflection 2 floors SB: bending (NO inner moment, final row) min state cost PB: deflection

20

Primary beam

Value!#Design value 17.139 !# 17.143

PB: bending moment, construction state SB: bending moment, final state PB: bending moment construction state SB: bending moment, final state PB: bending moment, construction state SB: bending moment, final state PB: bending moment, final state SB: bending moment, final state PB: bending moment, construction state SB: bending moment, final state

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Table 4. Structural members

3.010e 8!#3.071e 8

2.523e 8 !#2.562e 8

36 w 9.6 m 2 floors (NO inner row) 36 w 9.6 m 7 floors (NO inner row)

40.5 w 12.15 m 2 floors (NO inner row)

2.022e 8!#2.121e 8

3.10e 8!#3.22e 8

6.634e 7!#6.707e 7

1.95e 8!#2.04e 8

3.755e 8!#3.875e 8

value, which is the weighted sum of environmental criteria, while the other one is the costs based on the consumption of materials. The optimization against the EPD value and the costs provide the same or very similar results. For a narrow office building without inner columns, the most economical solution is expected to have most likely a slightly inferior environmental performance than the most ecological solution. The optimization also shows that the number of floors has a minimal influence on the structure, i.e. the same floor layouts with different floors tend to have similar structures.

References

21.7!#22 1.51e 8!#1.53e 8

4.141e 8!#4.173e 8 21.61!#23.14 3.1238e 8!#3.1239e 8 21.809!#23.143

[1] MacGinley, T. J.: Steel Structures: Practical design studies, 2nd ed., E & FN Spon, London/New York, 1997.

[2] Huang, L.; Mensinger, M.: Optimization Based Preliminary Structure Generation for Early Stage Design Estimation of Composite Office Buildings, Proc. of EUROSTEEL 2014, Naples.

[3] Mensinger, M.; Huang, L.; Stroetmann, R.; Podgorski, C.; Eisele, J.; Trautmann, B.; Feldmann, M.; Pyschny, D.; Lingnau, V.; Kokot, K.; Zink, K. J.; Baudach, T.: Project Report of P881: Sustainable office and administrative buildings in steel and steel composite construction, Germany, 2016.

M. Mensinger/L. Huang · Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer

Fig. 3. Resulting structures of examples 1–9

[4] DIN EN 1991-1-1 (Eurocode 1): Einwirkungen auf Tragwerke – Teil 1-1: Allgemeine Einwirkungen auf Tragwerke – Wichten, Eigengewicht und Nutzlasten im Hochbau, Dec 2010.

[5] DIN EN 1994-1-1 (Eurocode 4): Bemessung und Konstruktion von Verbundtragwerken aus Stahl und Beton – Teil 1-1: Allgemeine Bemessungsregeln und Anwendungsregeln für den Hochbau, Dec 2010.

[6] DIN EN 1993-1-1 (Eurocode 3): Bemessung und Konstruktion von Stahlbauten – Teil 1-1: Allgemeine Bemessungsregeln und Regeln für den Hochbau, Dec 2010.

[7] Stroetmann, R.; Podgorski, C.; Mensinger, M.; Huang, L.; Eisele, J.; Trautmann, B.; Feldmann, M.; Pyschny, D.; Lingnau, V.; Kokot, K.; Zink, K. J.; Baudach, T.: Ganzheitliche Planung nachhaltiger Bürogebäude in Stahl- und Verbundbauweise. Stahlbau 83 (2014), pp. 429–440. DOI: 10.1002/ stab.201410176

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M. Mensinger/L. Huang · Optimized preliminary structural design of steel composite buildings using the Sustainable Office Designer

[8] Stroetmann, R.; Podgorski, C.: Zur Nachhaltigkeit von Stahl- und Verbundkonstruktionen bei Büro- und Verwaltungsgebäuden – Teil 1: Tragkonstruktionen. Stahlbau 83 (2014), pp. 245–256. DOI: 10.1002/stab.201410143 [9] Ökobau.dat (2013), www.nachhaltigesbauen.de/oekobau dat, accessed Jan 2015.

[10] Stahlbau Zentrum Schweiz: Steelwork C1/11 Verbundbau-Bemessungstafeln Tables de dimensionnement pour la construction mixte, Zurich, 2011.

Keywords: preliminary structural design; Sustainable Office Designer; SOD; structural optimization; EPD; steel composite; automated structural design

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Authors Prof. Dr.-Ing. Dipl.-Wirt.-Ing. Martin Mensinger Lehrstuhl für Metallbau Arcisstr. 21 80333 Munich mensginer@tum.de Li Huang, M.Sc. Lehrstuhl für Metallbau Arcisstr. 21 80333 Munich li.huang@tum.de

Articles Daniel Pak* Hetty Bigelow Markus Feldmann

DOI: 10.1002/stco.201710006

Design of composite bridges with integral abutments Bridges are of vital importance to Europe’s infrastructure and composite bridges have already become a popular solution in many countries, representing a well-established alternative to concrete bridges. Their competitiveness depends on several factors such as site conditions, local costs of materials and labour and the contractor’s experience. One outstanding advantage of composite bridges compared with concrete bridges is that the steel girders can carry the weight of the formwork and the fresh concrete during casting. Another major advantage is savings in construction time, which reduces disturbance to traffic and, consequently, saves money for the contractor, but even more so for road users – a fact that has been neglected for a long time. Recently, this factor has increasingly attracted attention as the latest studies show the need to take into account not only simple production costs, but also construction time and maintenance costs when deciding on a specific bridge type. All these needs are met by integral abutment bridges as well. In addition, this bridge type has the potential to outclass traditional bridges with transition joints as it not only reduces production and maintenance costs, but saves on economic and socio-economic costs as well.

1 Introduction Bridges with integral abutments may be characterized by two advantageous features, see Fig. 1: 1. Structurally, the superstructure is fully fixed with the abutment structure such that end moments are built up which counteract the span moments, which leads to slender cross-sections with a remarkably shallow depth and a decrease in the costs of materials, fabrication, transport and construction.

2. In construction terms, expansion joints are avoided such that a continuous transition from the solid ramp to the superstructure is possible, thus avoiding any durability or inspection problems caused by additional joint details. Furthermore, bearings are eliminated.

permitting an additional reduction in construction depth.

At traffic intersections, the vertical distance between the gradients may be reduced, which minimizes earth-moving operations. Furthermore, the costs of materials, fabrication, transport and construction can be optimized. The absence of the middle support simplifies the construction of the bridge essentially without interfering with the traffic underneath, as the road does not have to be closed. In addition, horizontal loads such as wind, earthquake or braking forces are introduced directly into the subsoil, which eliminates complex anchorage points [1].

2 Historical development of composite bridges with integral abutments Framed steel bridges as we have known them for more than 70 years can be regarded as the forerunners of integral abutment bridges. In the framed bridges, short exterior columns were vertically cantilevered, providing end moments that decrease the span moment, thus reducing the depth of the girder (Fig. 2). In Germany they were very popular in urban surroundings and for crossing motorways, where neither large depths nor intermediate columns could be allowed. However, those bridges required large foundation blocks to resist the horizontal forces in the horizontal foundation gap or on the foundation piles.

Transition joints were still necessary. However, modern solutions for integral abutment bridges tend to use composite sections for both the superstructure and the integration into the abutment foundations.

Nowadays, these composite integral abutment bridges are a well-established design type and already state of the art in the USA [6], [7], [8]. Bridges with integral abutments have been built more frequently in Europe as well within the last decade [2]–[4], [9].

3 Structural systems Generally, both points result in very robust and cost-effective bridge solutions for which the provision of maximum flexibility in the clearance underneath the bridge and minimum lifetime costs for the structure are typical.

As the whole system acts as a frame, the superstructure can be designed to be quite slender, with haunches * Corresponding author: pak@bau.uni-siegen.de

Integral abutment bridges are divided into fully integral and semi-integral structures. Bridges without bearings and joints are classified as integral structures. The loadbearing structure is designed as a whole, with the superstructure monolithically connected to the substructure (piles and abutments), whereas concrete joints are understood as monolithic connections (Fig. 5) [12]. For the concept of semi-integral bridges, there are different definitions in dif-

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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D. Pak/H. Bigelow/M. Feldmann · Design of composite bridges with integral abutments

Fig. 1. Typical composite bridge with integral abutments

Fig. 2. Old framed bridges in Germany dating from the 1930s: a) Wilhelmsbrücke crossing the River Neckar in Stuttgart-Cannstatt, b) crossing over Hardenbergstraße at Zoologischer Garten railway station in Berlin [5]

ferent countries. In the USA as well as in some European countries, bridges are called semi-integral when they have either expansion joints or bearings, but not both. Bearings or expansion joints are arranged solely at the abutments.

In Germany bridges that do not comply with the definition of fully integral bridges and in which the piles are monolithically connected to the superstructure on at least two axes are referred to as semi-integral bridges (Fig. 6) [12].

Integral abutment bridges are generally designed based on two different concepts: 1. Low flexural stiffness of piles / low degree of restraint In the USA especially, abutments and columns are supported by single rows of flexible steel piles. The bridge structure can be considered to be a continuous frame.

As the columns are quite flexible, the continuous superstructure may be assumed to have simple or hinged supports. Consequently, except for the design of the continuity connections at abutments and columns, frame action can be ignored when analysing the superstructure for superimposed dead and live loads [13]. Furthermore, as only low moments need to be conducted through the abutment’s corner, the design of that detail becomes rather simple. The main advantage here is seen to be the absence of bearings and joints.

2. High flexural stiffness of piles / high degree of restraint The more slender the superstructure is intended to be constructed, the stiffer the substructure system has to be [1]. In order to increase the corner moment of the bridge and to decrease the span moment, the horizontal member (the continuous superstructure) is partly restrained

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Steel Construction 10 (2017), No. 1

by the stiff vertical members. This frame concept is widely employed in Europe, calling for stiff concrete foundation piles. For bridges with short and medium spans especially, the main advantage here is seen to be the slender superstructure and the absence of the middle support (Fig. 1).

A typical integral abutment bridge has one span and can be founded on piles or footings (Fig. 3). In the case of long spans especially, piled foundations are preferred for reason of their more flexible horizontal bedding, as constraint forces due to temperature loading and support settlement can be absorbed by flexible structures more effectively [10].

For reasons of aesthetics, but also to optimize visibility for traffic, the abutments are often designed with inclined faces (Fig. 4). Inclining the abutments to the back effectively creates a smaller mid-span moment as the superstructure is dimensioned for span ls2, which results in superstructures that are very slender visually [10].

In the case of multi-span structures, a deep foundation should be provided for the columns and, most importantly, for the abutments (Fig. 5). Alternatively, the abutments can be separated from the superstructure by bearings, resulting in a semi-integral structure (Fig. 6), although this kind of bearing system is less efficient in some cases. For example, in the case of semi-integral bridges, high braking forces from railway traffic can only be absorbed by the piles. In the case of fully integral bridges, braking forces are transferred directly by the pile block supporting the abutment into backfill and subsoil (Fig. 7) [10].

D. Pak/H. Bigelow/M. Feldmann · Design of composite bridges with integral abutments

Fig. 3. Single-span integral abutment bridge [10]

forces (Fig. 8). Deformations between sections can be absorbed without rail expansion joints [10].

4 Special design issues The analysis of an integral bridge differs from that of other bridges because the abutment is rigidly connected to the superstructure. Horizontal forces and displacements are transferred from the superstructure to the abutment and foundation piles. Therefore, the piles, which are interacting with the subsoil, are not only loaded by vertical forces, but also by shear forces and moments as well as forced displacements and rotations. Furthermore, the abutment interacts with the backfill, resulting in an additional restraint in the structure.

Fig. 4. Integral abutment bridge with inclined abutments [10]

Long integral abutment bridges spanning low valleys are generally divided into sections. These sections are provided with central pile trestles to absorb the high braking

4.1 Soil-foundation pile interaction The choice of a sufficient foundation pile system is very much linked with the design concept chosen. In the USA, the UK and Sweden, piles with a low flexural stiffness are preferred for minimizing the flexural effects due to lateral movements and rotations of the abutments. Therefore,

Fig. 5. Multi-span integral abutment bridge [10]

Fig. 6. Multi-span semi-integral abutment bridge [10]

Fig. 7. Multi-span structure deformed by railway braking loads [10]

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D. Pak/H. Bigelow/M. Feldmann · Design of composite bridges with integral abutments

Fig. 8. Section of long railway viaduct [10]

H-piles are the first choice for integral bridges in the USA, especially for longer bridges [13]. Nowadays, they are mostly orientated for weak axis bending to minimize the stresses in the abutments and to make sure that local buckling of the flanges does not occur, even if the soil does not support the pile laterally [14], [15]. Additionally, oversized pre-drilled holes can be used to reduce the horizontal resistance to lateral displacements at the top of the piles.

These holes are filled with a low-stiffness material that surrounds the top of the piles [11]. Examples of loose materials that are used are sand and bentonite slurry [16]; compressible foam has also been used [17]. A study in the USA from the early 1980s showed that only four out of 29 states that were building integral bridges were normally using pre-drilled holes. A newer survey from 2000 shows that 12 out of 30 states are now demanding the use of pre-drilled holes [18]. Based on investigations performed on a pilot bridge built by Ramböll Sverige AB over the River Leduån in Sweden [11], the benefit of pre-drilled holes could be proven within the scope of the European research project INTAB as well.

In Germany piles with a high flexural stiffness are preferred in order to increase the corner moment of the bridge, decrease the span moment and avoid the middle support, with concrete piles being quite common. In general, they can be categorized as precast or cast-in-place piles, with the precast concrete piles being either conventionally reinforced or prestressed. Their quality is controlled by, for example, dynamic loading tests or integrity checks. Piles and pile caps are connected via reinforcement to form a monolithic structure.

Sheet pile walls as foundation members remain a less common alternative. They can be taken into consideration if the sheet pile wall needs to be constructed anyway. In that case the existing sheet pile wall can be used as a piling system for the integral abutment bridge, avoiding additional piling. Owing to the high stiffness of the sheet pile wall, special care needs to be taken during the construction of the abutment-sheet pile connection (Fig. 9). On the one hand, slippage between sheet pile wall and abutment needs to be minimized, e.g. by installing vertical studs welded to the sheet pile wall. On the other hand, particular attention has to be paid to the problem of concrete cracking within the region of the sheet pile wall. Studies have shown that a restraint of 50 % has the most beneficial influence on the choice of rolled beams for the composite superstructure, whereas for higher degrees of restraint, crack width design at the supports becomes problematic when tapered haunches are not used [11].

For design purposes, the soil surrounding the piles can be represented by springs that are applied to the piles in regions where sufficient lateral support from the surrounding soil is expected. To reproduce the non-linear behaviour of the soil, non-linear springs can be used.

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Fig. 9. INTAB sheet pile connection [11]

Here, the p-y method is a widely used empirical method, with soil pressure p being considered a non-linear function of pile deflection y. Essentially, the method involves introducing a series of p-y curves to represent the true behaviour of the soil by considering the non-linearity of the constrained soil modulus Es,k [19] based on the geotechnical survey. However, due to non-linearities, that concept does not allow for a superposition of load cases.

Therefore, the application of linear springs is a commonly used variant. In that case the possible distribution of spring constants over the depth is reduced to two cases (A: dense soil, B: loose soil, Fig. 10), which are considered separately taking into account the upper and lower limits of the coefficient of subgrade reaction ks or Es as given by the geological survey: – loose soil: config. B: 1–5 m: 0..ks,max, 5 m–base: ks,max – dense soil: config. A: 0–1 m: ks,min, 1 m–base: ks,max If the coefficient of subgrade reaction is not given, it can be estimated with the following equation based on the pile diameter Ds (for Ds # 1.00 m, set Ds " 1.00 m) [20]: ks,k =

Es,k Ds

(1)

The concepts shown here describe the consideration of the subsoil in a static analysis. However, a dynamic structural analysis of the superstructure also requires knowledge of the dynamic response of the soil structure, which in turn

D. Pak/H. Bigelow/M. Feldmann · Design of composite bridges with integral abutments

The additive load cases can be combined with the load cases “)T summer” and “)T winter” to reduce the number of load combinations, as the internal forces and deformations caused by these load cases are adequately affine.

The earth pressure “at rest” is determined according to EN 1997-1 [23]: q0(z) = σ 0(z) = K 0 ⋅ γ soil ⋅ z

Fig. 10. Application of springs to foundation piles

relies on dynamic soil properties measured in situ and including all the environmental factors at the time of testing [21]. The correct application of this spring-and-damper system is a focal point of the recent research project P1099 “Dynamische Auslegung von Verbundbrücken mit integralen Widerlagern” [dynamic design of composite bridges with integral abutments]. An overview of the project is given in section 5 of this paper.

4.2 Soil-abutment interaction To take the soil-abutment interaction into account for static calculations, a different approach has to be chosen.

The use of linear springs representing the backfill is not possible in this case, as…

– the deformations are too large, which calls for the use of non-linear springs at least in the upper part of the abutment for wall movements vh # vpassive,50 % (although these non-linear springs contradict the principle of load case superposition), and – the soil only supports the abutment on one side, thus the forces in the springs may become negative, calling for non-linear springs in any case.

Therefore, the soil is generally represented by an external loading. The self-weight of the consolidated backfill results in a horizontal earth pressure component qh (at-rest earth pressure). This load is increased by the temperature-induced deformations of the structure in the summer (earth pressure summer) and decreased in the winter (earth pressure winter). The load case “at rest” is defined as “lateral earth pressure due to surcharge”. The summer as well as winter load cases are mixed load cases, whereas the additive components () earth pressure summer, ) earth pressure winter) represent “another variable action” caused by temperature loading. To allow for a consistent application of safety factors, the winter load case and the summer load case are separated into the permanent load case “at rest” (LG,sup,ULS " 1.35, LG,inf,ULS " 1.0) and the additive load cases “) earth pressure summer” and “) earth pressure winter” (LQ,sup,ULS " 1.5, LQ,inf,ULS " 0) [22].

(2)

The earth pressure, which increases with increasing depth z, is based on the coefficient of earth pressure at rest K0 (according to Annex A of DIN EN 1997), the specific weight of the soil Lsoil.

The limiting value of passive earth pressure Xp(z) serves as a basis for the variable summer load case. However, its full application is far too conservative for bridges with short and medium spans, as the complete passive earth pressure is certainly not activated during the summer. Therefore, an approach proposed by Vogt [24] is adopted to determine the “mobilized” passive earth pressure Xp,mob(z) on the back wall, based on the maximum movement vp(z) of the abutment during the summer: qsummer(z) = σ p,mob(z) – σ 0(z)

(3)

σ p,mob(z) = K p,mob(z) ⋅ γ soil ⋅ z

(4)

K p,mob(z) = K 0 + (K p – K 0 ) ⋅

v p(z) a ⋅ z + v p(z)

(5)

The mobilized passive earth pressure is additionally dependent on the passive earth pressure coefficient Kp (according to Annex A of DIN EN 1997 and coefficient a: a " 0.1 for loose soil, a " 0.01 for dense soil; conservative assumption according to [25]: a " 0.01). The movement of the abutment vp(z) over the abutment’s height z is determined by applying the characteristic temperature load case “)T Summer” (L " 1.0) to the structural model, conservatively disregarding the earth pressure behind the abutment.

An iterative determination of the abutment deformation by considering the earth pressure is recommended for structures with low restraint (piles with low stiffness, shallow foundation).

The active earth pressure is a function of the active earth pressure coefficient Ka (according to Annex A of DIN EN 1997): q winter(z) = 1/2 ⋅ σ a(z) – σ 0(z)

(6)

σ a(z) = K a ⋅ γ soil ⋅ z

(7)

As this limiting value is already activated by a relatively small movement va of the abutment, it provides the basis for the variable winter load case without considering the actual movement of the abutment va(z). Conservatively, the active earth pressure is halved to ½ Xa(z).

However, this concept of equivalent loadings cannot be applied to dynamic problems. In that case a system of springs and dampers has to be applied based on the dynamic soil properties of the backfill. The correct applica-

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D. Pak/H. Bigelow/M. Feldmann · Design of composite bridges with integral abutments

Fig. 11. Teltow Canal bridge (photos: SSF Ingenieure AG)

tion of this spring-and-damper system is a focal point of the aforementioned research project P1099 “Dynamische Auslegung von Verbundbrücken mit integralen Widerlagern“.

5 Integral abutment bridges in Germany and Austria The share of long-span integral abutment bridges in Germany is rather poor. For instance, in 2005 just 2.1 % of the large highway bridges in Bavaria were integral or semi-integral ones. Some 78 % of those bridges were built between 1995 and 2005, which at least shows that integral bridges have become more and more important [26].

The main field of application for integral abutment bridges in Germany is the crossing of small side roads. In 1999 the German Federal Ministry of Transport, Building and Urban Affairs (BMVBS) published a list of 10 single-span prototype bridges [9] and recommended their use to the authorities. Eight of these 10 bridges were integral ones with a maximum span of 45 m.

In 2003 many recommendations from the BMVBS were withdrawn [27] and replaced by the Eurocode-based DIN Fachbericht 101 to 104 [28] – [31]. As no specific rules concerning integral bridges can be found in these standards, at present there is no specific standard or guideline for integral road bridges in Germany.

For the design of new integral abutment bridges, designers have to rely on their experience, which leads to the situation that only a small group of engineering companies has specialized in frame and/or integral bridges. It is common practice in Germany to use vaulted-tapered girders and beams, which allows for an easier transfer of the bending moments at the frame edge and leads to a very winsome bridge shape. In combination with very rigid foundations, quite long single-span bridges can be achieved compared with conventional bridges with bearings and joints.

The German railway company Deutsche Bahn AG has released its own codes and guidelines, which cover the design of integral concrete and conventional composite bridges at least. For the combination of both, designers still have to pass an internal approval procedure. Fig. 11 shows a composite railway bridge with integral abutments designed by SSF Ingenieure AG over the Teltow Canal in Berlin. Prefabricated elements were used for its construction. For the authorization procedure, the main focus was set on the engineering design of the frame edge [32].

Another example of an integral abutment bridge recently built in Austria is the railway bridge over the River

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Fig. 12. Composite railway bridge over the River Salzach (Austria), under construction (March 2016)

Salzach, situated in Neumarkt-St. Veit im Pongau (Fig. 12).

It was built by the Austrian railway company Österreichische Bundesbahnen (ÖBB). Besides its distinctive characteristics as an integral abutment bridge, it is also a so-called twin bridge. Twin bridges consist of two decks each carrying one track which are separated by a longitudinal gap but share a continuous ballast layer. This shared ballast layer leads to interaction between the two decks. The characteristics of twin bridges designed as filler-beam bridges are described in detail in [33], while [34] gives an insight into the ballast interaction behaviour.

Its structural properties at different construction stages are being examined experimentally within the scope of the aforementioned research project P1099 “Dynamische Auslegung von Verbundbrücken mit integralen Widerlagern“ (see Fig. 13). A focal point is the contribution of the backfill and the subsoil surrounding the foundation piles to the dynamic properties of the structural system.

The analysis of the measurement results combined with further parametric studies performed on calibrated models will allow for the elaboration of design procedures that consider dynamic soil-structure interaction. Therefore, tri-directional accelerations are measured as well as the relative displacements between the new deck and its twin, which is already in operation. Besides reactions to ambient excitations (e.g. wind, traffic on the second – not yet connected – track or on the road close by), the bridge is subjected to targeted excitation. A hydraulic vibrator excites the bridge at frequencies in a range between 0.5 and 30 Hz. Fig. 13 shows the measurement concept. The first construction stage investigated was the “naked” bridge with-

D. Pak/H. Bigelow/M. Feldmann · Design of composite bridges with integral abutments

Fig. 13. Measurement concept for the composite railway bridge over the River Salzach (Austria)

out backfill behind its abutments and without finishing elements such as ballast, sleepers, rails, etc. Therefore, there is no interaction between this deck and its twin either; the dynamic properties are not influenced by the backfill behind the abutments. In the second construction stage investigated, the backfill had been added. Comparison of the measurement results and numerical models of the bridge followed by parametric studies will increase knowledge about the contribution of the (unconsolidated) backfill to eigenfrequencies, eigenmodes and damping. The third measurement takes place after the bridge is completed, shortly before its final acceptance and the opening for traffic. The comparison of these two construction stages will increase the knowledge about the contribution of finishing elements to the dynamic properties of the bridge. This includes the interaction with the twin deck through the ballast bed already mentioned as well as the contributions of other elements, e.g. the restraining effects of the rails as described in [35].

The first three measurements have already taken place. The last measurement will be conducted approximately half a year after the third to investigate the influence of the – by then – consolidated backfill.

6 Conclusions Composite bridges with integral abutments combine several specific advantages. They are robust structures that do not require bearings and expansion joints. If constructed with a high degree of restraint, very slender superstructures can be achieved, combined with the need for no intermediate support. This kind of bridge, for example, is becoming more and more common for motorway crossings in southern Germany. However, care has to be taken regarding the consideration of soil-structure interaction.

For foundation piles as well as for backfill, different approaches are used to take the bedding and varying earth pressure distribution into account. These concepts are applicable in design offices, reflecting the complex behaviour of soil-structure interaction in a conservative yet still economic manner. While road bridges with integral abutments can thus be efficiently designed, the design of composite railway bridges with integral abutments calls for answers regarding the dynamic soil-structure interaction. This issue

is examined further in the FOSTA research project P1099 “Dynamische Auslegung von Verbundbrücken mit integralen Widerlagern“.

Acknowledgements The authors gratefully acknowledge the financial support of FOSTA (Forschungsvereinigung Stahlanwendungen, P1099 “Dynamische Auslegung von Verbundbrücken mit integralen Widerlagern“) and the European Research Fund for Coal and Steel (RFCS INTAB, RFCS INTAB , “Economic and Durable Design of Composite Bridges with Integral Abutments”).

Furthermore, the authors wish to thank their partners from P1099 (Lehrstuhl für Geotechnik im Bauwesen, RWTH Aachen University; Institut für Massivbau, Leibniz Universität Hannover; SSF Ingenieure AG; IBW Ingenieurbüro für Bauwerkserhaltung Weimar GmbH), Strabag SE and the railway companies ÖBB and DB. References [1] Braun, A.; Seidl, G.; Weizenegger, G.: Rahmentragwerke im Brückenbau. Beton- und Stahlbetonbau 101, No. 3, 2006, pp.

187–197. DOI: 10.1002/best.200600469 [2] Pak, D.; Feldmann, M.; Hechler, O.: Integral Abutment Bridges. Proc. of 5th European Conf. on Steel & Composite Structures, Graz, 3–5 Sept 2008. ECCS European Convention for Constructional Steelwork, Brussels, 2008, pp. 189–194.

[3] Collin, P: Some trends in Swedish bridge construction. Proc. of Intl. Conf. on Welded Structures, Budapest, 2–3 Sept 1996, pp. 163–172.

[4] Collin, P.; Johansson, B.: Wettbewerbsfähige Brücken in Verbundbauweise. Stahlbau 68, No. 11, 1999, pp. 908–918.

DOI: 10.1002/stab.199903180 [5] Schleicher, F.: Taschenbuch für Bauingenieure. Springer Verlag, Berlin/Göttingen/Heidelberg, 1949.

[6] White, H.: Integral Abutment bridges: comparison of current practice between European countries and the United States of America. Special Report 152, Transportation Research & Development Bureau, New York State Dept. of Transportation, 2007.

[7] Collin, P.; Veljkovic, M.; Petursson, H.: International Workshop on the Bridges with Integral Abutments. Technical Report 2006:14, Luleå University of Technology.

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[8] Glynn, A. C.: Bridge Manual, 4th ed. New York State Dept. of Transportation, Office of Structures, 2008.

[9] Bundesministerium für Verkehr, Bau- und Wohnungswesen: Allgemeines Rundschreiben Straßenbau 23/1999 – Musterentwürfe für einfeldrige Verbundüberbauten zur Überführung eines Wirtschaftsweges und eines Straßenquerschnittes RQ 10,5. Verkehrsblatt-Verlag, Dortmund.

[10] Feldmann, M.; Naumes, J.; Pak, D.; Veljkovic, M.; Eriksen, J.; Hechler, O.; et al.: Design Guide INTAB – Economic and Durable Design of Composite Bridges with Integral Abutments. RWTH Aachen University, Institut für Stahlbau, 2010.

[11] Feldmann, M.; Naumes, J.; Pak, D.; Veljkovic, M.; Nilsson, M.; Eriksen, J.; et al.: Economic and durable design of composite bridges with integral abutments. Final Report EUR24224. European Commission, Research Fund for Coal and Steel Unit (RFCS), Luxembourg, 2010.

[12] Glitsch, W.: Richtlinie “Integrale Bauwerke” – Sachstandsbericht. Stahlbau 82, No. 10, 2013, pp. 708–714.

DOI: 10.1002/stab.201310094 [13] Burke Jr., M.: Integral and Semi-Integral Bridges. John Wiley & Sons, Oxford, 2009.

[14] Arsoy, S.: Experimental and Analytical Investigations of Piles and Abutments of Integral Abutment Bridges. Doctoral Thesis, Virginia Polytechnic Institute & State University, Blacksburg, 2000.

[15] Maruri, R.; Petro, S.: Integral Abutments and Jointless Bridges (IAJB) 2004 Survey Summary. Proc. of 2005 – FHWA Conference (IAJB 2005), Baltimore, Maryland (USA), 16–18 Mar 2005, pp. 12–29.

[16] Dicleni, M.; Albhaisi, S.: Maximum length of integral bridges supported on steel H-piles driven in sand. Engineering Structures 25, 2003, pp. 1491–1504.

[17] Connal, J.: Integral Abutment Bridges – Australian and US Practice. Proc. of Austroads 5th Bridge Conf., Hobart, 19–21 May 2004.

[18] Abendroth, R. E.; Greimann, L. F.: Field Testing of Integral Abutments. Report, Iowa DOT Project, Iowa State University & Iowa Dept. of Transportation, 2005.

[19] Hällmark, R.: Low-cycle fatigue of Steel Piles in Integral Abutment Bridges. Thesis. Luleå University of Technology, 2006.

[20] DIN 1054 (Oct 2010): Baugrund – Sicherheitsnachweise im Erd- und Grundbau – Ergänzende Regelungen zu DIN EN 1997-1. Beuth Verlag GmbH, Berlin.

[21] Luna, R.; Jadi, H.: Determination of Dynamic Soil Properties Using Geophysical Methods. Proc. of 1st Intl. Conf. on the Application of Geophysical & NDT Methodologies to Transportation Facilities & Infrastructure, St. Louis, MO, Dec 2000.

[22] Pak, D.: Zu Stahl-Verbundbrücken mit integralen Widerlagern. Dissertation, RWTH Aachen University, Lehrstuhl für Stahlbau, Shaker Verlag, 2012, ISBN 978-3-8440-0362-8.

[23] DIN EN 1997 (Sept 2009): Eurocode 7: Entwurf, Berechnung und Bemessung in der Geotechnik – Teil 1: Allgemeine Regeln; Deutsche Fassung EN 1997-1:2004 AC:2009. Beuth Verlag GmbH, Berlin.

[24] Vogt, N.: Erdwiderstandsermittlung bei monotonen und wiederholten Wandbewegungen in Sand. Mitteilungen des Baugrundinstituts Stuttgart, No. 22, 1984.

[25] Berger, D.; Graubner, C.-A.; Pelke, E.; Zink, M.: Fugenloses Bauen – Entwurfshilfen für integrale Widerlagerbrücken. Hes-

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sisches Landesamt für Straßen- und Verkehrswesen, Wiesbaden, 2003.

[26] Schiefer, S.; et al.: Besonderheiten beim Entwurf semi-integraler Spannbetonbrücken: Eine Alternative im Brückenbau mit zunehmender Bedeutung – aufgezeigt am Beispiel der Fahrbachtalbrücke im Zuge der BAB A3 bei Aschaffenburg.

Beton- und Stahlbetonbau 101, No. 10, 2006, pp. 790–802, DOI: 10.1002/best.200600500 [27] Bundesministerium für Verkehr, Bau- und Wohnungswesen: Allgemeines Rundschreiben Straßenbau 13/2003 – Technische Baubestimmungen – DIN-Fachbericht 104 “Verbundbrücken”. Verkehrsblatt-Verlag, Dortmund.

[28] DIN-Fachbericht 101: Einwirkungen auf Brücken (Actions on bridges) (Mar 2009). Beuth Verlag GmbH, Berlin.

[29] DIN-Fachbericht 102: Betonbrücken (Concrete Bridges) (Mar 2009). Beuth Verlag GmbH, Berlin.

[30] DIN-Fachbericht 103: Stahlbrücken (Steel Bridges) (Mar 2009). Beuth Verlag GmbH, Berlin.

[31] DIN-Fachbericht 104: Verbundbrücken (Composite steel and concrete bridges) (Mar 2009). Beuth Verlag GmbH, Berlin.

[32] Koch, E.: Eisenbahnbrücken bei der Deutschen Bahn AG. Stahlbau 75, No. 10, 2006, DOI: 10.1002/stab.200610085 [33] Rauert, T.; Bigelow, H.; Hoffmeister, B.; Feldmann, M.; Patz, R.; Lippert, P.: Zum Einfluss baulicher Randbedingungen auf das dynamische Verhalten von WIB-Eisenbahnbrücken, Teil 1: Einführung und Messuntersuchungen an WiB-Brücken. Bautechnik 87, No. 11, 2010, pp. 665–672.

DOI: 10.1002/bate.201010044 [34] Rauert, T.; Bigelow, H.; Hoffmeister, B.; Feldmann, M.: On the prediction of the interaction effect caused by continuous ballast on filler beam railway bridges by experimentally supported numerical studies. Engineering Structures 2010, No. 12, 2010, pp. 3981–3988.

[35] Bigelow, H.; Feldmann, M.; Hoffmeister, B.; Zabel, V.: Zur Einspannwirkung von Eisenbahngleisen. Bautechnik 93, No. 7, 2016, pp. 462–469. DOI: 10.1002/bate.201600006 Keywords: integral abutment bridges; soil-structure interaction

Authors Prof. Dr.-Ing. Daniel Pak pak@bau.uni-siegen.de Universität Siegen Naturwissenschaftlich-Technische Fakultät Department Bauingenieurwesen Lehrstuhl für Stahlbau und Verbundbau Paul-Bonatz-Str. 9–11 57076 Siegen Germany Dipl.-Ing. Hetty Bigelow bigelow@stb.rwth-aachen.de Prof. Dr-Ing. Markus Feldmann feldmann@stb.rwth-aachen.de RWTH Aachen University Institut für Stahlbau und Lehrstuhl für Stahlbau und Leichtmetallbau Mies-van-der-Rohe-Str. 1 52074 Aachen Germany

Articles Behnoush Golchinfar Dimitri Donskoy Julius Pavlov Marcus Rutner*

DOI: 10.1002/stco.201710007

Remote monitoring of structural health in composites Dedicated to Prof. Dr. Eng. Akimitsu Kurita on his 70th birthday, in honour of his scientific achievements, guidance and the education of his students.

This paper explores a new interdisciplinary method for internal damage detection and tracking in composite materials using thermo-chemical sensing. A micro-sized network of strings is interwoven into the composites. Each string consists of a pair of tubes containing one of two different non-polar reactants. A local defect within the composites causes straining and cracking of the tube shell, resulting in direct contact between the two non-polar reactants. The latter undergo a chemical reaction resulting in a polar product. When exposed to a microwave energy source, a polar product heats up dramatically within seconds in comparison to the surrounding composite material or the non-polar reactants. This localized thermal signature can be rendered visible by an infrared camera.

This study summarizes the findings of an in-depth computational and experimental study of this sensing technology which is expected to be applicable across industries using composites, among them aerospace, automotive, offshore and bridge engineering. Potential applications in steel offshore or steel bridge engineering involve using composite sensing patches to cover fatigue fracture-critical components. Defects initiating on the steel substrate surface are expected to be sensed on demand with this proposed sensing technology.

1 Introduction There is a strong interest in measuring structural performance, which can be done automatically or on demand.

Although structural health monitoring is an established research field, there are still significant challenges to be tackled, shortcomings in current sensing technology and research demands with respect to sensing defects in structures built with composites [1], [2]. The motivation for this research is the lack of a single methodology for detecting internal defects in composite plates which is accurate, efficient, cost-effective and robust.

The research objectives are: – to develop a new thermo-chemical sensing and monitoring technique to detect internal defects in composites – through large-coverage sensing and – with no power source required on the structure to be monitored, – to develop a manufacturing procedure for an embedded sensing string network, and

* Corresponding author: mrutner@stevens.edu

– to develop diagnostics and prognostics that support integrity monitoring.

This research was primarily started due to the authors’ interest in damage detection in structural members built from composites. Composites are superior to other materials with regard to load-weight ratios and life cycle cost-savings, and are favoured as primary structural materials in many industries, such as the automotive, aerospace and wind turbine sectors, to name just a few. Composites are not the material of first choice for some industries, such as bridge or offshore engineering, although some studies have investigated using fibre-reinforced polymer composites for structural members in bridge engineering [3]. Composites are used for retrofitting and strengthening measures on metal structures [4]. The authors see one area of application for the proposed technology in composites used as load-carrying structural components. The proposed sensing technology uses smart materials, i.e. materials that exhibit a coupling between multiple physical domains [5].

Common examples are piezoelectric materials showing the direct and converse piezoelectric effect, which means they are capable of converting mechanical deformations into electrical signals and vice versa. The technology proposed here introduces a smart material that couples three physical domains: mechanical, chemical and thermal, as shown in Fig. 1. There is no literature on this proposed kind of coupling of the thermal, chemical and mechanical domains for structural health monitoring.

This paper builds on the preliminary findings achieved in [6] and conducts a feasibility study of the proposed technology with the aim of developing a robust, large-coverage, power- and maintenance-free on-demand inspection technology. The new sensing methodology is introduced in section 2. Analysis and experimental test results are presented in section 3. Potential applications for this sensing methodology are discussed in section 4 and the conclusions follow in section 5.

2 Sensing methodology 2.1 The idea The core of the concept of the proposed thermo-chemical sensing is the implementation of parallel channels containing chemicals on a micro-scale form factor in the form of fibres, which we call the “sensing strings”, as depicted in

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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B. Golchinfar/D. Donskoy/J. Pavlov/M. Rutner · Remote monitoring of structural health in composites

Table 1. Dielectric properties [7] 3 GHz

Fig. 1. Structural health monitoring using smart materials: new coupling of three domains

Fig. 2. The sensing string consists of a tube with two parallel channels containing one of two different molecular non-polar reactants. Multiple strings are embedded in the composite panel during the panel manufacturing process.

In this manufacturing process, the strings are rigidly bonded to the panel, yet are engineered to accommodate the strains experienced by the panel in normal operation.

Upon the formation of a crack or in the presence of excessive strain, yielding or other damaging event, the sensing string fractures, releasing small quantities of the chemicals encapsulated in the channels which then react to yield a molecular polar product, e.g. water. Upon controlled exposure to low-level microwave (MW) energy, the molecular polar substance heats up dramatically in contrast to the ambient composite material or the non-polar reactants due to high microwave absorption. Table 1 illustrates this contrast (three orders of magnitude in tanI) between typical materials used in composites and a polar material such as water [7].

The localized high thermal signature on the dry component surface above the internal defect can be imaged in

Material

J’/J0

tanI

J’/J0

tanI

PTFE polytetrafluoroethylene

2.1

0.00015

2.08

0.00037

PMMA polymethyl methacrylate

2.75

0.0078

2.75

0.0083

PE polyethylene

2.26

0.00031

2.26

0.00036

SiO2 fused quartz glass

3.78

0.00006

3.78

0.0001

Water

76.7

0.157

55

0.54

real time with a commercially available infrared (IR) camera. This thermo-chemical detection approach is shown schematically in Fig. 2. The MW heating and IR scanning and non-contact operation at a distance from the composite surface enable a very high component throughput during inspection. Depending on the MW source coverage and power, the proposed approach may achieve a scanning rate of tens of square inches per second, far exceeding the throughput of conventional techniques. Obviously, in order to be able to use MW and IR effectively and get meaningful results, thermal signature scanning requires the structure surface to be dry. The composites tested within this project were glass, Kevlar and carbon fibre-reinforced matrix composites with embedded sensing strings.

In addition to the high throughput potential, the proposed technology enables high resolution in the visualization of defects, thus supporting the characterization of defect size, type and growth rate. Further, the inspection methodology, being fully independent of electronic support circuitry or interfaces, is reliable and robust with regard to manufacturing and implementation in the composites, and achieves a service life as long as that of the structure. The proposed technology is expected to overcome many of the drawbacks of current inspection technology for internal defect detection in composites, thus contribut-

Fig. 2. Proposed structural health monitoring technology for internal defects in composites

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10 GHz

B. Golchinfar/D. Donskoy/J. Pavlov/M. Rutner · Remote monitoring of structural health in composites

Fig. 3. Temperature sensitivity of chemical product

ing to the availability of composite infrastructure or composite-strengthened metal infrastructure. The following subsections give insights into the physics behind the proposed inspection methodology.

2.2 Choice of chemical reactants This subsection describes fundamental research to find adequate non-polar reactants to maximize the large-coverage sensing sensitivity and efficiency of the technology. The chemical and physical properties of water make it uniquely suitable as a microwave-reactive temperature indicator. It is formed in a wide variety of reactions, reliably and with a high degree of control over the reaction rate. It is one of the safest products, since issues such as deflagration due to local overheating, toxicity or disposal measures are non-existent.

Water is, in fact, the optimum indicator molecule for microwave-induced temperature measurements. Heating up the cross-section and reading the thermal signature (using an IR camera) gives accurate information about important quantities in inspection, i.e. 1) internal defect location, 2) internal defect extent and 3) defect type. Further, the proposed rapid inspection enables high-frequency sensing on successive occasions to provide information about the defect growth rate.

For the purposes of preliminary feasibility testing, the authors employed the neutralization reaction of acetic acid (CH3COOH) with an alkali carbonate (potassium bicarbonate, KHCO3). An alkali carbonate was chosen as a test substance because of the following advantages: a) It is a salt of a weak acid and therefore reacts quickly with numerous organic acids, and b) the carbonic acid, which is the product of the reaction, decomposes into carbon dioxide and water, thus making the reaction effectively irreversible.

The reaction rate can be effectively regulated, for example, by using an organic acid with a higher molecular weight, such as heptanoic acid, as a matrix to suspend and dilute the acetic acid. It should be noted that the heptanoic acid will also react with the carbonate (as will any acid that is stronger than carbonic acid); however, the reaction rate is much lower and, for all practical purposes, insignificant compared with that of acetic acid. Using a diluent matrix of low polarity (such as that possessed by the longer-chain heptanoic acid) has the added effect of reducing the overall microwave absorption by the acetic acid, and therefore its temperature change upon microwave irradiation. In this way it can be ensured that the greatest temperature change when a reaction occurs is due to the product water absorbing the incident microwave radiation. Fig. 3 shows the significant temperature increase in the polar material (marked with crosses) within seconds when exposed to MW radiation. However, the non-polar materials, marked with triangles and squares, only show a minor temperature increase.

3 Analysis and results 3.1 Test specimen The test specimens were Kevlar fibre-reinforced epoxy matrix composites 100 mm long w 50 mm wide, as shown in Fig. 4a. Two different composite layups were investigated: the 2-ply layup and the 6-ply layup, as shown in Fig. 4b.

The manufacturing materials comprised Kevlar fibre plain weave (Fibre Glast®), epoxy resin (number 2000) and hardener (number 2020). The ratio of epoxy resin to hardener was 100:23 by weight. Each specimen was air-cured for 24 hours before being used for testing. For this test series, the sensing string tube was manufactured by inserting a syringe needle into the composite layup during laying up

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Table 2. Dielectric properties

Fig. 4. Kevlar fibre/epoxy specimen: a) geometry and b) layup

and curing. The syringe needle was removed after the composite panel had been cured. In order to test the sensitivity of defect detection, well-defined volumes of polar material were filled into the 35 mm long sensing string of the composite panel simulating the macro-size and micro-size defects, as shown in Fig. 4a. For the tests described here, the syringe needle diameter was 1.0 mm and the infill volumes studied were 0.1 and 0.2 ml. Test series currently ongoing using vacuum bagging are expected to produce sensing strings with even smaller diameters in the range of fractions of a millimetre.

3.2 Computational study The specimen as shown in Fig. 4 was also modelled within a computational study, particularly for the purpose of the feasibility and optimization of the sensing methodology.

Two different physics were coupled in the computation: electromagnetic heating and heat transfer. One challenge prior to running the analysis was certainly defining the input parameters required to run a successful simulation, specifically: dielectric constant, electrical conductivity and thermal conductivity. Table 2 shows the properties of the composite material and water as used in the simulation. Fig. 5

Unit

Kevlar/Epoxy

Water

Density

[kg/m3]

1450

1000

Thermal conductance

[W/(mK)]

0.65

0.6

Heat capacity at constant pressure

[J/(kgK)]

1420

4181.8

Relative dielectric permittivity

1

4 – 0.12j

80.36– 9.36j

Electric conductivity

[S/m]

0.8 w 10–3

0.025

shows four contour plots of the simulation results. The top of each contour plot refers to the respective layup of the composite specimen, i.e. 6- or 2-ply, and the volume of the polar product. The bottom of each contour plot refers to the maximum temperature of the thermal signal on the composite surface. It should be noted that the 0.2 ml polar product volume results in a stronger temperature signal than the 0.1 ml volume. Further, the 6-ply layup reduces the thermal signal being sensed on the composite surface compared to the 2-ply layup. The most important finding is that the method clearly represents the damage location, size and growth rate.

3.3 Experimental study All Kevlar composite specimens with different layups and polar material volumes were exposed to microwave radiation (2.45 GHz frequency, 1200 W) for 5 s. The thermal signature on the composite specimen surface was measured by a Satir PK160 infrared camera (detector resolution of 160w12). Fig. 6a shows the temperature contours on the specimen surface for each case. Fig. 6b summarizes the results with a temperature-time history.

The two major findings of the diagram in Fig. 6a are as follows. Firstly, the methodology is highly sensitive to the detection of multi-scale defects (from micro- to macro-size) in composite panels. As the sharp contour plots demonstrate, hand-held infrared camera photos have enough resolution to pick up internal defects in composite panels. Secondly, the composite layup and the volume of polar material affect the test outcome and resolution of the result. The authors also tested the crossing of sensing

Fig. 5. Simulation of thermal signal on surface of Kevlar/epoxy matrix composite specimen; two different composite layups (see Fig. 4b); two different volumes of non-polar material in sensing string, i.e. 0.1 and 0.2 ml

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Steel Construction 10 (2017), No. 1

B. Golchinfar/D. Donskoy/J. Pavlov/M. Rutner · Remote monitoring of structural health in composites

a)

b)

Fig. 6. a) Infrared photos of the Kevlar composite specimen; b) variation in polar material volume and composite layup

Fig. 7. Manufacture of Kevlar composite sensing patch: a) crossing sensing strings, b) IR image of defect indicated by crossing sensing strings, c) vacuum bagging and d) vacuum bagging pump

area remote scanning. The authors also see potential for this technology in detecting defects in structural steel components via composite patches covering fatigue fracture-critical structural components. Fatigue fracture-critical components and connections can be found in steel bridges (Fig. 8a) or in metal thin-walled structures, as shown here with an example from the aerospace sector, a Piper PA-46 (Fig. 8b). The reader should note the similarity of structural detailing between bridge and aerospace engineering [8]. Fig. 8c shows a schematic sketch of a sensing patch covering a fatigue fracture-critical connection which provides valuable information with respect to the four main questions of structural health monitoring: location of defect, size of defect, growth rate of defect and type of defect. Fig. 8d shows a schematic sketch of section A-A. The steel substrate is covered by the Kevlar composite patch, which has sensing strings interwoven into the composite material. The sensing patch is bonded to the steel substrate with an epoxy. Deformations of the steel substrate surface cause deformations in the epoxy and the thin composite patch. Owing to the localized deformation, the sensing strings become stretched and finally fracture when a deformation or strain threshold is exceeded, leading to the chemical reaction and the production of polar materials that can be sensed, as described in section 3. The sensing patch is currently under development in the laboratory.

5 Conclusions strings embedded in Kevlar composites, as marked in Fig. 7a by broken lines. Injected quantities of water (0.1 and 0.2 ml) indicate potential defects, as shown in the IR scan in Fig. 7b. The composite specimen was built with the vacuum bagging method, as shown in Figs. 7c and 7d.

4 Applications Potential applications for this new sensing technology can be found in many industries – industries where composite plates are used for load-carrying structural components, such as in the wind turbine, aerospace or automotive industries, as well as in composite armour plating.

The proposed sensing technology can be used to detect and monitor defects in composite components via large-

A new sensing methodology has been introduced which uses a smart material that couples thermal, chemical and mechanical domains. This methodology has multiple new features, since the method enables the detection of internal defects in composites a) through large-coverage sensing, b) through fast scanning that allows a throughput of tens of square centimetres per second, and c) it works with no power source required on the structure to be monitored.

Experimental and numerical studies have been undertaken to deliver proof of the concept and the results have been discussed. Several test parameters, such as the polar medium volume and the composite layup, have been found to affect the sensitivity of the sensing signal. These parameters have been varied in order to quantify the effect. It was noticed that the 0.2 ml polar product volume results in a

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B. Golchinfar/D. Donskoy/J. Pavlov/M. Rutner · Remote monitoring of structural health in composites

Fig. 8. Analogy between structural detailing in civil and aerospace engineering [8] – use of sensing patch on metal substrate

stronger temperature signal than the 0.1 ml volume. Further, the 6-ply layup reduces the thermal signal being sensed on the composite surface compared with the 2-ply layup. The method is able to represent clearly the damage location, size and growth rate.

Acknowledgements The granting of the Innovation and Entrepreneurship (I&E) Doctoral Fellowship, provided by Stevens Institute of Technology, to conduct the feasibility study and deliver proof of concept is much appreciated.

Intl. Workshop on Structural Health Monitoring (IWSHM), Stanford University, 1–3 Sept 2015.

[7] Paulauskas, F.; Bigelow, T.; Yarborough, K.; Meek, T.: Manufacturing of Carbon Fibers Using Microwave-Assisted Plasma Technology. SAE Technical Paper 2000-01-1527, Wiley, 2000.

DOI: 10.4271/2000-01-1527.

[8] FAA Aviation Safety: Special Airworthiness Information Bulletin, Wing Spar – Main Spar Cap Cracks, CE-15-18, 15 Jun 2015.

Keywords: structural health monitoring; composites; aerospace; bridges; crack; damage precursor

References Authors [1] Wenzel, H.: Health Monitoring of Bridges, Wiley, 2009.

[2] Farrar, C. F.; Worden, K.: Structural Health Monitoring, Wiley, 2013.

[3] Potyrala, P. B.: Use of Fibre Reinforced Polymer Composites in Bridge Construction. State of the Art in Hybrid and All-Composite Structures, Universitat Polytecnica de Catalunya, 2011.

[4] Chajes, M. J.; Chacon, A. P.; Swinehart, M. W.; Richardson, D. R.; Wenczel, G. C.; Liu, W.: Applications of Advanced Composites to Steel Bridges: A Case Study on the Ashland Bridge (Delaware-USA), DCT 181, Delaware Center for Transportation, University of Delaware, Mar 2005.

[5] Leo, D. J.: Engineering Analysis of Smart Material Systems, Wiley, 2007.

[6] Rutner, M.; Donskoy, D.; Pavlov, J.; Besser, R.: Internal Defect Detection in Composite Plates on Demand. Proc. of 10th

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Steel Construction 10 (2017), No. 1

Behnoush Golchinfar bgolchin@stevens.edu Dimitri Donskoy Ph.D., Associate Professor ddonskoy@stevens.edu Julius Pavlov, Ph.D. jpavlov@stevens.edu Dr.-Ing. habil. Marcus Rutner, Associate Professor mrutner@stevens.edu Postal address for all authors: Stevens Institute of Technology Schaefer School of Engineering & Science Department of Civil, Environmental & Ocean Engineering 1 Castle Point on Hudson Hoboken, NJ 07030-5991 USA

Articles Wei Zhang* Mengxue Wu Jin Zhu

DOI: 10.1002/stco.201710008

Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges In current bridge design codes or specifications, the dynamic effects of vehicles are considered by using a dynamic amplification factor (DAF) or dynamic load allowance (IM). However, a DAF is defined based on the ratio of the maximum dynamic load responses to the static load responses, and it is more appropriate for maximum value-based strength design. For fatigue design, stress cycles other than the maximum stress ranges could contribute to fatigue damage accumulations. Meanwhile, on the capacity side, a reduction in fatigue strength due to structural deterioration, which is related to local environmental conditions, including temperature, humidity, etc., could introduce more uncertainties into structural safety and reliability evaluation.

However, such multiple stress range effects and structural deterioration are not included in current bridge fatigue design. To evaluate the vehicular dynamic effects for the life cycle fatigue design of short-span bridges, the present study proposes a new dynamic amplification factor for life cycle bridge fatigue design (DALC), which is defined as the ratio of the life cycle nominal live load stress range to the maximum static stress range. In contrast to other traditionally defined dynamic factors, the newly defined DALC includes information about both the structural loading and the structural capacity. Therefore, the multiple stress cycles from vehicle-induced vibrations and the structural deteriorations from road surface conditions and corrosion of structural members are included. Parametric studies of DALC were carried out for multiple parameters and variables in the bridge’s design life cycle, for instance, possible faulting days in each year, fatigue strength exponent, corrosion parameters and corrosion level. The stochastic properties and uncertainties from these variables are also considered in the DALC calculation.

1 Introduction Fatigue is one of the typical structural failure modes and has attracted much research attention, especially given the current situation that the deteriorated infrastructures only received an overall grade of D based on the American Society of Civil Engineers’ (ASCE) 2013 Report Card [1].

As important nodes in the transportation network, bridges can experience cyclic loadings from surrounding environments, such as vehicles, wind, waves, ice, etc. Fatigue damage can accumulate progressively at structural details and lead to possible failures of structural components or catastrophic failures of the entire structural system. In the cur* Corresponding author: wzhang@uconn.edu

rent bridge design codes and specifications, such as the AASHTO LRFD specification [2], the fatigue limit state is defined as a load combination of fatigue and fracture relating to repetitive gravitational live and dynamic responses under a single design truck. The vehicle dynamic effects are included in the fatigue design by introducing the dynamic load allowance (IM), which is defined as an increment to be applied to the static wheel load to account for the dynamic impact of moving vehicles. The IM values used in codes and specifications are based on the vehicle– bridge interactions for an average road surface condition [3]. However, many researchers found that the IM values are greatly underestimated for poor road surface conditions, high vehicle speeds or their combined conditions [4]–[7]. A detailed review of dynamic amplification factors can be found in [8]. In the current AASHTO LRFD code, the number of stress cycles for each truck passage is set as one or two for most short- and medium-span bridges. However, in many analytical and experimental studies, more than one or two stress ranges have been found for each truck passage over a bridge [7], [9], [10]. In the present study, for simplification, the term “each truck passage” is used to illustrate the dynamic interactions of the bridge with one vehicle moving along the bridge. In addition, the IM only reflects the largest stress amplitude for one vehicle passing over the bridge, and the other smaller stress cycles cannot be included in the IM value as defined in the current codes or specifications. As a result, fatigue damage due to stress cycles other than the largest one are excluded.

Therefore, the maximum value-based dynamic amplification factor might not be appropriate for evaluating bridge fatigue damage due to vehicle loads with multiple stress cycles.

In addition, corrosion of structural components can start during their fabrication processes and much research has been carried out to investigate the deterioration of structural components or systems, such as steel bridge girders [11]–[13], bridge stay cables [14], [15], reinforcing bars in concrete structures [16], [17], offshore structures, port facilities [18]–[20] and underground or subsea pipelines [21]–[23] in various environmental conditions, such as atmospheric, coastal, marine or underground soil environments. As an infrastructure system often serves as a vital lifeline, the costs of structural maintenance and rehabilitation have been continuously increasing with the ongoing deterioration of the existing infrastructure systems due to

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

safety requirements. For example, the total annual direct and indirect costs (due to traffic delays and lost productivity) of corrosion for the replacement and maintenance of existing bridges could reach $91 billion according to the U.S. Federal Highway Administration (FHWA) and NACE International [24]. Further, corrosion due to corrosive environments and fatigue due to ambient conditions could interact with each other and result in severe damage, possibly leading to catastrophic failures, such as the tragedy of the Silver Bridge collapse in 1967, which was attributed to the combined action of a corrosive environment and cyclic stresses [25].

Owing to the remarkable progress in vehicle–bridge dynamics and metal corrosion analysis, it is possible to evaluate structural fatigue damage in the life cycle of the bridge due to random vehicle–bridge interactions, environmental deteriorations and combinations thereof. Based on the vehicle–bridge dynamic system, a dynamic amplification factor on stress ranges (DAFS) was proposed as the ratio of the nominal live load stress range to the maximum static stress range [8]. In contrast to the traditionally defined dynamic amplification factor (DAF) or IM, the nominal live load stress range is based on equivalent stress ranges obtained from the vehicle–bridge dynamic analysis with different randomly generated road profiles and vehicle types and speeds. In addition, as the bridge deteriorates progressively over its life cycle, the nominal stress range due to dynamic loads and the fatigue strength reduction due to corrosion should be updated. In the present study, to evaluate vehicular dynamic effects for the life cycle fatigue design of short-span bridges, a new dynamic amplification factor for lifecycle bridge fatigue design (DALC) is proposed for life cycle bridge fatigue design. The effects due to time-variant structural deterioration, varying vehicle–bridge interactions with different vehicle speeds and types and different road surface conditions are included in this single parameter for easy implementation in fatigue design. Nevertheless, with a direct correlation with the cumulative probability of failure, the DALC could be used in life cycle performance-based design with a target reliability level for bridge fatigue design.

2 Modelling time-variant vehicle–bridge dynamic systems 2.1 Modelling a vehicle–bridge dynamic system Modelling vehicle impacts on bridges started with including the vehicle as a constant moving force or moving mass [26], [27]. Later, the vehicle–bridge dynamic analysis was established after a fully coupled vehicle–bridge or vehicle– bridge–wind system was set up [28], [29]. The coupled equations of motions were set up to model the dynamic structural system, which can be found in the literature [8], [30]. The coupling forces between bridge and vehicles, modelled as coupling forces between the tyres and the randomly generated road surface, were proved to be significantly affected by the vehicle speed and road roughness conditions. Significant effects on the dynamic responses of short-span bridges were found in previous studies, and a systematic fatigue damage assessment approach was set up to include the effects of the progressively deteriorating road conditions and random dynamic vehicle loads over a bridge’s life cycle [7].

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2.2 Modelling corrosion of bridge components Research on bridge metal corrosion started in the 1940s.

The rate of material loss of metal specimens in various environments was examined by Albrecht and Naeemi [11] based on the large number of datasets collected. Later, in the 1980s, possible bridge corrosion types were classified

by Kayser and Nowak [31], [32] and fatigue strength reduction curves were defined after a deterioration model was developed. The inherent randomness in the deterioration process was believed to be the major source of large uncertainties in structural performance. Some major parameters for corrosion of structural members, such as deterioration rate (annual loss) and pattern (roughening and pitting), were identified [33]. To consider corrosion in engineering practice, several corrosion models were proposed based on experiments [34], such as the exponential model [35], the mass-loss model [36], the bilinear model [37] and the modified bilinear model [38]. In the present study, a two-segment model is used to estimate the average corrosion loss of the bridge components. In the first 10 years of exposure to a surrounding environment, the corrosion rate is assumed to be a constant value and the average corrosion depth of the bridge components for the first 10 years of exposure can be obtained from Eq. (1), which is based on a linear model [37]. After 10 years of service, deterioration will have progressed at certain locations and an accelerated corrosion process will take place, so the exponential model might be better for estimating corrosion penetration [35], [39], see Eq. (2). Therefore, in the present study, the average corrosion loss in the two-segment model is expressed in the following two equations: For the first 10 years of exposure: R = t ⋅ rav

(1)

After the first 10 years of exposure:

(

)

R = 10 ⋅ rav + A t − 10

B

(2)

where: R average corrosion penetration [mm] t number of years average corrosion rate during first 10 years of exporav sure [mm/year] A, B parameters determined from the analysis of experimental data Three corrosion levels (high, medium and low) were defined, corresponding to marine, urban and rural environments [11], [39]. The statistical parameters for A and B are listed in Table 1 for carbon steel [11]. The average corrosion rates rav during the first 10 years of exposure are defined in ISO 9224 [37]. The upper and lower bounds for corrosion rates for different corrosion levels are listed in Table 2 [37].

Based on several field surveys, corrosion was found to occur most likely at two locations on the bridge girders, namely, the top surface of the bottom flange and the web near supports [32], [40]. Surface corrosion-induced section loss is usually modelled with a sectional material loss, and such effects could be included in the equations of motions

W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

Table 1. Statistical parameters for A and B for carbon steel Environment Rural (low) Urban (medium) Marine (high)

Parameter

Mean value

Coefficient of variation

(

)

Bound

Average corrosion rate rav (mm)

A

0.034 mm

0.09

Low

Medium

High

B

0.65

0.10

Upper

0.005

0.012

0.03

A

0.0802 mm

0.42

Lower

0.0005

0.005

0.012

B

0.593

0.4

A

0.0706 mm

0.66

B

0.789

0.49

via structural stiffness matrices. Based on the fatigue test data on corroded samples with specific materials and corrosion environments, the fatigue category for corroded steels could downgrade gradually when the steel plates are heavily pitted [41], [42]. Even though the corrosion effect on fatigue category has been addressed in many codes or specifications [43], [44], such corrosion effects are not addressed in current bridge design codes or specifications.

In 1994 Hahin defined a general corrosion fatigue equation that takes into account general corrosion and pitting [42]: N = A ' Scorr

Table 2. Average corrosion rates of carbon steel during the first 10 years

−m

(3)

roughness coefficient (RRC) [7], [47]. The local road discontinuities can be modelled with a step up or down for the faulting between approach slab and pavement and between bridge deck and approach slab. As the bridge ages, surface abrasion due to heavily loaded vehicles and environmental erosion due to infiltrating water or de-icing salts causes the road surface to deteriorate gradually. In order to consider the time-dependent deterioration of the road surface conditions, a progressive road roughness deterioration model for the bridge deck surface was introduced and the road surface profiles were generated randomly for different conditions in the bridge’s life cycle [7]. As the road profiles are introduced into the equations of motion for the vehicle–bridge dynamic system by affecting the contact forces between vehicle tyre and bridge deck, the calculated dynamic stress ranges could include the time-variant deterioration of road surface conditions.

3 Vehicular dynamic effects for life cycle fatigue design 3.1 Dynamic amplification factor (DAF)

where: A′ = A/K f Kf " 1.2 11.54 rt fatigue reduction factor, related to pitting r corrosion rate [inch/year] t time [year] stress range, and should be adjusted for Scorr moment of inertia or section loss due to corrosion m fatigue strength exponent, which is approx. 3.26 for a corrosion environment [42] For comparison, the relationship between number of cycles to failure N and equivalent constant amplitude stress range Sre for steel without corrosion is given below [45]:

( )

N = A ⋅ Sre

−m

When a vehicle travels along a bridge, the dynamic effects caused by the moving vehicle, such as increases in deflections and stresses, are considered by using the dynamic load allowance (IM) defined in the current AASHTO LRFD [2]. The IM is defined as an increment in the static load to account for the dynamic effects of moving vehicles.

Therefore, the maximum dynamic response of the moving vehicles can be obtained as follows [48]: R dyn = R sta ⋅ (1 + IM/100)

(5)

where: maximum static response Rsta IM/100 dynamic amplification (DA) (1 IM/100) dynamic amplification factor (DAF) for the bridge

(4)

where A is the detail constant that is typically defined in design codes, such as Table 6.6.1.2.5-1 in AASHTO [2]. For the corrosion fatigue of high-tensile steel plates, it is worth noting that even a small stress fluctuation could lead to fatigue damage accumulation since there is no fatigue limit for corrosion fatigue [46].

2.3 Modelling of road surface condition A twofold road surface condition including the effects from both long undulations and local defects in road profiles is used in the vehicle–bridge dynamic system. The long undulations in the roadway pavement were generated through an inverse Fourier transformation using the road

For example, a DAF value of 1.15 corresponds to a DA of 0.15 and an IM of 15. The IMs adopted for the fatigue and fracture limit state and all the other limit states for bridge components, except the joints, are 15 and 33 respectively in AASHTO LRFD [2].

To calculate bridge deflections, moments, shears and stresses considering dynamic load effects, Billing [4] presented the equations for computing dynamic amplifications (DA) when dealing with the bridge responses due to a passing vehicle. As the DA is defined based on maximum structural displacement, it is not strictly applicable to bridge fatigue design, where multiple stress cycles generated by moving vehicles could introduce fatigue damage accumulations. Further, the dynamic amplification effects could also be different at different locations on the bridge,

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W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

such as the negative and positive moment regions, and the underestimated negative region dynamic response could potentially lead to a fatigue life overestimate amounting to 29 % [4]. Green [49] also presented an interesting example of defining a zero DA for large dynamic responses based on the current definition of DA, which suggests that DA might not be an effective measure for considering dynamic effects in fatigue design. In addition, compared with other parameters, such as the first natural frequency of the bridge, vehicle speed, vehicle suspension systems and initial vehicle vibrations, which have effects on the dynamic amplification factor (DAF) or DA, the road surface profiles were found to have a tremendous effect on DAF or DA [48], [50], [51]. Even though the deck surface roughness is considered as a major factor for vehicular dynamic effects on bridges, estimating the effects of long-term time-variant deck deteriorations at the bridge design stage is still challenging.

3.2 Revised equivalent stress range As discussed earlier, two correlated parameters, i.e. the equivalent stress range and the number of stress cycles per truck passage, have been used to evaluate vehicular dynamic impacts on bridges. For cyclic loading-induced variable stress ranges, the Palmgren-Miner damage law is often used for fatigue damage accumulations [52]. The accumulated damage is defined as D " 8ni/Ni, where ni " number of stress cycles of stress range Si and Ni " number of stress cycles to failure in structural component if the stress range is Si. Fatigue life could be expressed in terms of cycles to failure Ni = A ⋅ (Si )−m [45]. To combine the two parameters into one for simplicity, a revised equivalent stress range Sw is defined on the basis of equivalent fatigue damage [7]. Therefore, the fatigue damage of multiple stress cycles due to each truck passage is simplified to that of a single stress cycle Sw [7]. Based on this definition, the fatigue damage due to one stress cycle Sw is D = A −1 ⋅ (S w )m , and is equal to the fatigue damage due to multiple variable stress ranges D = A −1 ⋅ n ⋅ (Sre )m . For one truck passage j, the revised equivalent stress range is

( )

j = Nj Sw c

1/m

j ⋅ Sre

(6)

where: Ncj number of stress cycles due to the passage of the jth truck j Sre equivalent stress range of the stress cycles due to the jth truck m material constant, which could be assumed as 3.0 for all fatigue categories [53] When analysing the stress history for vehicle loadings, the rain-flow counting method is usually used to obtain the numbers of cycles per truck passage. As the stress range cut-off levels greatly affect the number of cycles due to vehicle loadings, a stress range threshold is usually defined. For example, Kwon and Frangopol [45] defined a range for the stress cut-off level as from 3.45 Mpa (0.5 ksi) to 33 % of the constant amplitude fatigue limit (CAFL).

However, after implementing an equivalent stress range concept for fatigue design, the cut-off stress range has only

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Steel Construction 10 (2017), No. 1

a limited effect if a small cut-off range (! 3.45 Mpa or 0.5 ksi) is defined, since the fatigue damage contribution of smaller stress ranges is very small when the PalmgrenMiner damage law is applied.

3.3 Dynamic amplification factor for life cycle bridge design For bridge fatigue design, structural details, such as bridge welds, should satisfy the following equation in order to consider load-induced fatigue [2]:

( ) ( )n

γ Δf ≤ ΔF

(7)

where: L load factor )f live load stress range due to passage of fatigue load nominal fatigue resistance ΔF

( )n

Owing to the complex vehicle–bridge interactions, the vehicle-induced dynamic stress ranges could be treated as a random variable considering different bridge types, vehicle types and speeds and road surface conditions. Certain distribution types, such as normal or lognormal distributions, could be used to demonstrate the randomness of the stress ranges for different scenarios with certain bridge, vehicle and road surface conditions [7]. Therefore, the revised equivalent stress range for a given bridge during its service life can be defined as

∑ (pj) ⋅ (Swj)

⎡ Slc w = ⎢ ⎢⎣

j

1/m m⎤

(8)

⎥ ⎥⎦

where: pj the probability of case j (and here case j is defined as a combination of vehicle type, vehicle speed, road roughness condition and number of lanes) j Sw revised equivalent stress range for case j, which can be assumed to follow a normal distribution [7] m fatigue strength exponent, which is a random variable As discussed earlier, structural deteriorations, which appear on the capacity side of the structural design equations, are not included in the current bridge fatigue design codes. To simplify the design process, a new dynamic amplification factor for life cycle bridge design (DALC) will therefore be proposed in the present study to consider both sides of the design equations, namely, the loading side, such as vehicle types and speeds and road surface conditions, and the capacity side, such as the fatigue strength, which could reduce as corrosion progresses. For corroded members, the relationship between the structural fatigue life in terms of cycles to failure and stress ranges is updated from Eq. (4) to Eq. (3), and Eq. (8) is updated to the following for discrete time integration: 1/m

∑ ∑ (qk ) ⋅ (pj) ⋅ (Swj )m⎥⎤

Slc = ⎡ ⎢ ⎣

k

j

(9)

⎦

where qk is a time-variant coefficient for corrosion and is defined as q k = A / A′

(10)

W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

where A ' is a time-dependent detail constant as shown in Eq. (3). It is worth noting that the fatigue strength reduction, which demonstrates the structural capacity, is included in the equivalent stress ranges after implementing Eq. (10).

Owing to the uncertainties associated with the revised j , the fatigue strength exponent equivalent stress range S w m and the corrosion parameters A and B (shown in Table 1), a fit-of-goodness test, such as the chi-square test, is carried out to check the distribution type of the revised equivalent stress range for the whole design life Slc. Based on Sturges’ rule and the number of samples (500) used in the present study, the number of intervals was chosen as 10 and the degree of freedom 7. With a 5 % significance level, the test limit for the chi-square test was calculated as C1−α,f = C0.95,7 = 14.07.

With a known distribution for Slc, a nominal equivalent stress range in the life cycle Slc n can be defined with a designated reliability index G, such as 3.5, which is typically used in AASHTO LRFD [2] for ultimate strength limit states. If the cumulative distribution function of the revised equivalent stress range in the life cycle is defined as F, the nominal equivalent stress range in the life cycle Slc n can be calculated as (11) () where Φ ($) denotes the standard normal cumulative dis-

−1 ⎡ ⎤ Slc n = F ⎣Φ β ⎦

tribution function.

The reliability-based dynamic amplification factor for the revised equivalent stress range for life cycle bridge fatigue design (DALC) is defined in Eq. (12). For comparison, the definition of the dynamic amplification factor (DAF) based on maximum responses [48] is shown in Eq. (13).

DALC =

Slc n Sst

(12)

R dyn R sta

To demonstrate the procedures required to obtain the newly defined DALC, one short-span slab-on-girder bridge is used, which is designed in accordance with the AASHTO LRFD bridge design specification [2]. The bridge has a span of 12 m and a width of 13 m, and it can accommodate two vehicle lanes in the same direction.

The concrete deck is 0.19 m thick, and the haunch is 40 mm high. All of the six steel girders are W27w94 and are evenly spaced at 2.3 m. Some other design parameters of the bridge model can be found in the literature [8]. For the vehicle modes, three 3D numerical models of trucks with two, three and five axles were used, and the average daily truck traffic for those trucks was assumed to be 600, 400 and 1000 respectively. Owing to the bridge’s short span, only one truck was assumed to pass over the whole bridge at one time and the vehicle speeds for the different trucks were assumed to follow normal distributions with the same mean value and standard derivation. The three types of truck with two, three and five axles were based on AASHTO H20-44, HS20-44 and 3S2 [54]. The geometry, mass distribution, damping and stiffness of the tyres and suspension systems of the three types of truck can be found in the literature [8], [47]. In the vehicle–bridge dynamic system, the road profile was generated randomly through the superposition of a random road profile, which was assumed to be a zero-mean stationary Gaussian random process, to represent long undulations in the roadway pavement and the faulting data to represent road surface discontinuities that cause hammer effects, such as uneven joints, potholes and bumps. In addition, the road roughness coefficient was treated as a function of time [7]. In a full roadway pavement maintenance cycle of 13 years in the present study, 7, 2, 2, 1 and 1 years are classified as very good, good, average, poor and very poor road conditions respectively.

4.2 Revised equivalent stress ranges

where Sst is the maximum static stress range due to the passage of the live loads without considering the dynamic effects.

DAF =

4 Numerical example 4.1 Bridge and vehicle model

= 1 + IM /100 =

Sdyn Sst

(13)

where Rdyn is the maximum dynamic response, Rsta the maximum static response and Sdyn the maximum dynamic stress range.

When the maximum static stress range Sst and DALC are known, it is possible to calculate the reliability-based nominal live load stress range for the whole design life Slc n, which includes the fatigue damage due to multiple stress range cycles caused by each vehicle passage with various vehicle speeds, various road conditions and different corrosion conditions of the structural members. Comparing the different definitions of DALC and DAF in Eqs. (12) and (13), DAF could possibly underestimate fatigue damage accumulations due to multiple stress cycles since it is based on a deterministic ratio of the maximum dynamic response and the maximum static response.

After solving the equations of motion for the vehicle– bridge dynamic system, the revised equivalent stress ranges are obtained for various vehicle speeds and road roughness conditions. Based on the previous study, normal distribution is one of the acceptable distribution types for the revised equivalent stress range [7], [8]. Generally, the revised equivalent stress ranges increase with the vehicle speed and the deterioration of the road surface condition.

Nevertheless, the dynamic displacement of bridges was found to change with the vehicle speed [29], [48], [55], [56].

Based on the random traffic samples collected, statistical techniques show that vehicle speed follows a normal distribution, which allows the vehicle speed to be conveniently described with two parameters: mean and standard deviation. In the present study the 85th percentile speed was approximated as the sum of the mean value and one standard deviation for simplification. For a speed limit of 26.8 m/s (60 mph) and a coefficient of variation of vehicle speed of 0.2, the mean vehicle speed obtained was 22.3 m/s (50 mph). In order to simplify the calculations, the randomly generated vehicle speeds were grouped into six

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W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

ranges represented by vehicle speeds from 10 m/s (22.4 mph) to 60 m/s (134.4 mph).

4.3 Parametric study 4.3.1 Number of days with faulting Over a bridge’s life cycle, the vehicle–bridge dynamic system could have different vehicle speeds, vehicle types and vehicle numbers. Further, the structural material could deteriorate at a different rate throughout the life cycle and the road surface condition could change as well. As a result, the DALC could vary greatly for different bridges with the different parameters mentioned earlier for the vehicle– bridge dynamic system. For the road surface conditions, uneven joints and, possibly, potholes and faulting could exist in the bridge’s life cycle even with continuous maintenance. Effective rehabilitation and maintenance could minimize the frequency of the occurrence these discontinuities and the time needed to eliminate them. However, downtimes could still last from less than one day to several days or several months in a whole year if the maintenance actions are not carried out in time. As a result, numbers of days with faulting could be an important parameter for the DALC. As discussed in section 3.3, distribution types for the revised equivalent stress range for the whole design life Slc should be checked first in order to get the DALC. For the cases with different numbers of days with faulting, the chi-square tests for the revised equivalent stress range Slc were carried out to check whether or not normal and lognormal distributions are acceptable. The results are listed in Table 3. It is worth noting that the results shown in the table are based on the medium corrosion category of structural members and the fatigue strength exponent m is assumed to have a constant value of 3.00 or 3.26. As shown in the table, both normal and lognormal distributions are acceptable for Slc with different numbers of days with faulting. In the present study the lognormal type was adopted to describe the distribution of the revised equivalent stress range in the life cycle. Therefore, the nominal equivalent stress range Slc n corresponding to the designated reliability index G " 3.5 can be obtained using Eq. (11).

Based on Eq. (12), the DALC values for different numbers of days with faulting in the bridge’s design life are obtained, see Fig. 1. As shown in the figure, DALC values are found to increase with the number of days with faulting. When there is no faulting or only one day with faultTable 3. Chi-square tests for Slc with different numbers of days with faulting m " 3.00

Faulting days

Normal

Lognormal

0

3.3

1

7.2

7

42

m " 3.26 Normal

Lognormal

2.3

5.9

5.8

6.1

10.2

8.3

8.4

6.4

11.1

9.3

15

7.2

6.8

10.0

8.5

30

10.8

9.1

6.2

6.3

60

6.4

5.2

5.1

4.0

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Fig. 1. Effects of faulting days on DALC

ing, the DALC value is about 5.4 for m " 3.00 and 5.6 for m " 3.26. However, with an increase in the faulting days to two months, the DALC values only increase by 3.7 % and 3.6 % respectively, to 5.6 for m " 3.00 and 5.8 for m " 3.26.

This suggests that the number of days with faulting has only a limited effect on the DALC when the corrosion of bridge components is considered. In addition, with a fatigue strength exponent m " 3.26, the DALC could be about 3.7 % more than for the condition of m " 3.00, which indicates that the fatigue strength exponent also has a limited effect on the DALC for such an increase from m " 3.0 to m " 3.26.

4.3.2 Fatigue strength exponent and corrosion level Hahin [42] found that the fatigue strength exponent m could vary between 2.9 and 3.8 with few exceptions, and the mean value of m could be set as 3.26 for the corrosive environment. To check the effects of the fatigue strength exponent on the DALC, different deterministic values of fatigue strength exponent m were adopted to evaluate the DALC, with three corrosion environments being considered: rural (low corrosion level), urban (medium) and marine (high). The statistical values for the corrosion parameters A and B [11] and the values of corrosion rate [37] as shown in Tables 1 and 2 were adopted for the different corrosion categories. The results are shown in Fig. 2. It is worth noting that the results are based on one faulting day in each year. As shown in Fig. 2, the fatigue strength exponent m could have a large impact on DALC values when m increases from 2.9 to 3.8. The DALC could increase by 10–11 % when the fatigue strength exponent m increases from 2.9 to 3.8 at each corrosion level, which indicates greater fatigue damage and a shorter fatigue life expectation. In addition, the DALC values also increase with the level of corrosion due to rural, urban and marine conditions. From the rural corrosion level to the urban corrosion level and marine corrosion level, the mean value of DALC increment could increase by about 3.7 % and 9.0 % for the same fatigue strength exponent m. As a result, fatigue damage accumulations could accelerate under corro-

W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

sive environments and the fatigue life for the bridge could be reduced accordingly.

Since the fatigue strength exponent should be treated as a random variable, m is assumed to follow a normal distribution in order to improve our understanding of its effects on the DALC. As listed in Table 4, the mean values of the fatigue strength exponent m are set to 3.00 and 3.26, and the COVs are set to 0.00, 0.025, 0.05, 0.075 and 0.10.

Similarly, the chi-square tests for the revised equivalent stress range of the whole design life Slc were carried out and the results are listed in Table 4 as well. As shown in the table, when the COV of m is # 0.05, normal distribution is not acceptable for Slc in many cases; only the lognormal distribution is acceptable. The effects of a random fatigue strength exponent m on the DALC are shown in Fig. 3. As shown in the figure, the COV has significant effects on the DALC values and the DALC value could increase by 1.2 when the COV increases from 0.00 to 0.10 for a mean m value of 3.26. In addition, when the mean value of m increases from 3.00 to 3.26, the DALC values increase by about 0.2 for the same COV value of m. Overall, the DALC values will be influenced significantly by the fatigue strength exponent m. Higher mean values and/or higher derivations of m will lead to a larger DALC. Therefore, the fatigue strength exponent m could be a very important parameter for obtaining a reasonable DALC in bridge fatigue design.

Fig. 2. Effects of fatigue strength exponent and corrosion level on DALC Table 4. Chi-square tests for and COVs of m

Slc

with different mean values

Mean value of m COV of m

3.00

3.26

Fig. 3. Effects of random fatigue strength exponent m on DALC

4.3.3 Corrosion parameters Corrosion parameters A and B are obtained from experiments and uncertainties could be introduced into the corrosion model and the DALC. In the present study, to consider the random effects, the corrosion parameters A and B are assumed to follow a normal distribution with the mean values and COVs shown in Table 1. Four cases were defined to calculate the DALC value: both of the parameters A and B are deterministic (case AB-00), only parameter A is random (case AB-10), only parameter B is random (case AB-01), and both parameters A and B are random (case AB-11). Three corrosion environments were adopted for the structural components: rural (low corrosion level), urban (medium) and marine (high). For all of the cases, the fatigue strength exponent m was set as a deterministic value. The chi-square tests of revised equivalent stress range Slc for a medium corrosion level are listed in Table 5.

As shown in the Table, both normal and lognormal distributions are acceptable for Slc in different cases with different corrosion parameters A and B. Table 5. Chi-square tests for Slc with different cases of corrosion parameters A and B (N " normal; LN " lognormal) m 2.9

Case AB-00

Case AB-10

Case AB-01

Case AB-11

N

LN

N

LN

N

LN

N

LN

3.8

2.3

1.4

1.8

11.0

7.3

11.7

7.8

3.0

7.2

6.1

6.4

4.8

7.3

4.9

4.9

3.1

3.1

9.4

8.9

7.7

6.0

9.6

7.9

13.8

10.1

3.2

5.8

4.7

4.4

5.7

6.7

4.3

12.3

9.5

Normal

Lognormal

Normal

Lognormal

3.3

8.1

5.7

10.9

7.6

6.4

4.4

10.3

10.0

0.00

7.2

6.1

9.4

6.8

3.4

8.7

5.5

11.6

8.9

13.9

9.9

4.1

2.1

0.025

5.0

5.3

9.3

8.0

3.5

8.3

5.6

3.6

2.8

5.3

5.8

10.3

6.0

0.05

7.4

4.7

12.1

8.6

3.6

9.6

6.4

9.5

7.8

9.2

6.1

10.9

6.2

0.075

14.1

9.5

11.2

7.2

3.7

11.4

8.5

10.4

7.3

7.8

5.8

9.9

8.0

0.10

16.3

10.8

16.9

11.0

3.8

6.4

3.9

11.1

8.4

13.4

9.7

12.1

8.3

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W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

The effects of corrosion parameters A and B on the DALC in different corrosion environments are shown in Fig. 4. As shown in Fig. 4a, the DALC values for four different cases are almost the same, which indicates that the randomness of the corrosion parameters has no effect on the DALC in the low corrosion environment. As shown in Fig. 4b, for the urban environment (medium corrosion level) there are almost no differences in the DALC values between cases AB-00 and AB-10. Similarly, only limited differences can be found in the DALC values between cases AB-01 and AB-11. However, from case AB-00 to case AB-11, the DALC values increase by about 0.2 (3.7 %) for the same fatigue strength exponent m. This suggests that the randomness of the corrosion parameters has effects on the DALC, but they are very limited in the medium corrosion environment. As shown in Fig. 4c, although there are very few differences between DALC values for cases AB-00 and AB-10, also between cases AB-01 and AB-11, the DALC values increase by about 25.4 % from case AB-00 to case AB-11 for the same fatigue strength exponent m. This indicates that the randomness of the corrosion parameters could have large impacts on the DALC in the high corrosion environment. In particular, parameter B has a more significant effect on the DALC than parameter A at the high corrosion level.

5 Concluding remarks Moving from a traditionally defined dynamic factor (DAF) and the dynamic amplification factor on stress ranges (DAFS), a new dynamic amplification factor, DALC, is proposed for evaluating vehicle dynamic effects in the life cycle fatigue design of short-span bridges. Information about both environmental loading and structural capacity are included in the newly defined parameters. A numerical simulation for solving a coupled vehicle–bridge system, including a 3D suspension vehicle model and a 3D dynamic bridge model, is used to obtain the revised equivalent stress range. The model includes fatigue damage due to multiple stress range cycles caused by vehicle passages at various vehicle speeds under various road conditions in the bridge’s life cycle as well as the effects of the long-term deck deterioration and the corrosion-induced fatigue strength reduction. Parametric studies of the DALC were carried out to discover the effects due to multiple variables in the bridge’s life cycle, e.g. faulting days per year, fatigue strength exponent, corrosion parameters and corrosion level. The stochastic properties and uncertainties due to these variables were also considered in the DALC calculation. The present study allows the following conclusions to be drawn: – The DALC can be used for life cycle bridge fatigue design with a given reliability level. A larger DALC value suggests a larger fatigue damage accumulation and a shorter fatigue life.

– The fatigue strength exponent could be a very important parameter for obtaining a reasonable DALC in bridge fatigue design. The effects of the randomness of the corrosion parameters on DALC could increase as the corrosion level intensifies. When considering the random effect of corrosion parameter B in the high corrosion environment, the DALC could increase greatly.

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Fig. 4. Effects of corrosion parameters on DALC for different corrosion levels

W. Zhang/M. Wu/J. Zhu · Evaluation of vehicular dynamic effects for the life cycle fatigue design of short-span bridges

– The DALC increases with the corrosion level. Fatigue damage accumulation could accelerate under corrosive conditions and the bridge’s life could be greatly shortened.

As a further expansion of the DAFS defined previously, the new DALC includes the corrosion-induced fatigue strength reduction and preserves the effects due to road surface conditions and traffic information, such as vehicle speed and type. In contrast to DAF and DAFS, the newly proposed DALC includes both the loading information calculated from the vehicle–bridge dynamic system and structural capacity information when implementing the equivalent fatigue damage concept. Moreover, different fatigue strength exponents, corrosion parameters and their randomness could lead to different DALC results. In order to obtain a reasonable DALC for bridge fatigue design, more numerical simulations and sensitivity studies are needed to obtain the DALC as a function of the multiple parameters.

Nevertheless, as the corrosion conditions greatly affect the fatigue life, it is extremely important to consider historical bridge maintenance activities and future maintenance plans in order to reach a better fatigue life estimation.

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Keywords: dynamic amplification factor; life cycle bridge design; fatigue; environmental corrosion; structural dynamic analysis; revised equivalent stress range

Authors Wei Zhang, Assistant Professor Deptartment of Civil and Environmental Engineering University of Connecticut, Unit 3037 261 Glenbrook Rd. Connecticut, CT06269 USA Mengxue Wu, Lecturer Department of Civil Engineering and Architecture Southwest Petroleum University Chengdu, China Jin Zhu Deptartment of Civil and Environmental Engineering University of Connecticut jin.zhu@uconn.edu Postal address for Mengxue Wu and Jin Zhu: 180C Foster Drive, Willimantic Windham, CT 06226 USA

Articles Xi Li Branko Glisic*

DOI: 10.1002/stco.201710009

Evaluating early-age shrinkage effects in steel-concrete composite beam-like structures The shrinkage of concrete is a material phenomenon that occurs due to the reduction in the concrete’s volume over time. In composite beam-like structures, the shrinkage of concrete elements affects strains and stresses in the overall composite section. The ability to identify and estimate early-age shrinkage in beam-like structures allows the creation of a baseline for strain-based structural health monitoring, and enables a more thorough understanding of structural performance and condition. In this project, the early-age shrinkage behaviour of steel-concrete composite beam structures was studied with an integrated approach using monitoring data from a reduced-scale test structure and simulation results from finite element analysis. A simplified analytical expression was developed to study further the range of magnitude of shrinkage in the concrete slab and the resultant strain distribution in steel girders. The goal of this paper is to create an innovative, comprehensive and widely applicable procedure that identifies and quantifies early-age shrinkage in steel-concrete composite beam-like structures and its effects on the overall composite cross-section.

1 Introduction Structural health monitoring is an important engineering approach that enables the non-destructive assessment of monitored structures. In the past few decades, the United States has been experiencing a growing problem with the ageing of the transport infrastructure. In 2013, some 11 % of all bridge structures in the United States were rated to be structurally deficient [1]. With the availability of infrastructure funding being limited, there has been an increasing need for the development of efficient and effective structural health monitoring methods to aid the process of repairing and replacing existing structures. According to statistics from the US Department of Transportation, 60 % of the transportation network in the United States is made up of beam bridges [2]. Composite beam bridges, such as steel-concrete composite structures, are one of the most common types of beam structure in the United States.

Hence, it is vital to develop and improve structural health monitoring methods that enable accurate assessment of structural behaviour in composite beam structures.

Specifically, this research aims to assess the early-age shrinkage effects in the strain-based monitoring of * Corresponding author: bglisic@Princeton.edu

steel-concrete composite beam-like structures. As a structural material, concrete is subjected to long-term shrinkage due to the reduction in volume that begins shortly after casting [3]. In steel-concrete beam-like structures, concrete shrinkage could affect the strain distribution in the overall composite structure. It is commonly assumed that a structure has close to zero deformations in its initial state prior to any external loading. However, for steel-concrete composite structures, factors such as early-age effects in concrete can prompt non-zero initial strain conditions in the structure almost immediately after construction [4].

The general trends in the long-term shrinkage of concrete have been studied in depth in the literature and incorporated into design standards [5]. However, strains due to early-age shrinkage in steel-concrete composite beams within the first few days after casting, which could induce bending and axial strains in the structure, have not been well modelled [5]. In the context of structural health monitoring, it is of great interest to identify and quantify the effects of early-age shrinkage in order to improve our understanding of the behaviour of the monitored structure and create a reference point for evaluating its structural condition.

For a newly cast steel-concrete composite beam resting on a horizontal surface (e.g. on the ground) and not fully restrained, early-age shrinkage effects in the concrete slab create bending and upward lifting at the two ends of the structural member in the longitudinal direction, resulting in a “double-cantilever” effect [6], as shown in Fig. 1.

This upward deflection then induces partial activation of the self-weight of the structure, which lowers the ends of the structure towards the ground. Although these deformations are relatively small in magnitude and unnoticeable to the eye, they result in non-zero axial and bending stresses and strain distributions prior to any external loading. It should be noted that for beam structures that are simply supported, fixed at the two ends or with other typical boundary conditions, the self-weight of the structure is immediately activated after pouring the concrete. However, for a beam structure that is fully or partially supported by the ground, the behaviour of the structure in terms of deformation, stress and strain becomes more complex, as the amount of dead load that is applied to the structure depends on the upward lift of the ends of the structural member.

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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Fig. 1. Composite beam structures under the influence of shrinkage-induced initial strains [6]

Fig. 2. Installation of long-gauge fibre optic sensors on steel girder (left) and in concrete slab (right)

The goal of this research is to develop a practical method for evaluating initial strains and stresses due to early-age shrinkage in steel-concrete structures which can be used universally in the structural health monitoring of composite beam-like structures. The research presented combines physical tests on a scale-model test structure in close-to-real conditions and numerical simulations with 3D finite element modelling. The first part of the analysis consisted of finding the initial strain state due to early-age shrinkage through elastic strain measurements from static loading events and finite element simulations. Next, a simplified analytical approach was employed to determine the reasonable range of early-age shrinkage and associated deflections. Finally, finite element modelling was used to simulate the behaviour of the test structure under shrinkage effects, and quantify the magnitude of shrinkage in the concrete slab which best approximates the observed behaviour of the test structure.

2 Test structure and monitoring methodology The model structure used for the evaluation of early-age shrinkage was a multi-girder steel-concrete composite beam created as part of the Automated Non-Destructive Evaluation and Rehabilitation System (ANDERS) programme by Rutgers University. The model structure had a span of 30 ft. (9.14 m) and a width of 12 ft (3.66 m). It consisted of an 8 inch (0.20 m) thick concrete slab supported by three W16w57 steel girders and was monitored by three sets of long-gauge fibre optic strain and temperature sensors. Each sensor set included four strain sensors with a

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gauge length of 19.7 inch (0.50 m) placed in parallel, two mounted on steel girder flanges using aluminium angles, and two attached to the slab reinforcement prior to casting of the concrete slab. Fig. 2 shows the installation of the monitoring system.

The sensor network was installed during the construction of the model structure. One of the goals of the ANDERS project was to use the monitoring system to detect prescribed minute damage inside the concrete slab [7].

Therefore, sensors were placed near two locations of minute damage and one healthy reference location. The arrangement of the monitoring system with respect to the cross-section of the model structure is shown in Fig. 3.

Based on uncertainty analysis and manufacturer’s specifications, the limits of error of the monitoring system are evaluated to t5.1 RJ [8].

To avoid any ambiguity, it should be noted that for the purpose of studying the effect of early-age shrinkage, this project focused on the monitoring data at the healthy reference location (girder F). The test structure was monitored during a series of static and dynamic loading events.

For the scope of the current research, the monitoring data from two static loading events – seven days after the casting of the concrete slab – were studied: – Load case 1: Structure lifted at two intermediate jacking locations – Load case 2: Structure placed on temporary wooden supports (simply supported) After the removal of the formwork, the test structure was placed on level ground, where it experienced the combined

X. Li/B. Glisic · Evaluating early-age shrinkage effects in steel-concrete composite beam-like structures

Fig. 3. Sensor positions on test structure [6]

Fig. 4. Evolution of elastic strain over time and extraction of static loading events [9]

influence of the “double-cantilever” effect and partial activation of self-weight (see Fig. 1). The evolution of elastic strain over time at girder F during the monitoring session, as well as the extraction of the two static load cases [9], are shown in Fig. 4.

As strain sensor readings shown in Fig. 4 are relative to the reference point when the structure was placed on the ground, the sensor readings Jsensor, the strains due to loading Jdue to loading and shrinkage-induced initial strains Jinitial condition have the following relationship [6]: εsensor = εdue to loading – ε initial condition

concrete or attached to steel girders with aluminium angles). All surface connections were modelled using surface tie interactions, which prevents any relative movement of the nodes on either surface. The Young’s modulus of the concrete slab seven days after casting was estimated based on compressive strength cylinder tests performed in parallel with the curing of the concrete slab. The empirical equation from ACI 318-08 for estimating Young’s modulus for normal-weight concrete was used [8], as shown in Eq. (2), where fce represents the compressive strength of the concrete specimen in psi [10]:

(1)

Based on the integrated method, the strains due to loading could be evaluated using finite element simulations of the two static load cases. Thus, the initial strains due to early-age shrinkage, which are the only unknown variables, can be calculated from Eq. (1).

3 Creation and validation of numerical model To understand the anticipated strain distributions of the test structure during the two static load cases, a 3D finite element model was created with ABAQUS CAE. Eightnode solid elements were used for the modelling of the concrete slab and steel I-girders, and steel reinforcement in the concrete slab was incorporated as 2D embedded wire mesh. Intermediate steel diaphragms were modelled using 2D shell members, and strain sensors in the monitoring system were simulated as 2D truss elements (embedded in

Ec = 57000 fc′

(2)

The material properties used in the finite element model are summarized in Table 1.

The FEM-simulated sensor strains were compared with actual measured strains at girder F for both load cases, and the resultant strain profiles are illustrated in Fig. 5.

Table 1. Finite element model material properties [9] Material

Concrete

Steel

Aluminium

Density (kg/m3)

2400

7800

2799

Young’s modulus (MPa)

34 000

203 000

69 000

Poisson’s ratio

0.2

0.3

0.33

Material strength (MPa)

46.8

400

200

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Fig. 5. Elastic strain profiles from actual sensor measurements and finite element simulations [6]

Fig. 6. Early-age shrinkage-induced strain profile at girder location [6]

Fig. 5 shows that for both simulated and actual sensor measurements, the strain profiles over the depth of the cross-section appear to have strong linear trends, which confirms the assumed healthy composite behaviour based on linear theory. It can also be seen that the magnitude of change in strain between load cases 1 and 2 simulated by the finite element model agrees very well with actual sensor measurements, which further validates the material stiffness definition in the finite element model. For each of the two load cases, the FEM-simulated strains are lower than the measured sensor strains by approx. 25–30 RJ near the bottom of the cross-section and 0–5 RJ near the top of the concrete slab. This consistent difference can be attributed to the combined effect of early-age shrinkage and partial activation of self-weight, as explained previously with Eq. (1). Based on these results, the average strain profile due to early-age shrinkage can be extracted, as shown in Fig. 6, where consistent values at all three sensor locations can be seen (including the two locations with artificially created damage). In order to evaluate the early-age shrinkage that caused this

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strain distribution, a simplified analysis was performed to obtain a first approximation of the range of the concrete shrinkage. Finite element analysis was performed subsequently to estimate the magnitude of the shrinkage and the resultant strain distribution.

4 Simplified analysis for early-age shrinkage In a typical steel-concrete composite beam structure where the geometric and material stiffness properties of all the structural components are known, a simplified numerical method can be created to estimate the composite action due to early-age shrinkage based on a set of compatibility equations. This simplified analysis assumes that the effect of early-age shrinkage mostly affects the strains in the longitudinal direction, and the resultant bending and axial strains have linear distributions.

As illustrated in Fig. 7, prior to any composite interaction, the concrete slab and steel girders can be treated as independent structural elements. The concrete deck is then free to deform in the longitudinal direction due to

X. Li/B. Glisic · Evaluating early-age shrinkage effects in steel-concrete composite beam-like structures

Fig. 7. Typical steel-concrete composite beam structure [6]

early-age shrinkage effects to a shorter length Lc, while the steel girders remain undeformed with a length Ls. In order to establish compatibility at the boundary surfaces between the concrete slab and the steel girder, normal forces and bending moments resulting from stresses generated at extremities of the beams are applied to the ends of the steel-concrete interface, denoted as N and M in Fig. 7.

After casting, the steel and concrete elements have the same length. Thus, free shrinkage in the concrete slab can be defined as εshrinkage =

Ls − Lc Ls

(3)

To achieve compatibility in the composite structure, the following two criteria must be met: a) the lengths of the deformed concrete slab and steel girders at the interface are equal: Ldc = Lds = Ld , see Eq. (4), and b) the curvatures of the deformed concrete slab and steel girders at the interface are identical: κ dc = κ ds = κ d [6], see Eq. 5.

1+

N M N M e )L e )L = (1 − − − EsA s EsIs s s EcA c EcIc c c

Nec – M Nes + M = EcIc EsIs

(4)

(5)

In Eqs. (4) and (5), EiAi and EiIi are the axial and bending stiffness of each structural element respectively (i " c for concrete and i " s for steel). The variable ei represents the eccentricity between the centroid of each structural element and the composite interface. As the geometrical properties and Young’s moduli of the steel and concrete elements could be estimated, the compatibility equations contain two unknown variables: N and M. Since the two equations are independent, the values of both variables can be solved to achieve unique solutions. The vertical deflection at either end of the structure due to the “double-cantilever” effect, I, can also be calculated: δ = R − R * cos(

Ld /2 ) R

(6)

where R is the radius of curvature of the overall composite section:

R=

1 κd

(7)

Based on the above analytical expressions, if the hypothesized scenario that the test structure is under partial activation of self-weight is to be valid, then the upward deflection I due to shrinkage of the concrete slab should be ! 1.417 mm.

This number corresponds to an imposed shrinkage of 0.0024 % (24 RJ), a reasonable amount for normal-weight concrete after seven days [11]. In general, when the geometrical and material properties of a composite structure are available, this simplified analytical method can be used for the effective prediction of the range of curvature P and deflection I in a composite structure induced by early-age shrinkage due to the “double-cantilever” effect.

5 Numerical simulation of early-age shrinkage in composite structures A finite element analysis was performed to simulate the behaviour of the test structure under the combined influence of early-age shrinkage in the concrete deck and selfweight. The goal of this analysis was to estimate the amount of early-age shrinkage in the concrete slab which would induce the strain distribution observed from actual sensor data shown earlier in Fig. 6. As a better representation of the boundary conditions of the test structure in the realistic situation, a non-linear contact problem was created between the test structure and the ground surface [6].

In the finite element model, the level ground was simulated using an analytical surface in ABAQUS CAE, and initialized to be in full contact with the bottom of the test structure. The boundary condition allows the test structure to lift away from the ground surface, but allows no sinking or penetration into the plane. The simulation included four loading steps: 1. Initializing contact between ground surface and test structure 2. Imposing shrinkage strain on concrete slab 3. Initializing monitoring of bending strains and stresses in the test structure 4. Applying self-weight to the test structure The simulated deformations of the test structure under the applied loading steps are shown in Fig. 8.

For illustrative purposes, a series of five values of imposed strain in the concrete slab were considered, and the

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X. Li/B. Glisic · Evaluating early-age shrinkage effects in steel-concrete composite beam-like structures

Fig. 8. Finite element simulation of “double-cantilever” effect and activation of self-weight [6]

Fig. 9. FEM-simulated initial strains due to early-age shrinkage of concrete slab [6]

corresponding elastic strains at the top and bottom sensor locations at girder F are summarized in Fig. 9. For each value of imposed strain, all five cases produce compressive strains near the bottom of the steel girders and small tensile strains in the top of the concrete slab, which agrees qualitatively with the measured structural behaviour due to the “double-cantilever” effect and partial activation of self-weight shown earlier in Fig. 6. According to the results, strains induced due to early-age shrinkage at sensor locations are linearly proportional to the value of imposed strain in the concrete slab.

To approximate the shrinkage strains experienced by the actual structure, simple linear regression was used to construct the initial strain profile at girder F and estimate the magnitude of initial strain experienced by the structure at the top and bottom sensor locations. Furthermore, the two shaded areas in Fig. 9 represent the limits of errors of the physical monitoring system (t5.1 RJ). Based on this analysis, a magnitude of imposed shrinkage of 23 RJ, as marked by the orange line in the figure, is the best estimate of the imposed shrinkage strain in the physical structure.

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This result corroborates very well with the simplified numerical analysis.

6 Conclusions Early-age shrinkage in composite steel-concrete beam-like structures can lead to partial activation of the beam’s selfweight, and create non-zero initial conditions in the overall composite structure. A deeper understanding of this phenomenon enables better comprehension of the early-age behaviour of composite structures, which is especially important for the purpose of strain-based structural health monitoring. The goal of the current research was to develop an efficient, widely applicable analysis method that couples the physical monitoring system with numerical simulations to identify and measure strains created by concrete early-age shrinkage.

The strain monitoring data from two static loading events of a model test structure were extracted, and simulations with a 3D finite element model of the test structure were implemented. A simplified analytical method was de-

X. Li/B. Glisic · Evaluating early-age shrinkage effects in steel-concrete composite beam-like structures

veloped to estimate the range of early-age shrinkage and relate the magnitude of concrete shrinkage to the deformations of a steel-concrete composite beam. Furthermore, a series of finite element analyses was performed to simulate the combined influence of the “double-cantilever” effect and partial activation of self-weight. Based on the analysis results, the process presented was able to quantify strains induced by early-age concrete shrinkage in a steel-concrete composite structure effectively. It should be noted that for the test structure studied for this project, the magnitude of early-age shrinkage-induced strain was up to 30 RJ, a significant value that constitutes 27 % of the magnitude of maximum bending strain due to the self-weight of the structure, which emphasizes the importance of accurately identifying this phenomenon. Future research should be performed on steel-concrete structures of other types and scales to improve further the understanding of the behaviour of composite beam structures in the presence of early-age shrinkage of concrete components in general.

Acknowledgements The material presented in this paper was supported by the USDOT-RITA UTC programme (grant No. DTRT12-GUTC16), facilitated by the Centre for Advanced Infrastructure and Transportation (CAIT) at Rutgers University and the National Science Foundation (grant No. CMMI1362723). Special thanks to Dr. Dorotea Sigurdardottir for her help with starting the current research. The authors would like to acknowledge the assistance of Prof. Nenad Gucunski and his research group at Rutgers University for their collaboration and support on the testing of the ANDERS slab.

Disclaimer The opinions, findings, conclusions and suggestions made or implied by the authors represent the views of the authors only. They do not reflect the opinions of the USDOT-RITA UTC programme, the Centre for Advanced Infrastructure and Transportation at Rutgers University or the National Science Foundation.

References [1] American Society of Civil Engineers: 2013 Report Card for America’s Infrastructure. http://www.infrastructurereport card.org/a/#p/home.

[2] US Department of Transportation: Federal Highways Administration 2013, Structure Type by State National Bridge Inventory (NBI). www.fhwa.dot.gov/bridge/struct.Cfm. [3] Mokarem, D. W.; et al.: Development of Performance Specification for Shrinkage of Portland Cement Concrete. Transportation Research Record 1834, Paper No. 03-2420, 2003, pp. 40–47.

[4] Holt, E.; Janssen, D.: Influence of Early Age Volume Changes on Long-Term Concrete Shrinkage. Transportation Research Record, Journal of the Transportation Research Board 1610(1), 1997, pp. 28–32.

[5] ACI Committee 209: Prediction of Creep, Shrinkage, and Temperature Effects in Concrete Structures. American Concrete Institute, 2008 (reapproved 2008).

[6] Li, X.: Validation of the Method of Neutral Axis for the Identification of Minute Damage in Composite Beam Structures.

M.S.E. thesis, Princeton University, Princeton, NJ, 2016.

[7] Sigurdardottir, D. H.; Glisic, B.: Detecting minute damage in beam-like structures using the neutral axis location. Smart Mater. Struct. 23 (2014), 125042.

[8] Sigurdardottir, D. H.; Glisic, B.: Neutral axis as damage sensitive feature. Smart Mater. Struct. 22 (2013), 085030.

[9] Li, X.; Glisic, B.: Finite Element Verification of the Method of Neutral Axis for Damage Detection in Composite Beam Structures. 8th European Workshop on Structural Health Monitoring, 2016.

[10] American Concrete Institute: Building Code Requirements for Structural Concrete (ACI318-08) and Commentary. Farmington Hills, MI, 2008, (an ACI Standard).

[11] Almeida, J. F.; et al.: FIP Practical Designs of Structural Concrete. FIP Recommendations, 1999.

Keywords: structural health monitoring; steel-concrete composite structure; fibre optic sensors; early-age shrinkage; finite element analysis; strain-based monitoring

Authors Xi Li Master of Science student Dept. of Civil & Environmental Engineering Princeton University 100 Christopher Columbus Drive, Apr 2604 Jersey City, NJ 07302 USA Branko Glisic Associate Professor Dept. of Civil & Environmental Engineering Princeton University E330 EQuad Princeton, NJ 08544 USA

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Articles Ingbert Mangerig* Robert Kroyer Matthias Koller

DOI: 10.1002/stco.201710010

Experimental and numerical analyses of the effectiveness of high-frequency peening processes To assess fatigue behaviour, experimental tests were conducted on material specimens furnished with hammer peening tracks, with two different types of treatment being employed. The results of fatigue tests on treated specimens were compared with those of untreated specimens and assessed. Non-linear analyses with the FE program ADINA, taking into account the stress–distortion behaviour of a material subjected to fatigue, were conducted as part of the test concept, specimen analysis, evaluation of the test results and probing the mode of action of both techniques.

1 Introduction Processing technology for metal components includes a variety of known surface strengthening procedures aimed at increasing the stress capacity and service life of a workpiece. Many of the structural components requiring improvement are subjected to dynamic loading and thus exposed to cumulative damage due to cyclic effects – the consequence being fatigue fracture. In the field of mechanical engineering, shot peening is the most common method of improving fatigue strength. In the broadest sense, thermal or mechanical treatments such as deep rolling, familiar from reshaping technology, as well as a combination of thermal and mechanical processes can be considered as specific measures to improve a component’s surface characteristics and thus fatigue strength.

Improvements brought about by plastic deformation involve inducing residual stresses near the surface and alterations to the crystalline structure. Therefore, one issue requiring examination, particularly under oscillating loads, is the magnitude of the induced residual stresses as well as their dissipation through relaxation or, additionally, through degradation.

In the case of metals, plastic deformations are caused by lattice defects, or dislocations. If the loads exceed the bonding forces in the metallic elementary cells, the dislocations travel along predefined gliding planes, thus causing irreversible distortion. Pronounced changes in plastic behaviour are used to characterize the strength properties of metals. Dislocations are therefore not flaws, but welcome tools for controlling the stress–distortion behaviour of metals. Selectively induced plastic deformation alters the stress–distortion behaviour and strength characteristics. As * Corresponding author: ingbert.mangerig@unibw.de

54

these effects are achieved by dislocation movements, the effects can also be lost again through relaxation if the positions of the dislocations are not permanently blocked. In principle, plastic deformations are usually accompanied not only by increases in strength, but also by limits to ductility.

One and the same material can exhibit different strength and deformation properties due to different lattice errors and differing dislocation concentrations. One example that could be cited is the weld seam transition including regions of increasing hardness due to fast cooling from the heat of fusion. However, normalized and tempered zones with differing strength characteristics due to heat input from subsequent weld layers also arise within weld seams.

For some years, high-frequency peening techniques have been used to increase the fatigue strength of welded steel structures in structural engineering. Weld seam transitions undergo plastic deformation in this case. In these zones, notches in the weld seams are levelled, the material is strengthened and residual stresses are induced. Two techniques based on different principles have essentially prevailed for the treatment of weld seams [1], [2].

The different operating principles of both techniques are also evident in the surface texture of the induced plastic deformations. Fig. 2 shows the treatment tracks of both techniques. To work out differences and identify changes in the metallic structure, the treatment tracks were incorporated not at the borders of weld seams, but on metal surfaces. The intention was to avoid overlapping effects from heat input during weld seam production and alterations to the material in the treatment track.

Differences [3] are not only evident in the form of the treatment track, but also in the fatigue behaviour of structures whose weld seam transitions were treated using these techniques. This differing behaviour was the motivation for the findings presented in the scope of this publication to evaluate the effectiveness of high-frequency treatment techniques. To analyse the specific characteristics of both techniques, experimental investigations were carried at the Institute of Structural Engineering, University of the German Armed Forces Munich [4]. The evaluations had to pay special attention to the original state of residual stress in the base material as well as that in the weld metal.

Both techniques are based on the idea of influencing fatigue strength positively through plastic deformations at

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 1. Processing weld seam transitions to improve fatigue strength

Fig. 2. Treatment tracks of two techniques on the surface of sheet steel of grade S355J2

the weld seam transition – an idea that appears very plausible, despite some unresolved issues concerning the reliability of weld seams treated using hammering techniques when interpreting test results from real structures. An objective consideration of the topic reveals that both techniques involve a very complex interaction between practical implementation, material and solid mechanics and measures to ensure the quality of implementation. To help address issues relating to material behaviour and mechanical stress, fatigue tests [4] were conducted and their results evaluated by means of numerical analyses carried out using the FE program ADINA [5], [6].

tions in the region of the treatment tracks. Isotropic strengthening of the base material can be expected to occur near the surface, so that the transition to the plastic zone should lie above the yield point of the untreated base material. Fig. 3 shows the stress–cycle relationships for the tests conducted and the resultant, approximated S/N curves. A representation of notch case 160 as defined per DIN EN 1993-1-9 is included for the purpose of comparison. Although only a few tests have been conducted so far, the effects observed have been explained through the identification of different survival probabilities. The diagrams in Fig. 3 also show the results of the two fatigue tests conducted on non-welded specimens.

2 Fatigue tests Surface tracks as preparation for fatigue tests were made using treatment techniques 1 and 2 on test specimens made of sheet steel grade S355J2 whose surfaces were free of rolling residue and incipient rust. The test specimens were subsequently subjected to a pulsating tensile stress by a controlled force up to the point of complete fatigue failure. For the purpose of comparison, specimens without a treatment track were prepared from the same steel consignment and subjected to the same pulsating tensile stress.

Based on the hammering technique, it was expected that the strength of the treated materials would be raised because the tests [4] were conducted at maximum stress limits near the yield point. Fig. 2 shows the plastic deforma-

3 Interpretation of scanning electron microscope investigation The aim of this investigation was to analyse and to obtain a better understanding of two different methods of ultrasonic impact treatment methods as post-processing for weld seams. The first step in the evaluation process was to discuss the fracture surfaces with respect to the micromechanical structure after post-treatment, cyclic loading and rupture. Figs. 4 to 6 present the fracture surfaces of all test specimens for techniques 1 and 2 as a function of the applied upper stress level Xo in the stress relation R ~ 0.1.

The comparison of the fracture surfaces in Figs. 4 to 6 shows the typical fine fatigue striations in the grain struc-

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 3. )X-N relationships of the high-frequency peened test specimens for both treatment techniques in comparison to Notch Case 160 of DIN V ENV 1993: left – overview, right – extract

Fig. 4. Surfaces of specimens Z1, Z4, Z6, Z7, Z9 and Z10 for the applied stress level Xo " 383 MPa

Fig. 5. Fracture surfaces of specimens Z2 and Z5 for the applied stress level Xo " 420 MPa

Fig. 6. Fracture surfaces of specimens Z3 and Z8 for the applied stress level Xo " 450 MPa

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 7. Fracture surfaces of specimens Z1 and Z7 for the applied stress level Xo " 383 MPa (view normal to fatigue cycle loading direction)

ture as well as the very coarse microcracks due to cyclic loading. Although a systematic failure mechanism cannot be deduced, a coarse tendency for the different failure characteristic between the two high-frequency peening techniques can be observed in the fracture surfaces presented in Figs. 4 to 6. The fracture surfaces after the treatment using method 2, i.e. Z1_Method 2 (Fig. 4), Z2_ Method 2 (Fig. 5) and Z3_Method 2 (Fig. 6), show significant vertical crack initiations along the rupture surface. In comparison to method 2, the fracture surfaces of high-frequency peening technique method 1, i.e. Z6_Method 1 (Fig. 4), Z5_Method 1 (Fig. 5) and Z8_Method 1 (Fig. 6), the tendency is the creation of significant local fine grain structure cells. The most expected and typical microstructure can be observed in the Z9_S355J2 and Z9_S355J2 “non-high-frequency peening treatment” test specimens.

Already, the coarse evaluation of the pictures (Figs. 4 to 6) shows that differences between high-frequency peening processes must exist.

From a view normal to the fracture surface, see Fig. 7, it can be assumed that test specimen Z1-Method 2 fails very coarsely in the depth direction, whereas test specimen Z7-Method 1 fails harmonically in the depth direction.

For reference, the probable failure mechanism of the non-high-frequency peening treatment test specimen is presented in Fig. 8. The crack initiation starts in zone 1 and propagates to zone 3 accompanied by fatigue striations of an elliptical shape with an increasing number of cracks up to a thin, violent failure region.

In comparison to the fracture surface of specimen Z9, the “non-high-frequency-treated” specimen, presented in Fig. 8, a fracture surface processed using high-frequency peening technique method 1, see Fig. 9, will be discussed in the following. Crack propagation in an elliptical shape with very fine fatigue striations was also observed; however, these damage regions are very small, local and separate. It is assumed that local residual stress states based on the normal high-frequency hammer may be responsible for the nucleation of these failure mechanisms. Zooming one of these fracture zones in a higher resolution, an almost radial grain structure with respect to the probable crack initiation origin was also observed, see Figs. 9d and 9e.

Finally, the fracture surface effects of fracture Z4 posttreated with high-frequency peening technique method 2, presented in Fig. 10, will be discussed. It is obvious that there are also local elliptical cells with the fine fatigue striations; however, the cells seem to be slightly larger than those arising from high-frequency peening technique method 1. This may be not an issue but could be investi-

gated in further studies. However, some significant aspects must definitely be investigated in more detail, namely: a) the vertical sharp cracks in the local elliptical damage cells, b) the smeared material layers closed to the specimen outer surface, and c) the very non-homogeneous surface layer grain structure in comparison to the inner material layers, which looks like a delamination of the outer skin layer.

Further, in Fig. 10d, from the grain structure it looks like a material flow mechanism has occurred accompanied by long layered crack initiation.

Summarizing the discussion of the damage mechanism presented in Figs. 8, 9 and 10, differences in the failure and damage mechanisms of the high-frequency peening post-treatment methods investigated are obvious.

These differences can be also recognized in the fatigue tests at macromechanical level. The root cause, however, might be difficult to pinpoint because it may be an interaction of micro-meso-macro effects with multi-scale phenomena. As a first guess, the movement of the high-frequency post-treatment tool may be one reason for differences in the damage profile, also as a function of heat entry and local pressure.

4 Finite element analysis To determine the stresses in the regions of the material specimens treated by means of techniques 1 and 2, individual steps in the hammer treatment were simulated within the scope of FEM analyses. These calculations are intended to reveal deformations and stresses forming part of the elastic/plastic behaviour as a function of the hammering geometry.

Non-linear finite element analyses were performed using the FE program system ADINA [5], assuming an elastic/plastic material law, large deformations, large strains and 3D contact conditions. The simulations took place on the basis of assumptions derived from available process descriptions.

4.1 Modelling FE modelling was chosen to illustrate the different characteristics of both techniques, to observe the stress effects of the hammering techniques and to compare the simulation results obtained from both the techniques investigated.

The main modelling parameters were: specimen sheet

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 8. Fracture surface Z9 “non-high-frequency-treated” specimen, failure mechanism

thickness t " 15 mm, specimen cut-out width b " 6 mm, hammer diameter d1 " 4 mm, d2 " 3 mm and rounding radius r1 " 1 mm. In both models, the test specimens underwent plastic deformation by setting the hammer profile’s static displacement to uz " 0.12 mm in the direction of the positive Z axis.

4.2 Material law The material behaviour of the test specimen was described by a multi-linear material model, the cyclic material behaviour of steel S355J2 serving to supply the characteristic values assumed for the stress–distortion behaviour, see Fig. 12.

Material hardening for cyclic loading was described using the approach of Armstrong-Frederick [6]. The hardening parameters selected were Xy " 660 MPa, h " 1420 MPa and _ " 2.0. This material behaviour was cho-

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sen to simulate the altered strength characteristics when subjected to hydrostatic pressure. To evaluate the influence of material parameters as part of the theoretical analyses, the yield point, designated Xy in the Armstrong-Frederick approach, and the fracture criterion, expressed as the limit Jpl, were varied. The variations in material parameters show that, depending on the hammerhead’s geometry and the scattering of the hardening values, the resultant stress–distortion state is characterized by high local plastic distortion, but does not indicate structural failure as a cause.

4.3 Results of FEM simulation The modes of action of both techniques were examined by selecting individual, typical load situations. Load application involved a “quasi-static” stress, and the non-linear, elastic-plastic finite element equations and

I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 9. Fracture surface Z6, method 1 specimen, failure mechanism

corresponding contact problem being solved require the use a full Newton-Raphson iteration [6]. To obtain a representative non-linear elasto-plastic deformation state it was unconditionally necessary to use a material model which allows the simulation of large strain behavior.

The analysis model took into account residual stresses on a scale determined from values measured on individual test specimens. As a simplification, pressure distortions were induced to trigger a residual stress field in a state of equilibrium.

Fig. 2a. To simulate the procedure, a load history comprising the following steps was defined as part of the numerical analyses: 1. Positioning of the contact body at y " 0.0 mm on the Y axis; movement by uz in the Z direction.

2. Lifting of the contact body to z " 0.0 in the Z direction; movement by 1.0 mm along the Y axis; movement of the contact body by uz in the Z direction.

3. Lifting of the contact body to z " 0.0 in the Z direction; movement by –0.5 mm along the Y axis; movement of the contact body by uz in the Z direction.

4.3.1 Model 1, treatment technique 1

Fig. 13 shows a compilation of the individual steps in the load history and the corresponding contact situations and deformed specimen surfaces. The thickness of the FE discretization shown in this figure is 1.00 mm in the contact situation and approx. 0.34 mm in the deformed specimen surface.

A process control that sets neighbouring, overlapping pits generated by plastic deformation was assumed for the weld seam treatment method designated technique 1. This was meant to approximate the treatment track shown in

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 10. Fracture surface Z4, method 2 specimen, failure mechanism

4.3.2 Model 2, treatment technique 2 In contrast to technique 1, simulation calculations for the weld seam treatment method designated as technique 2 assumed a process control in which the hammerhead is initially pressed into the material and not subsequently lifted, instead moved through the material in millimetre steps. This was intended to approximate the treatment track shown in Fig. 2b. To simulate the procedure, a load history comprising the following steps was defined as part of the numerical analyses: 1. Positioning of the contact body at y " 0.0 mm on the Y axis; movement by uz in the Z direction.

2. Movement of the contact body along the Y axis by 1.0 mm, the penetration depth uz remaining constant.

3. Movement of the contact body along the Y axis by a further 1.0 mm, the penetration depth uz remaining constant.

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Steel Construction 10 (2017), No. 1

Fig. 14 shows the results of the numerical analysis as the steps in the load history with the contact situations described previously and the associated, deformed specimen surfaces.

4.3.3 Comparison of the simulation results obtained with treatment techniques 1 and 2 An evaluation of the analysis results makes it clear that the stress–distortion fields arising depending on the technique employed cannot be compared with each other. Possible at best is a comparison of the analysis results for the initial loading conditions. Every further stress–distortion state depends on the load history. Furthermore, evaluation with regard to the comparative stresses according to von Mises does not necessarily fulfil the objective of assessing significant stress–distortion states at the surface because compressive and tensile behaviour must be considered equally

I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 11. Finite element models for numerical analysis using different hammerhead profiles

Fig. 12. Stress–distortion relationship for steel grade S355J2 – cyclic material behaviour

when evaluating and rating results. Representative for a comparison of the techniques, therefore, are selected results for three load times ti and the corresponding relief periods ti Y. The stress–distortion state at time ti represents the relevant and critical time of failure of the specimen material.

However, the stress–distortion state at time ti Y is critical for subsequent kinematic hardening and fatigue behaviour.

The time increment Y chosen as 0.05 defines the chronological process between the maximum indentation by the hammer head, relief of the pressure point and movement of the hammerhead to the new starting position in technique 1, whereas the hammerhead in technique 2 is driven essentially with minor relief in the direction of the hammerhead track.

A graphical interpretation of the results makes it clear that both treatment techniques induce a complex, multi-axial state of stress that is determined by the load

history. By means of slight variations in material characteristic values, especially through the choice of the elastic-plastic failure criterion max. Jpl, typical centres of possible cleavage fracture can be demonstrated for both techniques: at the edge and in the centre of the spherical cap for technique 1 and along the top seam edge for technique 2. However, no rating criterion can be derived directly from this because of the state of high hydrostatic stress.

Stress tri-axiality [7] is therefore resorted to for a further condensation of the calculation results because stress conditions with identical main normal stresses at crystalline level do not result in sliding processes, so tensile stress on all sides can lead to cleavage fracture without deformation. A comparison of different stress states taking into account the size of the hydrostatic component therefore provides insights into possible failure mechanisms. Suitable for comparison here is stress multi-axiality, defined as the quotient of hydrostatic stress and the comparative stress according to von Mises [8]. The magnitude of the multi-axiality indicates whether ductile material behaviour is still possible. According to [9], stress multi-axiality, in addition to plastic distortion, is a critical variable for damage assessment.

The comparison shows that the analysis parameters chosen in technique 1 result in a prevalence of a nearly constant state of hydrostatic compressive stress in the treated area, while technique 2 results in narrow-bandwidth tensile stresses in the zone trailing the hammerhead.

Simplifications would have to be introduced in an effort to answer questions related to stress states in areas containing weld seam transitions treated with hammering techniques. Too many effects would have to be considered if the dynamic treatment process had been represented in detail. Accordingly, individual stress states in the production process were considered successively, taking into account the load history. The findings obtained therefore al-

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 13. Results of simulations on model 1 for treatment technique 1 with contact situations and associated deformed structure

Fig. 14. Results of simulations on model 2 for treatment technique 2 with contact situations and associated deformed structure

low an initial assessment of potential stresses in the region of the induced plastic deformations, but without being able to confirm reliably the distortions and stresses identified.

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Steel Construction 10 (2017), No. 1

Accordingly, the results of FE analyses for both treatment techniques indicate stresses that are very large – larger than the single-axis yield and fracture stresses of a

I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 15. Load state t1 " 1.0, maximum main stresses X1 and maximum main distortions J1: a) technique 1 and b) technique 2

Fig. 16. Load state t1 Y " 1.05, maximum main stresses X1 and accumulated plastic distortions J pl: a) technique 1 and b) technique 2

Fig. 17. Load state t2 Y " 2.05, maximum main stresses X1 and accumulated plastic distortions J pl: a) technique 1 and b) technique 2

grade S355 steel, but which, in analogy to the Hertzian pressure and the associated hydrostatic stress state, need not necessarily be included in a failure assessment. The characteristic distortions not only demonstrate plastic deformation, they also reveal a significant margin to the elongation at break of the steel considered. The FE analyses conducted confirm that it is possible to accomplish the objective of surface hardening through plastic distortion, as pursued by both treatment techniques. The calculated stresses and deformations result in an expected distribution of compressive and tensile stress for the initial load in each case, designated as time/load step 1.00 in the representations. The illustrations show that an internal equilibrium of stresses and forces is generated which is adequate for the deformation state. The results of the other load steps, identified here as time/load steps 2.00 to

3.00, depend on the technique. For a subsequent fatigue analysis, the cumulative stresses in individual steps represent the critical stress and distortion states. By means of additional simulation calculations, and considering results described previously, it was possible to demonstrate the ability to absorb the relatively high stress amplitudes in experiments described previously. This makes it possible to derive findings concerning the key characteristics of the techniques examined from fatigue tests conducted at relatively high stresses and a small number of load cycles, as in the case under consideration. What has yet to be clarified is why large-scale experiments involving treatment of weld seam transitions in some cases revealed a significant, incomprehensible scattering of load cycle counts. It is conjectured that plastic distortion could lead to a degradation of the intended positive effects of treat-

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

Fig. 18. Load state t3 Y " 3.05, maximum main stresses X1 and accumulated plastic distortions J pl: a) technique 1 and b) technique 2

Fig. 19. Load state t3 Y " 3.05, von Mises comparative stress XV and hydrostatic stress state XH : a) technique 1 and b) technique 2

Fig. 20. Load state t3 Y " 3.05, stress tri-axiality XH/XV: a) technique 1 and b) technique 2

ment. Further investigations were conducted against this background, which are described here only in abridged form.

To analyse possibly restricted ductility, comparative investigations were performed under the condition that a high value of stress tri-axiality can be used as a measure of the tendency towards brittle cracking of a material; extracts from the results obtained are shown in Figs. 20a and 20b. The illustrations show that, compared with treatment method 1, the treatment method designated 2 here involves a stress tri-axiality that is twice as high, so the sensitivity to brittle fracture is also twice as high. The distribution of the stress tri-axiality is striking, too. Evident in the case of the results shown for technique 1 in Fig. 20a are high values of stress tri-axiality on the top edge in the cen-

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tre of the spherical cap, with a low penetration of approx.

0.1–0.2 mm. Fig. 20b shows a much more heterogeneous characteristic for technique 2. The values of stress tri-axiality are not only significantly higher and positive without exception, but also exhibit a significant penetration at the edges of the treatment track, thus suggesting a significant stress concentration. Assuming the conditions described previously are fulfilled, these effects might be responsible for different fatigue behaviour of welding seam transitions depending on the high-frequency peening technique.

5 Summary For some years, high-frequency peening techniques have been used to increase the fatigue strength of welded steel

I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

structures. Weld seam transitions undergo plastic deformation in this case. In these zones, notches in the weld seams are levelled, the material is strengthened and residual stresses are induced. Two techniques based on different principles have essentially prevailed for the treatment of weld seams. The observed differences between welded pipe junctions served as a motivation for performing additional studies aimed at identifying sufficiently reliable information concerning stresses in the notch root of treated weld seam transitions.

On the basis of published research results as well as the authors’ own analyses of methods of treating weld seams, the fatigue behaviour of test specimens was examined and compared with treated, but non-welded, specimens. The steel sheets of grade S355J2, used for the presented investigations, were proportionally scaled to apply the peening treatment. To contribute to an understanding of the relationships of welded structures treated with hammering techniques, the distinctive properties of two techniques were analysed. An x-ray procedure was used to measure changes in induced residual stresses during fatigue tests. In view of a need to check for reductions in residual stresses induced by hammering techniques in zones exhibiting partial plastic deformation and, consequently, significant reductions in fatigue strength, the experiments were conducted using load-induced stresses near the yield point. The fatigue tests revealed differences between the fracture surfaces of treated and untreated test specimens. Numerical calculation methods were used to analyse the stresses expected in the case of high-frequency peening techniques.

Although the two techniques for treating weld seam transitions investigated pursue a common objective, a detailed examination revealed significant differences in zones of induced plastic deformation. Residual stress analysis (reserved for a further publication) indicates that the residual compressive stresses hammered in the treatment track’s longitudinal direction are much more pronounced than those aligned perpendicularly in the case of both techniques. The residual compressive stresses induced in the test specimens are different between the two analysis techniques. After being subjected to pulsating tensile loads with maximum stresses near the yield point, a significant reduction in residual compressive stress was measured in the direction of the post-treatment track and a slight increase in the direction of the load. In the case of the untreated specimens serving for comparison, surface measurements indicated nearly identical residual compressive stresses in both directions. After the introduction of uniaxial cyclic stress, again near the yield point, a significant reduction in residual stress in both directions was established, this being particularly pronounced in the direction of action, in some cases with inversion in the tensile range.

Finite element studies conducted parallel to the experimental studies showed that high stresses can occur in the treatment tracks of the test specimens without directly triggering fracture failure. Significantly influenced, however, is the internal stress state that is critical for the fatigue behaviour. The FE studies also showed that failure values of fracture distortion are not attained in the compression zones under the boundary conditions serving as a basis. This result plausibly demonstrates the basic idea and intention of

the techniques, i.e. to use compressive forces to override tensile stresses triggering fatigue cracks. The two techniques in some cases exhibit significant differences in terms of the characteristics and local arrangements of the highly stressed zones. FE studies of the hammering process indicate that one technique essentially generates closed compressive/tensile stress equilibrium concentrically around the hammerhead and the other technique generates closed compressive/tensile stress equilibrium only in the initial loading state, this effect being lost again when the hammerhead is moved. Movement of the hammerhead in technique 2 does not give rise to compressive/tensile stress equilibrium at the hammerhead’s fringes, instead between the frontal wave and the surface layer below the hammerhead’s flat portion. In the region of the frontal wave, one technique generates high compressive stresses which, however, might contribute to triggering microcracks at right-angles to the seam track, as indicated by the fracture images. Under alternating stress, these microcracks can unite to form macrocracks. The fracture images also show that these transverse cracks occur at irregular intervals, thus suggesting a relationship with the induced plastic deformations. In the case of both techniques, FE studies reveal tensile stresses on the component surfaces in the region of the treatment tracks. For reasons of equilibrium, these stresses are unavoidable.

The few tests conducted have already supplied interpretable results for making inferences about the fatigue behaviour of specimens treated with both techniques.

Changes in the initial stress–distortion states of the specimens examined can be observed very satisfactorily from the recordings and analyses of deformation and distortion behaviour using a special measurement method. Changes in distortion depending on load alternations were represented at suitable sectional planes in the region of the seam. Increases in distortion parallel to the direction of stress, which depend on changes in load, are observable here. The evaluation shows changes in distortion in the main direction, with high peaks along the edges of the seam starting from initially low to medium distortions, progressing to a levelling of the distortions at a high average value and culminating in a distinct distortion maximum at the crack plane followed by fracture.

Assuming that distortion due to production-related internal stresses plus load-induced stresses acts in the region of a specimen’s base material during the fatigue test, it can also be presumed that there is lower distortion in the region of treatment at the beginning of the tests due to the counteraction of the distortions induced by the loads – this effect also being desired. However, this advantage is lost in the course of the loading history and the distortions grow steadily.

Acknowledgements The authors would like to thank the staff in the laboratory of Prof. Dr.-Ing. H.-J. Gudladt at the University of the German Armed Forces Munich for implementing investigations with the scanning electron microscope.

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I. Mangerig/R. Kroyer/M. Koller · Experimental and numerical analyses of the effectiveness of high-frequency peening processes

References [1] Weich, I.: Ermüdungsverhalten mechanisch nachbehandelter Schweißverbindungen in Abhängigkeit des Randschichtzustands. Dissertation, Technische Universität Braunschweig, 2008.

[2] Mangerig, I.; Hess, A.: Zur Stabilität von oberflächennahen, durch plastisches Umformen eingeprägten Eigenspannungen.

Stahlbau 83 (2014), No. 4, pp. 236–244. DOI: 10.1002/ stab.201410156 [3] Mangerig, I.; Romen, N.: Ermüdungsverhalten geschweißter Rohrknoten von Fachwerkverbundbrücken. Stahlbau 78 (2009), No. 12, pp. 925–935. DOI: 10.1002/stab.200910108 [4] Koller, M.: Bewertung der Effektivität höherfrequenter Nachbehandlungsverfahren zur Steigerung der Ermüdungsfestigkeit von Schweißnähten im Brückenbau (assessment of the effectiveness of high-frequency treatment techniques for improving the fatigue strength of welded joints in bridge construction). Master’s thesis, Bundeswehr University, Institute of Structural Engineering – Steel Construction, Prof. Dr.-Ing. I. Mangerig, 2014.

[5] ADINA-A: Finite Element Program for Automatic Dynamic Incremental Nonlinear Analysis, ADINA R&D, Inc., 71 Elton Avenue, Watertown, MA 02472, USA. [6] ADINA-A: Theory and Modeling Guide TMG-A_90, ADINA R&D, Inc., 71 Elton Avenue, Watertown, MA 02472, USA. [7] Lemaitre, J.; Chaboche J., L.: Mechanics of solid materials, Cambridge University Press, 1990.

[8] Völling, A.: Berücksichtigung der Dehnungsbehinderung in bruchmechanischen Sicherheitsanalysen (consideration of strain constraints in safety analyses of fracture mechanics).

Diss., RWTH Aachen University, 2009.

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[9] Needleman, A.; Tvergaard, V.: An Analysis of Ductile Rupture in Notched Bars. Journal of Mechanics and Physics in Solids (1984), vol. 32.

[10] Ummenhofer, T.; Herion, S.; Weich, I.: Schweißnahtnachbehandlung mit höherfrequenten Hämmerverfahren – Ermüdungsfestigkeit, Qualitätssicherung, Bemessung. Stahlbau 78 (2009), No. 9, pp. 605–612. DOI: 10.1002/stab.200910074 Keywords: high-frequency peening process; material fatigue; fatigue test; FEM, plasticity

Authors Prof. Dr.-Ing. Ingbert Mangerig University of the German Armed Forces Munich Faculty of Civil Engineering & Environmental Sciences Institute & Laboratory for Structural Engineering Werner-Heisenberg-Weg 39 85577 Neubiberg Germany Dr.-Ing. Robert Kroyer MBDA Deutschland GmbH Senior Expert for Structural Mechanics/Dynamics, System Design Hagenauer Forst 27 86529 Schrobenhausen Germany robert.kroyer@mbda-systems.de Matthias Koller M.Sc. Rübezahlstraße 108 81739 München Germany ibkoller.matthias@t-online.de

Articles Jaap Wardenier* Peter de Vries Gerrit Timmerman

DOI: 10.1002/stco.201710011

Evaluation of cracks in an offshore crane runway girder This paper deals with the evaluation of fatigue cracks in a box type radial crane runway girder with full penetration welds between the web and flange. After 20 years of service fatigue cracks were observed which were initiated in the flange at the toe of the full penetration weld with the web. The observed cracks in the crane runway girder vary in length from a few mm to 330 mm with a summation of the lengths of all observed cracks being 750 mm, on a total length of 56000 mm, thus being only 1.3 %.

The investigation consists of the fatigue analysis described in this paper and additional experimental investigations with scale tests of approximately 1:2 on equivalent I sections with a concentrated load and with a line load to investigate whether cracks stop when they have grown through the residual tensile stress field. For this analyzed crane runway girder with multi-layered full penetration welds and with small cracks at the weld toes in the flange, based on EN 1993-6 the damage would be 1.0 for a design class 98 which is not far from the “class” 92 found by Kuhlmann et al, [1] for tests with rolling wheels, however, related to weld failures from the root of not fully penetrated fillet welds in combination with crack initiations at the weld toe.

1 Introduction This paper deals with the evaluation of fatigue cracks in a box type radial crane runway girder with full penetration welds between the web and flange. After 20 years of service fatigue cracks were observed which were initiated in the flange at the toe of the full penetration weld with the web. It is expected that they curve underneath the weld but, due to the limited possibility for inspection, this could not be clearly determined. The observed cracks in the crane runway girder vary in length from a few mm to 330 mm with a summation of the lengths of all observed cracks being 750 mm, on a total length of 56 000 mm, thus being only 1.3 %.

The objective of the evaluation is to explain the observed cracks and to investigate whether the cracks could become critical and repair is required. The investigation consists of the fatigue analysis described in this paper and additional experimental investigations with scale tests of about 1:2 on equivalent I sections with a concentrated load and with a line load to investigate whether the cracks stop when they have grown through the residual tensile

* Corresponding author: j.wardenier@tudelft.nl

stress field. This paper describes the assessment of the fatigue cracks in the actual crane runway girder; publication of the experimental investigations will follow.

The results of this analysis supports the results of the recently completed German AIF/Fosta research program [1] by the University of Stuttgart on the fatigue behaviour of crane runway girders under a rolling wheel. All other literature studied, only provides qualitative evidence but not evidence really based on a rolling wheel.

2 Background of the crane runway girder The crane runway girder with a cross-section shown in Fig. 1, consists of a box section with a 75 mm rail plate located on top of the 75 mm flange and above one of the 65 mm webs. The web is extended and fully supported beneath the box section, thus no in-plane bending stresses occur in the web and the overall shear stresses are neglectable.

The full penetration welds between the web and the flanges are made with a foot print nearly equal to the thickness of the web. Details of the dimensions and details of the steel grades are, as far as available, given in Table 1.

The loads with the number of cycles per year are given in

Fig. 1. Crane girder detail with location cracks with estimated weld detail

© Ernst & Sohn Verlag für Architektur und technische Wissenschaften GmbH & Co. KG, Berlin · Steel Construction 10 (2017), No. 1

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J. Wardenier/P. de Vries/G. Timmerman · Evaluation of cracks in an offshore crane runway girder

Table 1. Dimensions, steel grades and qualities of the test specimens

1

half wheel diameter

Dimension mm

Steel grade/ quality

actual fy N/mm2

900

30CrNiMo8

1000

2

rail

400w75

OX AR 360S

950

3

flange

700w75*

OX 602

500

4

web

300w65, then 30

PCD36

350

5

stiffeners

300x40; c.o.c. 900 mm

PCD36

350

*) machined to 75 mm

Table 2. Estimated number of cycles per year in the crane runway girder Wheel load (kN)

N cycles per year (16 wheels)

7700

32

7000

8

5200

400

4200

800

3300

2400

1900

104 000

Total number of cycles/year

107.640

N cycles per year in %

3.4 %

EN 1993-6 Table 5.1

EN 13001-3-1, Table C4

IIW-XIII Class 431

leff (mm)

411

350*)

350

Hoist

Wheel load

EN 1993-6 Table 5.1

EN 13001-3-1, Table C4

IIW-XIII Class 431

kN

kN

0

1900

71

84

84

2500

3300

124

145

145

96.6 % 100%

3 Stress analysis As shown in Fig. 2 and more extensively discussed by Euler and Kuhlmann in [2], a rolling wheel on a crane runway girder produces for a particular location a normal stress range )Xz and an additional local shear stress range )Yxz. Just before the wheel, both stresses are increasing from zero until the maximum shear stress Yxz is reached, then the normal stress Xz is still increasing but the shear stress Yxz is decreasing until the normal stress Xz reaches its maximum when the local shear stress Yxz becomes zero.

Thereafter, the normal stress is decreasing but the shear stress is now with opposite sign increasing to a maximum after which the shear stress is decreasing with the decreasing normal stress. Thus for every wheel passing there is a shift in the direction of the principal stress. Therefore, as stated in [1], [2] the summation of the individual damages for normal stresses and shear stresses as used in EN 19931-9 [3] seems not to be appropriate for these non-proportional multi-axial loadings.

In this paper the analysis of the crane runway girder is based on the nominal stress range. These “nominal” stress ranges )Xz are determined in Table 3 with the effec-

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Table 3. Comparison “nominal stresses” with EN 1993-6, EN 13001-3-1 and IIW-XIII [4], [5], [6] Actual crane girder

Table 2. As shown, 96.6 % of the cycles are caused by the wheel load of 1900 kN. Since the first cracks were observed after about 20 years of service with regular inspection other causes of cracking then fatigue, like lamellar tearing and cold cracking, can be excluded.

68

Fig. 2. Normal stress Xz and local shear stress Yxz due to wheel load [1], [2]

)Xz (N/mm2)

5000

4200

157

185

185

10 000

5200

195

229

229

25 000

7000

262

308

308

20 000

7700

288

338

338

*) In this case leff is the same as in the IIW-XIII Recommendation [6] because the limit 0.1D f 50 mm is governing

tive lengths leff according to EN 1993-6 [4], EN 13001-3-1 [5], IIW-XIII [6], see Fig. 3.

It is shown that both EN 13001-3-1 [5] and IIW-XIII [6] give a lower effective length than EN 1993-6 [4], resulting in nearly 20 % higher “nominal” stresses Xz but the current fatigue class at N " 2.106 cycles for “nominal” stresses in [5] and [6] is also higher, although depending on the weld quality class.

EN 1993-6 [4] states that the local additional shear stresses Yxz at both sides of the wheel may be assumed to be 20 % of the vertical stress Xz due to the wheel load, see Figs. 3 and 4.

EN 1993-1-9 [3] states that apart from the damage caused by the individual normal and shear stress ranges, as stated before, the summation has to be checked. Thus for a rolling wheel the actual shear stress range )Yxz is equal to 2 w 0.2)Xz " 40% of the vertical stress range )Xz due to the wheel load.

J. Wardenier/P. de Vries/G. Timmerman · Evaluation of cracks in an offshore crane runway girder

Fig. 3. Effective loaded length leff [4]

However, as explained by Kuhlmann et al. [1] the vertical wheel load stress Xz and the local shear stresses Yxz are not independent from each other and also not proportional with as a result that the summation equation for the damage due to axial stresses and shear stresses in EN 1993-1-9 [3] is not applicable here. Thus, it may be better to relate the fatigue damage to the “nominal” stress ranges )Xz according to EN 1993-6 caused by the wheel which indirectly take account of the local shear stresses.

4 Fatigue analysis For comparison with the experimental results of Kuhlmann et al. [1], the nominal stress ranges )Xz are based on EN 1993-6 [4] and recorded in Table 4. The fatigue check according to EN 1993-1-9 [3] without load and partial factors is based on Eq. (1): 3

⎛ Δσ E ⎞ ni ⎜ Δσ ⎟ ≤ 1.0 which is equal to N ≤ 1.0 ⎝ i c⎠

Fig. 4. Additional local shear stresses Yxz [4]

– tref " 25 mm and t is the thickness of the flange (75 mm) where cracks are present; – )XC,75 is the corrected fatigue class taking account of the thickness effect Using the nominal stresses calculated according to EN 1993-6 [4] and the fatigue class 71 according to EN 19931-9 [3] with a thickness effect coefficient of 0.2 the damage D would be 2.64 (Table 4). As shown in Table 5, assuming in accordance with EN 1993-6 [4] for the shear stresses a Table 5. Fatigue damage based on shear stress only with class 80 [3]

(1)

within Eq. (1) and Table 4: – )XE is the equivalent stress range at 2.106 cycles which gives the same damage as the considered stress range )Xi " )Xz with ni cycles; – )Xc is the fatigue class (stress range) at 2.106 cycles; – Ni is the number of cycles at the stress range )Xi following from the )XC – N curve for the considered class; – D " 8 ni/Ni is the cumulative damage according to Palmgren-Miner

)Y " Fz,Ed

)Xz )YC

2*0.2*)X

)XC,75 mm

Ni

ni/Ni

7700 288

80

115

64

107 246

0.006

7000 262

80

105

64

172 721

0.001

5200 195

80

78

64

763 516

0.010

4200 157

80

63

64

2 221 203

0.007

3300 124

80

49

64

7 417 626

0.006

1900

80

28

64

71

117 237 516 0.018 0.05

Table 4. Fatigue damage based on nominal stress with class 71 [3] and thickness correction EN 1993-6 and EN 1993-1-9 based on nominal stress, class 71 for 25 mm leff " 411 Fz,Ed

)Xz

ni/pyr

yr

total

)XC

(N/mm2)

D (tref/t)0,2

)XC,75

(N/mm2)

Ni

ni/Ni

(N/mm2)

7700

288

32

20

640

71

0.803

57.0

15 464

0.04

7000

262

8

20

160

71

0.803

57.0

20 583

0.01

5200

195

400

20

8000

71

0.803

57.0

50 210

0.16

4200

157

800

20

16 000

71

0.803

57.0

95 291

0.17

3300

124

2400

20

48 000

71

0.803

57.0

196 452

0.24

1900

71

104 000

20

2 080 000

71

0.803

57.0

1 029 288

2.02 2.64

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J. Wardenier/P. de Vries/G. Timmerman · Evaluation of cracks in an offshore crane runway girder

Table 6. Fatigue damage based on nominal stress only with class “98” EN 1993-6 and EN 1993-1-9 based on nominal stress l eff "411 Fz,Ed

)X

ni/pyr

yr

total

)XC

(tref/t)0,2

)XC,75 mm

Ni

ni/NI

7700

288

32

20

640

98

0.803

78.7

40 666

0.02

7000

262

8

20

160

98

0.803

78.7

54 126

0.00

5200

195

400

20

8000

98

0.803

78.7

132 036

0.06

4200

157

800

20

16 000

98

0.803

78.7

250 584

0.06

3300

124

2400

20

48 000

98

0.803

78.7

516 606

0.09

1900

71

104 000

20

2 080 000

98

0.803

78.7

2 706 700

0.77 1.00

Table 7. Comparison fatigue aspects in the rolling wheel tests [1] and those in the considered crane runway girder Rolling wheel tests [1]

Crane runway girder considered

Wheels

Rolling wheel

Rolling wheel

Loading

Constant amplitude

Variable amplitudes but in groups (also some high loadings)

Weld size

Small not full penetration welds with 2a " 11 mm

Full penetration welds 2a " 65 mm

Crack size

Crack through the weld and initiations at the weld toes

Crack initiations at the weld toe at the flange

)Y " 2 w 0.2)Xz " 0.4)Xz with class 80 and m " 5, according to EN 1993-1-9 [3] and a thickness exponent 0.2, gives a neglectable damage of 0.05.

Table 6 shows that relating the behaviour only to the “nominal” stress range )Xz due to the wheel load, a damage of D " 1.0 would agree with a design class of 98 which is not far from the “class” 92 found by Kuhlmann et al, [1] for the tests with rolling wheels. However, here for the crane runway girder it is related to small cracks at the toe of the full penetration weld whereas in [1] it is related to weld failures from the root of not fully penetrated fillet welds in combination with crack initiations at the weld toe.

As summarized in Table 7, there are differences to be considered: – In the tests with the rolling wheels [1] the cracks started from the root defects. However, at crack through the weld, also crack initiations at the weld toe in the web were observed.

– The weld sizes in the rolling wheel tests were made in one layer whereas the welds in the crane runway girder were made in many layers, resulting in totally different residual stress patterns. Depending on the fabrication, with welds made in one layer the whole weld may have been in residual tension whereas the weld in the crane runway girder has mainly tensile residual stresses in the outer layers and as a result compression stresses in the inner layers. The last situation may cause that crack growth stops after the crack has grown through the residual tensile field.

– Another aspect concerns the high loads in the crane runway girder. These high loads may have delayed the crack growth or may have reduced the residual stresses in such a way that under loading part of the stress range

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in the outer zones of the welds is not anymore in tension but due to the reduction in residual tensile stress, in compression.

5 Ultimate limit state analysis To check the effect of a full crack between the web and the flange on the ultimate capacity, also an ultimate limit state FE analysis was carried out by PT Structural with contact elements between the web and the flange to simulate a fully cracked weld in horizontal direction. Further a 20 mm wheel eccentricity was included. The wheel loading was increased by increasing the vertical wheel displacement in steps until no solution could be found anymore due to exceeding of the distortion limits. At this stage the loading was more than twice the maximum wheel loading of 7700 kN. Based on these analyses, it was concluded that a full crack between the web and the flange but with the stiffeners still fully connected, gives a margin of more than 2 between the occurring maximum wheel load and the loading at which distortion becomes excessive.

6 Summary and conclusions This paper describes the analysis of a crane runway girder where after 20 years of service small fatigue cracks were observed. Using the nominal stress method according to EN 1993-6 [4] in combination with EN 1993-1-9 [3] shows that the damage due to local shear stresses )Y " 2 w 0.2)X " 0.4)Xz is 0.05, thus being small. Kuhlmann et al. [1] stated that since the vertical wheel load stress Xz and the local shear stresses Yxz are not independent from each other and also not proportional, the interaction equation in EN 1993-1-9 [3] is not applicable. Thus it can be con-

J. Wardenier/P. de Vries/G. Timmerman · Evaluation of cracks in an offshore crane runway girder

cluded that a classification based on the nominal wheel load stress range )Xz is preferred, but the current class of 71 in EN 1993-6 and EN 1993-1-9 requires upgrading. For this analyzed crane runway girder with full penetration welds and with small cracks at the weld toes in the flange, the damage would be 1.0 for a design class 98 which is not far from the “class” 92 found by Kuhlmann et al, [1] for the tests with rolling wheels, however, related to weld failures from the root of not fully penetrated fillet welds, which will be more conservative than the current case with small cracks.

In the (to be published) additional experimental investigations, fatigue tests were carried out on scale models with concentrated loads and with line loads which showed that for these multi-layered welds, crack growth stops after the cracks have grown through the residual tensile field.

References [1] Kuhlmann, U.; Herter, K.-H.; Euler, M.; Retteneier, P.; Weihe, S.: Versuchsbasierte Ermüdungsfestigkeit von Konstruktionsdetails mit Radlasteinleitung. Stahlbau 84 (2015) No. 9, pp. 655–666. DOI: 10.1002/stab.201510312 [2] Euler, M; Kuhlmann, U.: Fatigue evaluation of crane rail welds using local concepts. International Journal of Fatigue, 33, Elsevier; 2011: 111-1126.

[3] EN 1993-1-9. Design of steel structures. Fatigue. CEN, Brussels, Belgium; 2006.

[4] EN 1993-6. Design of steel structures. Crane supporting structures. CEN, Brussels, Belgium; 2007.

[5] EN-13001-3-1 Cranes – General design-Part 3-1. Limit states and proof of competence of steel structures. CEN, Brussels, Belgium; 2013.

[6] Hobbacher, A.: Recommendations for fatigue design of welded joints and components. Revision of IIW Doc. XIII 2151r4-07. IIW Doc. XV-1440-13, IIW, Paris, France, 2013.

Key Words: Crane runway girder; compression load; fatigue; fatigue cracks; crack growth; residual tensile stresses

Authors em.Prof. dr. Jaap Wardenier Faculty of Civil Engineering and Geosciences Delft University of Technology P.O.Box 5048 2600GA Delft, The Netherlands and Visiting Professor at Centre for Offshore Research & Engineering and Department of Civil & Environmental Engineering National University of Singapore #E1A-07-03, 1 Engineering Drive 2, Kent Ridge Singapore 117576 Ir. Peter de Vries P.A.deVries@tudelft.nl Faculty of Civil Engineering and Geosciences, Delft University of Technology P.O.Box 5048 2600GA Delft, The Netherlands Ir. Gerrit Timmerman GTIMMERMAN@PT-STRUCTURAL.COM PT Structural Design & Analysis B.V. Boelewerf 22 2987 VD Ridderkerk, The Netherlands

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Reports DOI: 10.1002/stco.201720010

Design and construction of the complex steel structure for the Amager Bakke waste-to-energy plant Johannes Hauptenbuchner

Amager Bakke (English: Amager slope) is the name of Copenhagen’s new waste-to-energy-plant located on the Amager peninsula. Once finished, it will be one of the largest incinerators in northern Europe and will be used for the combined production of district heat and electricity. On top of the waste-to-energy plant there will be a landscaped park featuring artificial ski slopes and a viewing platform. The support structure is mainly formed by a steelwork. The model-based design and construction of the complex, three-dimensional steel structure proved to be a challenging task for all the engineers and companies involved.

1 Building and structure The inner space of the energy plant is ruled by the technical requirements. The external shape of the building is representing the strong idea to create a slope with its continuous way down from top. By integrating complex industrial technology into a compact building structure such as a mountain segment, Copenhagen-based architects BIG and structural engineers MOE A/S have created a structure with very high construction requirements. The building parts are arranged in line with the technological sequences of the power plant, from east to west, into the fuel delivery and storage areas and into a process building for the incineration, exhaust treatment and power generation.

The western end, at the highest part of the building, is formed by the chimney rising more than 60 m. A storey for administrative functions has been “pushed in” underneath the chimney (Figs. 1 and 2). As the support structure for this compact building, MOE A/S chose a steel frame structure, which sits up from a height of 17 to 30 m on a rein-

Fig. 1. AmagerBakke, aerial view © ARC-BIG

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forced concrete structure. For technological reasons, the waste bunker is surrounded by reinforced concrete walls up to 40 m high. The foundations to the delivery building located in the east form the supports for the steel columns at a height of 8 m. The steel structure is divided into different types of construction: a single-storey-shed construction in the delivery and waste bunker area, a tree-like structure in the process building and a multi-storey structure with prestressed concrete floors in the administrative area. It is worth highlighting the process area, where the structure largely spans over the industrial components with few possibilities for support and stabilization in the inner space.

The roof of the process building follows the three-dimensional and irregular shape of the skiing park. In the area of the waste bunker, the 11 m high, triangular roof trusses carry the heavy concrete roof 40 m above the waste bunker pit. The multi-storey administrative building is shielded from the process area by concrete elements to comply with fire protection requirements. The supporting shell of the roof is made of prestressed concrete elements on which the ski slope surfaces are modelled.

2 Contract of Züblin Stahlbau In March 2014 Züblin Stahlbau GmbH successfully passed the prequalification procedure for construction of the building’s structural steelwork and the precast concrete element construction and was awarded the contract for

Fig. 2. AmagerBakke, section © ARC-BIG

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Fig. 3. 3D-TEKLA model – contract scopes of Züblin Stahlbau (by Züblin Stahlbau)

these works in June 2014. The contractual work and services of Züblin Stahlbau GmbH included the fabrication planning with pro rata design of the structural joints and the fabrication and erection of 6500 t of building steelwork, 23 000 m2 of prestressed concrete hollow-core elements with in situ concrete work, 2000 m2 of reinforced concrete wall elements, 12 000 m2 of sandwich walls with fire protection classification, the drainage for the roof structure and the supply and erection of the chimney (Fig. 3). Apart from preparing the fabrication planning, Züblin Stahlbau had also taken on responsibility for the structural design of some parts of the work (“design and build”). That work included the prestressed concrete floors, the approx. 10 to 80 m high stair and lift towers, made of steel and sandwich construction, and all secondary steel structures and the roof drainage system. The complex engineering tasks were coordinated and executed by Züblin Stahlbau, but assistance was also provided by external engineering consultants. The sophisticated welds were produced in the company’s own workshops in Hosena.

The reinforced concrete elements and the chimney, including the production planning, were implemented by the Group’s internal departments of Ed. Züblin AG Direktion Ingenieurbau Nord und Züblin Chimney & Refractory GmbH. The steel structure was erected on the construction site in Copenhagen by our partner firm IMO Leipzig GmbH.

3 Design and interface clarification in 3D model The 3D program TEKLA structures was used for the geometric clarification and coordination of the interfaces and the fabrication planning of the main structure (Fig. 3).

Züblin Stahlbau has been using this program, in addition to other 3D software, for the fabrication planning of steel structures for about eight years. It is becoming increasingly clear that the exchange of model data with all participating engineers allows effective clarification of interfaces between construction stages and construction trades. In the project described here, a basic model from the client was available from the beginning of the fabrication planning.

The model had been created by consulting engineers MOE A/S, who had been appointed by the client to produce the design and structural calculations for the structure. As

MOE also uses the TEKLA program to edit building models, Züblin Stahlbau was able to import the data directly into the TEKLA program and working with interfaces for 3D data was unnecessary. Although the capabilities of interfaces for complex 3D building structures are being continuously developed, attributes of the objects are often lost during transfer of the data or time-consuming reworking is required. The direct transfer of model data into the TEKLA program provided decisive advantages and ensured the very fast and high-quality transfer of the geometric properties of all structural members and elements. In the basic model, the designers at Züblin Stahlbau were able to use the geometric data of the in situ and precast concrete elements, the steel structure of the main building and the chimney and the design data of interior work trades, e.g. drainage pipes. The positions of the members were defined in the axis system of the structure in the preliminary design phase and formed the basis for the structural calculations.

The stage-by-stage detailing of the structural steelwork and precast concrete elements as the basis for the shop drawings was carried out in line with the overall timetable determined by the client. The production planning for the fire-resistant façades, drainage works and the interior work trades in the office building followed with the necessary time lag. The technical coordination work was particularly effective thanks to the continuous feedback of 3D model data to the client’s consultant engineers. At the same time, Züblin Stahlbau provided the model data for the subcontractors involved. The status of the production planning based on this data was continuously fed into the 3D model in Züblin Stahlbau’s engineering office. Owing to the diversity of the 3D software used, the standardized Industry Foundation Classes (IFC) interface was one method used to import and export data. The approval of the production planning by the client’s design engineers was based on a model-based check of the design details and conflict checking in the 3D model incorporating all technical interior work trades, e.g. the plant and pipe construction.

The steel structure of the shed-type building is divided into columns, cantilevered and suspended beams, irregular heavy trusses and a tree-like support structure for stabilizing the roof trusses in the process building. The columns are welded box sections, which are stiffened by longitudinal ribs and bulkheads. The box sections are up to 1600 mm deep and 800 mm wide. The chords and diagonal members of the trusses are mostly welded tubular or box sections with depths, widths and diameters of 500 mm.

The coupling members in the walls and the tree-like structures of the process building are formed by tubular or box sections with a diameter of 200 to 800 mm (Fig. 4). Each of these structural members must be checked and at times also corrected at the beginning of the detailing with regard to their precise geometric position and dimensioning, according to the specifications from the main structural analysis system and the project drawings. At the same time, the general construction details from the main structural calculations of MOE, e.g. the anchoring of the columns or the truss joints, had to be adjusted or changed to suit the specific situations and according to the criteria of technically, economically and qualitatively optimum execution. The main changes to be agreed with the client’s design engineer included, for example, the execution of the fully welded

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Fig. 4. Tree-like roof structure for process building (by Züblin Stahlbau)

node joints of the verticals and diagonals with the chord sections of the trusses. To exploit the advantages of the modern fabrication facilities in Hosena optimally and to organize the erection sequences on schedule, the Züblin Stahlbau engineers proposed that all on-site joints should be bolted.

The research carried out in advance of the project had determined that transport by road was the optimum solution. In this case the dimensions for economical logistics prescribe maximum widths and lengths of 3.50 and 22.0 m respectively. The existing dimensions of the assemblies then had to be optimized accordingly regarding the layout of bolted joints. This primarily concerned the trusses in the process area with widths of 5 to 7 m, due to the shape of the roof, lengths of 30 m to the middle support and the roof trusses of the waste bunker, which are 11 m deep at the western support and 40 m long in total. The trusses designated in the project as welded assemblies could generally be converted into bolted individual units. However, very massive node joints resulted. The roof area of the building is accessible to the public. As a result, additional failure scenarios are taken into account for the support structures below it, e.g. failure of a complete truss. These lead to very high design loads, above all, to compressive loads on the members. To prevent stability failure of the gusset plates, additional stiffening was designed and verified geometrically. The transfer of the model data to MOE for checking of the geometric data, e.g. checking gusset plates for conflicts with pipes, facilitated the approval procedures. During the preparation of the structural joint calculations, Züblin Stahlbau also carried out geometric iteration by exchanging 3D information between structural engineers and designers, both in-house and with the external consulting engineers. Using the example of the waste bunker trusses (Fig. 5), however, the engineers reached the limits of feasible bolted joint design during the structural analyses of the chords. In the exceptional case of the failure load case of a truss, additional forces acted on the chords, which were already highly stressed within the normal load range. The tension loads on the bottom chords are up to 17 000 kN. The truss sections are therefore made of the higher material grade S460.

Directly above the chords there is an access level made of prestressed concrete elements with a concrete topping, and below them there is an overhead crane. In addition to the economic aspects regarding the production of

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Fig. 5. Preassembled truss for waste bunker (by Züblin Stahlbau)

Fig. 6. 3D section through roof structure for process building (by Züblin Stahlbau)

a bolted joint that can absorb these high tensile forces, the space available also limited the possible execution options.

The chord joints of the bunker trusses were therefore welded on site. To this end, a weld backing and an auxiliary construction were added in the workshop to fix the joints. The diagonals and verticals were designed with bolted joints and were bolted before the welding to the chords. Execution class EXC 4 was specified for the waste bunker structure due to the load, type of construction and use.

The coordination of the steel structure for the roof trusses with the geometry of the roof support shell made of prestressed concrete elements is particularly worth highlighting due to the especially time-saving and precise clarification of the interfaces. The roof support shell follows the irregular 3D shape of the skiing park and is supported on the top chords of the roof trusses. The gap between the precast elements will be closed with reinforced in situ concrete after the elements have been laid. As the prestressed concrete elements have a constant depth per element due to the production technique and the adjacent roof girders follow different pitches, gaps of up to 100 mm result along the edge of the chords. The elastic support strips bridge a tolerance of up to 10 mm (Fig. 6).

After modelling the steel trusses, the TEKLA model was exported in subsections and the concrete elements laid out by the precast concrete element planner according to the profile of the roof surface. The positions of the ele-

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Fig. 7. Upper chords of roof trusses for process building (by Züblin Stahlbau)

Fig. 8. Column head in tree-like structure (by Züblin Stahlbau)

ments had to be optimized so that an additional support element is required on one side only. After feeding this data back into the master model, the additional steel components were modelled to close the joints in order to carry the load. As a result, the top chords acquired an irregular toothed strip that could be dimensioned individually for each location along the roof truss. This procedure substantially shortened the time-consuming iteration between structural steelwork and concrete designers – a requirement for completion of the structural steelwork modelling and the shop drawings (Fig. 7).

With regard to the extensive interface clarifications in the 3D model, the coordination of the main structure with the interior work trades in the administrative area of the power plant will also be briefly discussed here. The administration building is a multi-storey building with a structural steel frame stiffened by the floors. The floors are made of composite steel girders on which precast prestressed concrete elements are laid. The voids between the precast elements, the edges and the composite girders, socalled delta beams, are filled with reinforced concrete on site. An in situ concrete floor is laid on the precast elements, which forms a support for the subsequent fixing of the structures for the lift shafts, the stairs and the glass balustrade. The innumerable geometric interfaces between the main steel structure, the composite girders, the precast pretensioned concrete elements and the interior work trades were coordinated in the 3D model by synchronizing the submodels exchanged with the engineering consultants of the respective suppliers.

order to enable continuous assignment to the final item numbers in the parts lists. Preparation of the shop drawings and generation of the NC data was semi-automated by means of software modules following completion of all modelling and positioning. Regardless of the complexity of the data, transfer of the NC data – in order to control the cutting to size of the component parts on the sawing/drilling and flame-cutting machines – can be fully automated.

However, assembly of the steelwork members in the workshop is based on the two-dimensional shop drawings. These must be manually reworked for individual structures. For this project, clear representation of the welds, very exacting in terms of workmanship, involved very considerable work.

The lead times – from provision of the shop drawings to the start of cutting the steel plates in the workshop – were a maximum of two weeks. As the preparation of the shop drawings could not always be completed on time, due to the high degree of complexity involved, accelerated handover of 3D model data and materials lists to the fabrication department was established to enable the job scheduling in the workshop to start exactly on time.

The NC data-controlled cutting machines enable precise and mostly error-free transfer of even complex geometries to the component parts, which are assembled in the workshop. The assembly of individual members with low repetition character, however, is subject to the classic laws of skilled manual work. Accordingly, the complex and extensively stiffened cross-sections and node joints required highly skilled manual input during assembly and welding.

This resulted in a great deal of work in the preparation and execution of the checks and tests regarding geometry and welding quality. When drawing up the extensive inspection and test plans, in addition to classic geometric checks, an additional control step using a 3D scanner was introduced in order to realize precise fitting of complex 3D substructures (Fig. 8).

The 3D scanner was set up with the coordinates of the 3D TEKLA model and referenced at pinchpoints along the main axes on the physical component. Following the assembly of a selected node joint, the measured data was recorded by scanning the edges of the steel plates of the planes and the drilled holes of all attached parts (Figs. 9 and 10). The set of points acquired in this way was compared separately by the software of the scanner for the geometric planes and the drilled holes with the design coordinates from the 3D TEKLA model.

4 Shop fabrication on the basis of the 3D model Depending on the degree of detailing, all steel member data for materials ordering, job scheduling and fabrication can be generated from the 3D TEKLA model. Owing to the long lead times for ordering the sheet metal segments, especially in grade S460, rough materials lists were generated from the model at a very early stage and orders placed with the suppliers. Project-based rolling of the required dimensions and quantities was therefore feasible. However, this approach requires very early design reliability with regard to the main structural dimensions and preliminary item designations of the individual members. The preliminary item numbers were included as attributes of the component parts in the model through to the detailed modelling in

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Fig. 9. 3D survey of column heads in fabrication shop (by Züblin Stahlbau)

Fig. 10. Report of 3D survey of column head planes © Hemminger

If deviations occurred outside the tolerance of # 1 mm, the measured coordinates were imported into the model to enable simulation of the assembled item. The version of the assembly thus created was used to check the effect of the deviations on the adjacent components in the model.

Where it was still possible with regard to the timing of the production sequence, corrections were incorporated in the model and the shop drawings. It was found that, on a caseby-case basis, deviations of 3 to 5 mm were tolerable, as the adjacent nodes of the spatially inclined, connected diagonal girders were mostly 15 m away and the deviations could be compensated for within the given area through the hole clearance or additional filler plates. The change in the assembled items on the basis of the measured data was discarded due to the experience acquired. There was no reduced risk of renewed deviations for a second assembly, as the measured data could not be used effectively for the correction due to the complex spatial positions of the component parts. The components were not finally welded together until after the measured data had been evaluated positively. In addition to checking the geometry of spatial nodes using a 3D scanner, the assembly of the trusses for the roof structure required conventional control through temporary preassembly during fabrication. To this end, all

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Fig. 11. Preassembly of the 40 m long truss chord for waste bunker in fabrication shop (by Züblin Stahlbau)

the bracing members of the trusses were produced completely; however, the chord beams of the truss were not finally welded until after preliminary assembly and successful measurement. In this way, geometric deviations due to weld shrinkage, unfavourable superimposition of tolerances or other errors could be analysed in good time and corrected immediately (Fig. 11).

An important control step in the quality monitoring of the fabrication sequence is the testing of the welds. For the parts of the building classified as execution class EXC 4 in particular, substantial test and inspection work arose, the organization of which had to be fitted in during the fabrication sequence. The scope of non-destructive weld tests was based on European standard EN 1090-2 in accordance with the project specifications. Consequently, the areas with classification EXC 4 and the areas with mostly full utilization of the capacities of the cross-sections (capacity utilization U # 50 %) required complete testing (100 %) of all penetration welds (butt joints). The tests were organized and verified on the basis of weld lists extracted from the 3D model. Clear assignment of the weld to the welder and the test results could thus be shown clearly and tracked. With the help of this data, it was possible to verify the statistical size of tests of other weld types, or rather, lower test requirements in other areas. Additional marking of the welds according to the standardized sequences with hard stamping also enabled clear assignment to the welder.

5 Erection planning based on 3D simulation To prepare for the erection work, the erection steps were mapped in the 3D TEKLA model. This offered the best conditions for interlinking the examination of the interim construction states with the hoist load planning and controlling the flow of materials to the construction site. For the processing of the work planning in the 3D TEKLA model, the building was divided into subsystems that were based on the system of the building structures and the time schedule of the client. Upon incorporating the time schedule, time-slots of approx. 2 to 4 weeks were collated as separate subsystem groups. Completion of the planned processing chain of a subsystem group in the engineering office, in the materials order and in the fabrication thus corresponded to the expected erection progress on the

Reports construction site. The detailed planning of deliveries was based on separate attributes that were assigned to the components with a lead time of approx. 4 weeks before delivery and the subsequent erection. These “call orders” were transferred to the transport units during the dispatch planning; the transport units were then taken on trucks by road to the construction site.

The final modelling of the aforementioned subsystem groups was followed by detailed planning of the individual erection steps depending on requirements. To do this, the subsystems were handed over to the erection planners of the erection contractor in TEKLA format for further processing. The display of individual steps by selective marking of the components enabled, in the further processing, the determination of weights and centres of gravity of assemblies, geometric checking of hoisting scenarios and the handover of information and model data, e.g. to external engineering consultants for the structural analysis of intermediate construction states.

In the case of intervention in the overall structural system, the information handed over was also used by the client‘s design engineers, MOE, to perform structural analyses. An erection step basically described the planned erection performance for a day, as the aim was to attain a structurally secure interim state at the end of each day.

This procedure resulted due to the wind conditions on the construction site, which is located directly on the Baltic Sea coast. The coastal wind blows irregularly and with high speeds, so large components are exposed to very high loads. In addition, rapidly changing wind conditions and gusts leading often to a failure of the large, wind-sensitive cranes. The large cranes used with crawler tracks and a safe working load of 130 to 750 t can only work up to certain wind speeds for safety reasons. At the height of the crane booms used (approx. 130 m), the wind speeds must not exceed 9 m/s (or 13 m/s for smaller cranes).

Thus, securing intermediate construction statuses by relieving the loads needing cranes is only possible for a short time. The assistance by crane was therefore not planned for use overnight. Hoisting work began according to the expected wind time-slot with adequately low speeds based on the weather report.

Using the example of the heavy truss construction for the waste bunker, the exacting erection preparations become particularly clear. The roof structure of the waste bunker consists of 40 m long triangular trusses positioned longitudinally with the roof slope, whose top chords are stiffened in the plane of the roof by transverse diagonals.

Vertical transverse trusses, on which the roof structure is supported, are located on the support axis of the highest point. The stiffening of the bottom chords and the bottom support for the roof structure is provided by a concrete slab made of prestressed concrete elements and in situ concrete in the bottom chord plane of the trusses. This structure does not have sufficient stability until the final step after the in situ concrete slab has cured. The erection sequence was therefore divided into 14 operations, which required more or less the entire structure and the necessary auxiliary construction works to be incorporated in the structural analyses. The substructures were defined in the 3D TEKLA model and processed further using the structural design software for frames accordingly.

Fig. 12. Intermediate construction step of the roof trusses for the waste bunker (by Züblin Stahlbau)

The first erection step was defined as the lowering of a single truss onto the two supports. The structural elements of the roof truss, including the temporary anti-tilt bracket, had to remain stable with a self-weight of 900 kN and the horizontal wind load of 300 kN. Based on this, the hoist load studies were carried out, the erection aids dimensioned and the erection description drawn up for approval of the erection works and for instructing the erection personnel. The other structural systems to be checked ranged from the partially completed sections of several trusses through to analysis of the structure during placement of the in situ concrete (Fig. 12). Similarly, an erection study for the entire roof structure of the process building was carried out on the basis of the daily erection progress.

Individual systems were derived from this for the structural analyses, which were coordinated and carried out by Züblin Stahlbau. The work on the 3D TEKLA model, which also took place online during the regular video conferences, meant that the client’s design engineers could be quickly and effectively integrated and involved in the determination of the internal forces and moments for larger building sections.

6 Structural steel erection The erection works on the construction site started on schedule in January 2015. To this end, large crawler cranes and safe working loads of 600 t, 300 t and 130 t were mobilized in order to be able to achieve the hoist weights and to reach all the locations in the structure.

Mobile cranes were used to enable permanent deployment on the construction site. Notwithstanding the extensive preparations and planning, the erection of the steel structure was a particular challenge. The implementation of the erection concept described for hoisting the trusses above the waste bunker from April to May 2015 was an outstanding event. The five 40 m long roof trusses weighing 90 t were preassembled upright and were lowered onto their supports by two crawler cranes with capacities of 300 and 600 t and an outreach of up to 40 m (Fig. 13).

Erecting the trusses following horizontal preassembly would have required extensive auxiliary construction works in order to prevent lateral buckling of the truss members during putting up the truss.

In addition to these heavy lifts, particular hoisting work repeatedly had to be carried out during the erection

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Fig. 16. Lifting unit 2 of the chimney (by Züblin Stahlbau)

Fig. 13. Lifting the first truss for the waste bunker (by Züblin Stahlbau)

Fig. 17. View of south side of Amager Bakke in April 2016 (by Züblin Stahlbau) Fig. 14. Lifting a preassembled unit for the process building (by Züblin Stahlbau)

Fig. 15. AmagerBakke, view from south-east towards waste bunker © Christoffer

of the trusses for the process building, which span over the boilers and filter units. Not only were the dimensions of the preassembled units – up to 40 w 15 w 10 m – a challenge for the steelwork erectors (Fig. 14), but also the access for erection crews at heights of up to 90 m, above the already installed power plant technology, also had to be planned in precise detail for each hoist (Fig. 15). A high-

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light in the erection work on the main structure was reached with the lifting into place of the 300 t, 60 m high chimney. The chimney was lifted in six parts with maximum weights of 80 t to a level of 130 m and was anchored to the building at three support points (Fig. 16). Before detaching the chimney parts from the crane, it was necessary to align the 72 drilled holes of the connection plates above each other at each support point, to partly drill in the building-side plates and to connect them with highstrength bolts with diameters of M36 to M48.

The wind was not only a serious influencing factor with regard to applied loads during the interim construction states. The time lost due to excessive wind speeds was also an organizational challenge during the erection work. During the winter months, as well as in the summer of 2015, downtimes amounting to as much as 50 % of the working hours occurred because of strong winds. In order to counteract the delays that resulted, an additional crawler crane with a safe working load of 750 t was also deployed at times. Using this measure it was possible for work on the process building to take place on two sides at the same time. However, the time schedule delays could not be fully made up. Nevertheless, the waste incineration plant went into comissioning on time in June 2016 (Figs. 17 and 18). Completion of the whole building structure will be concluded with the erection of the ramp structure, which leads skiers from the roof level at 30 m

Reports to the starting point. This is scheduled for mid-2017. Following that, the power plant operator plans to carry out the construction work for the ski park above the roof covering.

The numbers at a glance Steel structure: Number of steelwork components: Structural bolts up to M48: bolting assemblies Sandwich elements: Prestressed concrete hollow-core slabs: Precast reinforced concrete elements: Start of steelwork erection: Completion of main building:

approx. 6500 t approx. 43 000 approx. 71 000 high-tension approx. 14 000 m2 approx. 20 000 m2 approx. 2000 m2 beginning of 2015

Fig. 18. Amager Bakke in September 2016 (by Züblin Stahlbau)

September 2016

Construction team Client:

ARC Amager Resource Center Architects: BIG Bjarke Ingels Group Copenhagen Structural engineers: MOE A/S Consulting & Engineers Design, fabrication, supply, Züblin Stahlbau GmbH, erection: Hosena, Germany

Author Dipl.-Ing. Johannes Hauptenbuchner Züblin Stahlbau GmbH Bahnhofstr. 13 01996 Hosena Germany info@zueblin-stahlbau.de

Announcements International Conference on Architecture and Civil Engineering 2017

International Conference on Building Envelope Systems and Technologies

39th IABSE Symposium Engineering the Future

Location and Date: Singapore, Singapore, 10–12 March 2017

Location and date: Istanbul, Turkey, 15–18 May 2017

Location and date: Vancouver, Canada, 21–23 September 2017

Information and registration: http://icace.coreconferences.com/index. html

Information and registration: http://icbestistanbul.com/

Information and registration: www.iabse.org

International Conference on Wind Energy Harvesting 2017 (WINERCOST’17) Location and Date: Coimbra, Portugal, 20 and 21 April 2017

Eurosteel 2017 Location and date: Copenhagen, Denmark, 13 15 September 2017 Information and registration: www.steelconstruct.com

Information and registration: www.cmm.pt/WINERCOST17

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People

People Professor Gert Albrecht dies at 75

Colleagues and former students have lost a friend and mentor. Univ.-Prof. Dr.Ing. Gert Albrecht has died at the age of 75 on 16. January 2017. Our thoughts are with his wife Brigitte and his two daughters, Tanja and Christina, who have lost a wonderful husband and loving father.

Univ.-Prof. Dr.-Ing. Gert Albrecht was born in Chemnitz/Sachsen as son of builder Kurt Albrecht and lived from 1951 on in Berlin. After finishing high school in Berlin, he studied civil engineering at TU Berlin and graduated in 1968. During these years he came for the first time in touch with steel structures working as student assistant. The same year, Gert Albrecht was hired as engineer

by Friedr. Krupp Maschinen- und Stahlbau Rheinhausen. In the firm’s department of bridge construction, he dedicated his time to structural analysis and investigation of bridges, including moving and modular bridges, and challenging research themes, such as the design of long-span suspension bridges. During this time his interest and ambition for teaching became obvious when he became part-time teacher for mathematics at the Städtisch-Naturwissenschaftliches Gymnasium Rheinhausen.

In 1970, Gert Albrecht became doctoral research assistant at the Chair for Steel Structures led by Prof. Roik at TU Berlin. Albrecht followed Roik when he became Professor at the Ruhr-University Bochum in 1973. His studies on the effective width of T-beam sections under consideration of elastic-plastic behavior became his dissertation topic which re-

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ceived the summa cum laude honor. His time collaborating with Prof. Roik gained him experience and reputation in the field of forensic analysis and bridge engineering. Notable examples are the rehabilitation of the West-Gate-Bridge Melbourne, the forensic analysis and rehabilitation of the Tasman-Bridge in Hobart, Tasmania after partial collapse and structural analysis of the 370m-span- cable-stayed bridge Rheinbrücke Düsseldorf-Flehe.

Prof. Albrecht was engaged in committee work and standardization throughout his career. He contributed to the new DIN 18800 and the new DIN 1072, and was member of the working group “Effective Width” which was setup by the Secretary of State for Transportation.

Further, he was member of the group of experts on “Bonding Technology” at Deutsches Institut für Bautechnik between 1983 and 1990.

Prof. Albrecht became Partner of the consulting engineering firm Haensel-Roik-Albrecht (HRA) and received the proof expert for statics license for metal structures in 1990. At HRA, he contributed to significant structures, such as the 310m-span cable-stayed bridge Rheinbrücke Beeckerwerth, the widening of the Rheinbrücke Köln-Rodenkirchen, the railway brides crossing the Süderelbe in Hamburg, and the Sauertalbrücke Trier-Luxemburg.

On October 1, 1992 Gert Albrecht was appointed Professor at the Chair of Steel Structures, Department for Civil, Geo and Environmental Engineering at Technical University Munich (TUM). The same year, he founded an engineering consulting firm in Munich which became Engineering Consultants Albrecht-Doblies in 1996. As it was in Bochum, Professor Albrecht combined academic interests in science and education with engineering practice and consulting. The outcome of this quite time-consuming approach benefited in particular his students. Professor Albrecht knew on how to explaining rather dry theoretical contents using illustrative examples out of recent and exciting engineering practice which made his lectures enjoyable and fascinating. Through close collaboration with industry Professor Albrecht succeeded in establishing cutting-edge research projects at his Chair of Steel Structures and involved undergraduate and graduate students. Creating opportunities for his students, Prof. Albrecht offered many promising students a full-time position in his engineering firm. Notable engineering projects during that time are the design of the Strom- und Vorlandbrücke des Wasserstraßenkreuz Magdeburg, the proof engineering of the Lehrter Bahnhof Berlin, the research and engi-

neering consulting for the maglev TRANSRAPID as expert and proof engineer for the German Federal Railway Organization (EBA). Prof. Albrecht’s research interest and passion was in bridge engineering, among them steel-reinforced concrete single and double hybrid bridge design and performance assessment, orthotropic deck, long-term behavior, such as creep, shrinkage and long-term tension stiffening, structural glass and fatigue/life-cycle assessment and rehabilitation of bridges.

In 1996, Prof. Albrecht became member of the examining board for the accreditation of proof engineers in structural engineering in Bavaria. He was voted into the experts committee “Structural Engineering” of the IHK for Munich and Upper Bavaria and into the experts committee for “Structural Glass” of the Deutsches Institut für Bautechnik.

He became Chairman of the working group for EC 3-2 in 1997.

Professor Albrecht has been a dedicated mentor for Bachelor, Master, Doctoral and Visiting Students providing guidance, fostering talent, detecting potential in his students and creating enormous opportunities. For all of his students, his leadership laid the foundation for a strong career in engineering practice or in academia.

Professor Albrecht inspired international collaboration and knowledge transfer. He was Chairman of the biennial Japanese-German-Bridge Symposium taking place in Japan and in Germany with experts from academia, industry and government attending. Prof.

Albrecht had research collaborations and student exchange with Osaka Institute of Technology and the Technical University Timisoara.

After becoming Professor Emeritus TUM, Professor Albrecht stayed active attending conferences, chairing the JGBS. Too early and for many suddenly Professor Albrecht is pulled out of this life. He will be missed.

Friends and colleagues of Gert Albrecht view back to the shared times full of gratefulness. The authors of this obituary will always treasure the memories to a friend and mentor. The thoughts which move us are expressed in the words by the German poet Matthias Claudius: ”They buried a good man, but to me he meant much more.”

Marcus Rutner, Hoboken, USA Akimitsu Kurita, Osaka, Japan

ECCS news

Events

IFiress 2017

BESTInfra 2017

6–7 June 2017, Naples, Italy

21–22 September 2017, Prague, Czech Republic

The aim of the symposium is to collect and disseminate the latest results of scientific research concerning fire safety. It represents an opportunity to share research, technology and expertise among peers in an international forum. The workshop aims to address the international scientific community, but also the leading industrial and professional representatives in order to inspire the debate on critical issues concerning fire safety.

The event, organized by the Department of Structures for Engineering and Architecture and the International Council for Research and Innovation in Building and Construction, is the second time the International fire Safety Symposium has been held. The first event took place in Coimbra, Portugal, in 2015. The symposium will be held at the Federico II Convention Center.

The intention of the BESTInfra 2017 international conference is to enable an exchange of the latest ideas regarding transport infrastructure design, construction, operation and maintenance between the industrial, public and academic sectors, between various disciplines and between countries. The contributions will be divided into sessions focusing on the following main topics: – Low-energy and high-performance materials – Roads, bridges and tunnels with increased durability and an extended service life – Advanced technologies and smart solutions for railways – Systems for management, durability assessment and life cycle cost analysis in transport infrastructure – Environmental protection and green transport infrastructures – Safety and security of structures and structural diagnostics

WINERCOST’17 20–21 April 2017, Coimbra, Portugal

“WINERCOST’17 – THE INTERNATIONAL CONFERENCE ON WIND ENERGY HARVESTING 2017” will be held at the University of Coimbra, Portugal, on 20–21 April 2017.

WINERCOST’17 is dedicated to “Wind Energy Harvesting: From Aeolian Farms to Cities of the Future”. It will provide a forum for presenting and discussing different aspects of wind energy and wind energy technologies in urban and suburban built environments in order to enhance the concept of “smart future cities”.

The conference is integrated into the Transport and Urban Development (TUD) COST Action with the same designation. This COST Action aims to merge the efforts of the European research groups working on wind energy technology and the pathways to introduce it by means of robust applications for urban and suburban built environments, thus enhancing the concept of “smart future cities”. This action revisits safe, cost-effective and socially accepted wind energy technology for consideration in the design and development of future urban/suburban habitats. In addition, the most important issue in the social acceptance strategy is scrutinized in close collaboration with municipal authorities, industry, manufacturers and international wind energy organizations and platforms.

WINERCOST’17 will provide a forum for presenting and discussing different aspects of wind energy harvesting technologies and will address topics such as wind characteristics and loads, structures, materials and dynamics, grid integration, operation and control, markets, strategies, policies and socio-economics, smart cities and environmental aspects. www.cmm.pt/WINERCOST17

www.ifiress2017.unina.it

Eurosteel 2017 13–15 September 2017, Copenhagen, Denmark

This international conference, organized by the Technical University of Denmark (DTU) and the Danish Steel Institute (DSI), is the 8th European Conference on Steel and Composite Structures and will be held in Copenhagen on 13–15 September 2017. The previous Eurosteel conferences were held in Athens (Greece, 1995), Prague (Czech Republic, 1999), Coimbra (Portugal, 2002), Maastricht (The Netherlands, 2005), Graz (Austria, 2008), Budapest (Hungary, 2011) and Naples (Italy, 2014).

The Eurosteel Conference will be held in conjunction with ECCS Annual Meetings and 2017 European Steel Design Awards on 14 September 2017. More details on www.steelconstruct.com.

The conference is being organized by the Centre for Effective and Sustainable Transport Infrastructure (CESTI) and the Faculty of Civil Engineering, Czech Technical University, Prague.

3rd International Symposium on Connections between Steel and Concrete 27–29 September 2017, Shanghai, China

The 3rd International Symposium “Connections between Steel and Concrete” is an attempt to continue the tradition of the previous two successful conferences organized by the Institute of Construction Materials, University of Stuttgart.

The first symposium was held in 2001 and the second in 2007, with approx.

400 delegates participating from across the globe.

The aim of the symposium is to bring the expertise and opinions of educational institutes, researchers and industry together and provide a quality platform for discussing and harmonizing ad-

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ECCS news vancements in the field of connections between steel and concrete. www.consc2017.uni-stuttgart.de

Danish Steel Day

TWG 7.5 – Practical Improvement of Design Procedures

Danish Steel Day will be held on the 9 November 2017.

Chairwoman: Prof. Bettina Brune

XI Conference on Steel and Composite Construction

Dutch Steel Day

23–24 November 2017, Coimbra, Portugal

Dutch Steel Day will be held on the 10 October 2017.

Finnish Steel Day Finnish Steel Day will be held on the 21 November 2017.

Norwegian Steel Day Norwegian Steel Day will be held on 2 November 2017 in Hotel Bristol, Oslo.

Portugal Steel Day Following the success of the previous conferences organized by CMM, the XI Conference on Steel and Composite Construction intends to promote the most recent innovations and achievements in this type of construction and make a decisive contribution to the promotion, consolidation and development of the sector. The conference will also be a privileged encounter for the exchange of ideas and experiences among the different players in the design and construction of steel and composite structures as well as in the teaching and research activities in this sector. In the 2017 event, emphasis will be given to the theme of refurbishment of the built environment, which is considered as a priority in and strategic for Portugal. All intervention in this area requires the establishment of multidisciplinary teams involving different specialists whose interaction will be stimulated in this conference.

The 2017 event will give special relevance to the theme “Next Generation of Eurocodes”. www.cmm.pt/congresso11

Announcements Czech Steel Day Czech Steel Day is being organized by the Czech Constructional Steelwork Association and will be held on 2 November 2017. www.caok.cz

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XI Conference on Steel and composite constructio day will be held on 23 and 24 November 2017 in Coimbra, Portugal.

Sweden Steel Day Sweden Steel Day will be held on 26 October 2017.

Technical Committees (TC) activities PMB – Promotional Management Board Chairman: Mr. Yener Gur’es Date: 11 April 2017, Brussels, Belgium

AC3 – Bridge Committee Chairman: Dr. Oliver Hechler Vice-chairman: Mr. Pavel Ryjacek

TMB – Technical Management Board Chairman: Prof. Milan Veljkovic Date: 11 April 2017, Brussels, Belgium

TC3 – Fire Safety Chairman: Prof. Paulo Vila Real Secretary: Prof. Martin Mensinger Date: 5–6 October 2017, Zurich, Switzerland

TWG 7.9 – Sandwich Panels & Related Subjects Chairman: Dr. Thomas Misiek Date: 29–30 June 2017, Helsinki, Finland

TC8 – Structural Stability Chairman: Prof. Bert Snijder Vice-Chairman: Mr. Richard Stroetmann Secretary: Dr. Markus Knobloch Date: 19 May 2017, Dresden, Germany; 27 October 2017, Coimbra, Portugal

TWG 8.3 – Plate Buckling Chairwoman: Prof. Ulrike Kuhlmann Secretary: Dr. Bettina Braun

TWG 8.4 – Buckling of Shells Chairman: Prof. John Michael Rotter Secretary: Prof. Spyros Anthony Karamanos

TC9 – Execution & Quality Management Chairman: Mr. Kjetil Myrhe Date: 1 March 2017, Brussels, Belgium

TC10 – Structural Connections Chairman: Prof. Thomas Ummenhofer Secretary: Mr. Edwin Belder Date: 16–17 March 2017, Salerno, Italy; 9–10 October 2017, Berlin, Germany

TC11 – Composite Chairman: Prof. Jean-François Demonceau Secretary: Prof. Graziano Leoni Date: 28 April 2017

TC13 – Seismic Design Chairman: Prof. Raffaele Landolfo Secretary: Dr. Aurel Stratan Date: 2–3 February 2017

TC14 – Sustainability & Eco-Efficiency of Steel Construction Chairman: Prof. Luís Bragança Secretary: Ms. Barbara Rossi

TC16 – Wind Energy support structures Chairman: Prof. Peter Schaumann Vice-chairman: Prof. Milan Veljkovic Secretary: Ms. Anne Bechtel

TC6 – Fatigue & Fracture Chairman: Dr. Mladen Lukic Secretary: Prof. Stephen LochteHoltgreven Date: 4–5 May 2017, Stuttgart, Germany

TC7 – Cold-formed Thin-walled Sheet Steel in Buildings Chairman: Prof. Jörg Lange

TC news TC3 News Currently, TC3 consists of 1 honorary, 25 full and 19 corresponding members.

The committee meets once a year. The number of participants was 17 at the

ECCS news last ECCS TC3 Annual Meeting held at the Steel Construction Institute (SCI), Sunningdale Park, Ascot, UK on 22 and 23 September 2016.

Recent topics discussed within TC3 include the following: – Tensile membrane action of composite slabs in fire – Fire walls made of lightweight sandwich panels – Concrete-filled stainless steel tubular columns in fire – The influence of hot-dip galvanization on temperature development in unprotected steel members in fire – Fire resistance of composite integrated beams (slim-floor beams) – Temperature assessment of a vertical steel member subjected to localized fire – Fire scenarios in suspended ceilings and hollow floors – Fire behaviour of prefabricated composite floors with steel dowels – High-strength steel members in fire – Default critical temperatures of steel members with a class 4 cross-section – Fire design of steel structures with intumescent coatings – Behaviour of cold-formed steel elements in fire – Assessment of existing structures in fire The next meeting of TC3 is scheduled for 5 and 6 October 2017 in Zurich, Switzerland.

News from TC7 – TWG 7.9 ECCS TC7 TWG 7.9 has just started work on a European Recommendation (ER) for the design of sandwich panels carrying point or line loads. Examples of such loads are photovoltaic installations on roofs or additional cladding on walls, i.e. non-structural components fastened to the sandwich panel.

The ER deals with the design of the panels for these loads and also with the design of the fastenings for the nonstructural components. Design rules are given in the main text; basically, they are amendments to existing design rules (e.g. from EN 14509). Each subsection concludes with examples. The annex to the ER describes the determination of resistance values, each value in a separate annex. The determination may be based on calculations or tests (verification of calculated/theoretical values only or wholly based on tests).

TC8 News Twenty-six full members and 10 corresponding members from 18 European

and North American countries currently constitute TC8 “Structural Stability”.

The technical committee is active in the field of the stability of steel structures.

The aim of TC8 is to provide scientifically sound input on stability design rules to code-writing bodies – in particular, CEN/TC 250/SC 3 and its Project Team and Working Group on EN 19931-1 – and to consult with these bodies.

The main committee concentrates on the stability of individual members (beams and columns) and skeletal structures (frames and trusses), whereas its Technical Working Groups focus on the stability of plate (TWG 8.3, convenor: Prof. Ulrike Kuhlmann) and shell (TWG 8.4, convenor: Prof. J. Michael Rotter) structures.

The committee meets twice a year and the last meeting of TC8 was in Barcelona on 4 November 2016. The agenda included a discussion about the progress of an intended publication on the “Design of Slender Steel Structures

by FEM”. This publication will contain some theoretical background, recommendations for the numerical analysis of structural components and examples for validation and benchmarks, with a special focus on information relevant for practitioners. Another discussion focused on important items for investigation and research in the future. These items include the following: – numerical methods in stability – global behaviour of structures (system approach) – imperfections and residual stress distributions to be used for stability rules – general method – high-strength steel members During the meeting, seven other presentations were delivered by members and guests: Joachim Lindner gave a presentation on initial bow imperfections.

Jozsef Szalai and Ferenec Papp presented a study on the safety assessment of different stability design methods for beam-columns.

Alain Bureau reviewed a simplified method for verifying lateral torsional buckling resistance. This method is based on the principle that the member resistance may be verified by checking the lateral buckling resistance of the compressed flange.

Gunnar Solland gave a presentation on requirements for structural ductility.

Trayana Tankova and Luís Simoes da Silva reported about a recent experimental and numerical study on the buckling behaviour of web-tapered beams.

Benjamin Launert and Hartmut Pasternak gave a progress report on the development of methods for the consideration of welding imperfections in stability verification rules.

Andreas Taras presented an analytical formulation for in-plane beam-column buckling which makes use of the generalized slenderness concept.

The following members attended the meeting: David Brown, Marc Braham, Alain Bureau, Rolando Chacon (local host), Luís Simoes da Silva, Dan Dubina, Markus Knobloch (Technical Secretary), Ulrike Kuhlmann, Benjamin Launert (guest), Joachim Lindner, Enrique Mirambell (guest), Ferenc Papp, Bert Snijder (Chairman), Gunnar Solland, Richard Stroetmann, Jozsef Szalai, Trayana Tankova (guest), Andreas Taras, Carlo Urbano.

The next TC8 meeting will take place in Dresden on 19 May 2017.

News from the TC11 Prof. Riccardo Zandonini has stepped down as Chairman of TC11 in order to devote himself completely to a new engagement as Senator of the University of Trento.

A workshop on composite construction entitled “Latest developments in research, standardization and practice” was held in Delft on 20 October 2016.

The event was jointly organized by Milan Veljkovic and Roland Abspoel of TU Delft and by a TC11 committee to express the great appreciation for Prof.

Riccardo Zandonini’s outstanding and dedicated service to ECCS. Fifteen speakers, invited from among European scholars, presented the results of their research and standardization activities.

Prof. Roberto Leon also took part in the meeting by means of a video link and provided insights into recent developments in composite structures in the USA. The TC11 six-monthly meeting was held on 21 October. During the meeting, the state of the art document on shear connections was approved and the procedure for the final publication was set in motion. A second document on perforated shear connections was discussed and has achieved the preliminary approval of TC11.

At the end of the meeting, TC11 suggested Prof. Jean-François Demonceau as a possible new Chairman to the Technical Management Board, in accordance with ECCS rules.

The next meeting of ECCS-TC11 will be on 28 April 2017 (venue to be decided).

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ECCS news

TC13 News During 2016, TC13 activities were carried out together with CEN/TC 250/SC 8 – WG2 in order to start the official process of revising and amending chapters 6 and 7 of the current EN 1998-1.

In the meeting in Naples (Italy) on 7 and 8 April 2016, the Committee discussed the criticisms of the design rules and requirements provided by EC8 for

traditional systems and local ductility.

In addition, a document concerning the revision of second-order effects and pushover analysis was discussed and submitted to TC250/SC8.

Raffele Landolfo (TC13 chairman and CEN/TC 250/SC 8 – WG2 convenor) and Guo-Qiang Li (Professor at Tongji University in China and CSA representative) organized the First Euro-China workshop on the Earthquake

Resistance of Steel Structures, Shanghai (China), 26–28 October 2016. A delegation of five TC 13 experts (Raffaele Landolfo, Dan Dubina, Vincenzo Piluso, Ahmed Elghazouli, Herve Degee) were invited to present their research outcomes at this event.

The last TC13 meeting has been held in Aachen (Germany) on 2–3 February 2017.

Secretary. Mrs. De Boe is responsible for following up ECCS Technical Committees under the chairmanship of Prof.

Milan Veljkovic. She is reporting on European-funded projects (RFCS and H2020) and implementing the deliverables, developing technical activities such as training courses, apps, publications and the website.

her expertise in European lobbying and her commercial approach, making ECCS more visible to specific European Institutions, and will be developing commercial activities and sponsorship in training courses, the website, publications, apps and the ECCS Architectural Awards.

We are convinced that Head Office has now the skills it needs to make ECCS the benchmark organization at European level and to develop further activities for the benefit of its members and the structural steelwork sector as a whole.

Further ECCS news ECCS extends its engineering and lobbying skills with two new members of staff

ECCS Annual Meetings

Mrs. Hélène De Boe is a civil engineer from the Louvain School of Engineering, Belgium. She graduated in September 2016 after a traineeship in the Structural Long Products Department of Arcelormittal, conducting research on how horizontal structural systems in tall buildings impact on the environment by performing a life cycle assessment. She also had traineeship experience with BESIX, a very large Belgian general contractor where she implemented the Autodesk Vault software and modelled the construction phases of a lock with Revit (BIM Department). She joined ECCS on 17 October 2016 as Technical

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Mrs. Sophie Delair graduated in European Law at the Université de Droit in Tours, France, in 1993 and developed expertise in European institution lobbying and public affairs. She was also Secretary-General of the Pan and Pro Association (panels and profiles) until 2011, after which she developed a small business for importing products which she is still managing now. She will contribute

The Annual General Meetings were held in Stockholm at the kind invitation of the SBI (Stålbyggnadsinstitutet), which organized the event perfectly in the Jernkontoret Building.

The European Steel Bridges Awards Ceremony was held on the afternoon of 14 November 2016. It was a great success. Eight projects had been selected out of 29 submitted – three projects in each category, Road and Railway bridges and Pedestrian and Cyclist Bridges, out of which a winner was chosen for each category. This year, two special awards were submitted in new categories.

Details and photos are available on our website at www.steelconstruct.com.

ECCS news country, open the awards to European Ipos and to public voting. The new concept will be in operation for the next European Steel Design Awards Ceremony, which will be held in Copenhagen on 14 September 2017.

European Steel Bridge Awards 2016 – the winners The ECCS Steel Bridge Awards 2016 were awarded in Stockholm on Monday, 14 November 2016. The awards are given by the European Convention for Constructional Steelwork every two years to encourage the creative and outstanding use of steel in the construction of bridges. There are two main categories: Road and Railway Bridges and Pedestrian and Cyclist Bridges. In addition, the jury members decided to present two special awards: Special Social and Research Award and Special Engineering Award.

Winner, Pedestrian and Cyclist Bridges: The Schlossteg 2.0 The Schlosssteg 2.0 is a beautiful addition to its location. Taking into account the historical metallurgical heritage and its position on the “Iron Road”, the bridge suits its context and becomes a symbol of the past. This footbridge, with a single span of 60 m, has an aesthetically pleasing design and details. The lightweight and slender bridge was fabricated close to the site and slid into place using a smart erection process. Its integrated lighting suits the simple, unimposing architecture, and the choice of weathering steel represents a robust and long-lasting choice for the future.

The Technical Management Board saw the arrival of new members from the painting industry as an opportunity to re-activate the Technical Committee on Surface Protection (formerly TC4).

AGCCE is a global company active in Asia and Europe which produces resins for coatings. ECCS is pleased to open a new category of members in the shape of the paints and coatings community.

The objective is to improve the results of these products when applied to steel structures and, especially, to reduce the environmental impact of coatings when recycling or dismantling structures. This issue is of primary importance for our sector in the CIRCULAR ECONOMY trends.

It was also noticed that the finite elements method is of interest to all technical committees, which therefore must coordinate their work. A new series of design manuals will be launched, deal-

ing more practically with finite elements this time.

From the promotional point of view, it was decided to modify the organization of the awards and create a new concept, with more submissions per

Winner, Road and Railway Bridges: Sundsvall Bridge Sundsvall Bridge is a sophisticated solution to a complex problem. The bridge has a total length of 1420 m and spans varying from 88 to 170 m. With

Lasse Kilvær (right), Chair of the Architecture and Awards Committee, and Veronique Dehan (left), Secretary-General of ECCS, presented the award in the Pedestrian and Cyclist category to architect and engineer Ulrich Eder from Tragwerkstatt Ziviltechniker GmbH.

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ECCS news

Lasse Kilvær (left), Chair of the Architecture and Awards Committee, and Veronique Dehan (right), Secretary General of ECCS, presented the award in the Road and Railway category to architects Henrik Rundquist, Anders Hedås and Hanna Karasalo from &Rundquist and Magnus Lundberg from Trafikverket.

its curvature on plan and elevation and the low, horizontal profile, this large bridge manages to avoid dominating the setting, instead relating to the low undulating hills of Sundsvall. The detailing of the tapered pylons and the well-resolved heads allows the slender deck to appear to glide above the water. The steel superstructure was fabricated and erected in a remarkably short time using an impressive fabrication, logistics and erection concept. The bridge is a response to the high architectural expectations of its location and the reflections in the water bring this repetitive structure to life.

Special Social and Research Award: bridging MZAMBA The jury would like to give a special mention to the bridging MZAMBA project. The bridge, with a total length of 131 m, crosses a river that separates the inhabitants west of the river from necessary infrastructure, such as educational facilities, healthcare and general food supply, and thus improves their daily life. The design of the bridge accounts for the difficulties of the site conditions, with problematic soil at the abutments and difficult access. The project is a collaboration between the Carinthian University of Austria and a Swiss engineering firm, while the construction was carried out with mostly local labour and the NPO Build Collective. The construction of this bridge brought skills and knowledge of building techniques to local workers, enabling the possibility of

regular maintenance, and places the bridge firmly in the community of the River Mzamba.

Special Engineering Award: New Botlek Bridge The New Botlek Bridge is an example of outstanding structural and mechanical engineering. It demonstrates the possibilities of steel when building precise and complex structures. This new bridge brings substantial improvements to road, rail and ship traffic conditions in the area. The bridge has two lifting spans, each 102.5 m long, and a deck composed of three longitudinal truss girders. The steel structure enables compliance with the required precision and limited tolerances for the railway tracks and expansion joints. The concept is a convincing solution to this very complex situation. The challenging fabrication and erection processes were also remarkable achievements.

For more details, please visit the official website at steelbridgeawards.com

The 2016 Charles Massonnet Award This year, the President of the ECCS, Johan Löw, had the great honour and pleasure of presenting the 2016 Charles Massonnet Award to Mr. Jouko Kouhi from the Finnish Constructional Steelwork Association.

The jury appreciated his high level of expertise in connections, welding, stability and fatigue of steel structures. They underlined the fact that Mr. Kouhi had been an active member of various ECCS Technical Committees, TC6, TC8 and TC10, since 1979, also serving as the chair of the TC10 Connections Ad Hoc Group, ENV 1993 part 1.8, from 1999 to 2002.

Gala Dinner After the Awards Ceremony, ECCS members were able to attend the Gala Dinner organized by ECCS President Johan Löw, who had invited a number of distinguished guests. Everyone enjoyed a tasty typical Swedish meal while applauding our 2016 Charles Massonnet Awardee, Mr. Jouko Kouhi from Finland. The atmosphere was particularly friendly and warm in snowy Stockholm.

ECCS President Johan Löw (left) and Mr. Jouko Kouhi (right) from Finland

The jury emphasized his recognition at both national and European level and his expertise as an advisor for more than 18 different national and European organizations and Institutions such as: – Nordic Committee for Building Regulations – Federation of Finnish Metal, Engineering and Electrotechnical Industries (FIMET)

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ECCS news – Ministry of the Environment – European Committee for Standardization (CEN) – European Commission DGXII Science, Research and Development Despite his retirement, Jouko Kouhi is an untiring advisor/worker who is frequently seen in the offices of the Finnish Constructional Steelwork Association!

ECCS Annual Meetings The ECCS Board Members decided to adopt some resolutions that will have a substantial impact on the future of the global structural steel sector.

RESOLUTION 1 Following the Memorandum of Agreement signed with the Chinese Steel Construction Society in April 2015, ECCS has agreed to be a founding member of an international association of steel structures, the IASC (International Association of Steel Construction), together with 11 other associations at global level (AISC, ASI, CISC, CSCS, JSSC, KSSC, MSSA, SAISC, SCNZ, SSSS). The main objective is to achieve international exchange on technical issues and hold regular international technical conferences. The mission of this organization will be articulated around research, standards, promotion, awards and resolutions.

RESOLUTION 3 ECCS has agreed to organize a professional workshop in Brussels on 12 April 2017 in order to re-activate TC4 “Surface Protection”, inviting professionals from coating and painting manufacturers and their associated disciplines (resin, epoxy, polymer…) to participate in research and development in the field of the surface protection of steel structures. The objective will be to set up the mission and membership of this renewed technical committee and launch the first meeting in September 2017 in Copenhagen (within the framework of the Eurosteel Conference).

ECCS Presidents Jorn Nielsen, Director of the Danish Steel Institute (DSI) in Copenhagen, received the ECCS Medal of President from Johan Löw and will retain it until the next Annual Assembly on 14 September 2017.

RESOLUTION 2 ECCS has agreed to cooperate with SEI (Structural Engineering Institute) and ASCE (American Society of Civil Engineers) within the framework of the recently established GAD (Global Activity Division) in order to increase the visibility of the global issues that are impacting on the structural engineering profession. GAD and its mission will be introduced to ECCS members at the next AGM in Copenhagen and during the Eurosteel Conference. The main mission will be to act as the voice of structural engineers and promote high-quality services that are international in scope.

ECCS President Johan Löw (right) and Jorn Nielsen (left)

ECCS R&D activity in 2016

– Valorization of innovative anti-seismic devices – RFCS project

– 18 months (July 2016 q December 2017) – Total budget: €995,660 – ECCS budget: €74,152 – Coordinator: National Technical University of Athens, Greece – Partner institutions: UPT (RO), POLIMI (IT), UNINA (IT), UNIPI (IT), RWTH (DE), IST (PT), UACEG (BU), Uhasselt (BE), MSE (DE), ECCS (BE), UNICMI (IT) – Objective: Valorization action for 12 innovative steel-based dissipative devices recently developed in Europe within the framework of RFCS, EU and national research projects and suitable for buildings in seismic regions. A series of information, designand code-oriented documents will be produced and disseminated to architects, engineers, construction companies, students and other partners of the construction sector. The borders between EN 1998 and EN 15129 will be established. A procedure to determine reliable behaviour factors will be drafted and case studies with application examples prepared.

– Overall slenderness-based direct design for strength and stability of innovative hollow sections – RFCS project – 36 months (July 2016 q June 2019) – Total budget: €1,573,684 – ECCS budget: €69,834 – Coordinator: University of Munich (DE) – Partner institutions: Imperial (UK), IST (PT), ECCS (BE), CTICM (FR), CONDESA (ES), ULaval (CAN) – Objective: The aim of HOLLOSSTAB is to develop new, direct design rules regarding strength and stability checks based on the “overall interaction concept”. Addressing both usual steel grades as well as tomorrow’s high-strength steels, this project will allow for more economic, innovative, “thin-walled” hollow section members. Simpler and faster design procedures will be developed, and both numerical tools and the latest statistical production data will be made available to structural engineers. These goals will be tackled by means of extensive numerical, analytical and ex-

Steel Construction 10 (2017), No. 1

87

ECCS news / Book reviews

Book reviews Kido, E. M.: Aesthetics of Railway Stations in Europe and Japan (in Japanese: Tetsudo--eki no utsukushi-sa ~ Nihon no eki, Yo-roppa no eki ~), Institution for Transport Policy Studies (Ippan zaidanho-jin unyu so-go- kenkyu-jo) and CTI Engineering Co., Ltd., Research Center for Sustainable Communities (Kabushiki gaisha kensetsu gijutsu kenkyu-jo, Kokudo bunka kenkyu-jo) Tokyo, 2016, 172pp., 198 coloured illustrations, 21 w 30 cm, softcover, JPY 2500, ISBN 978-4-903876-68-9 Available by direct order from Institution for Transport Policy Studies, Tokyo: Fax 81 3 5470 8411, Tel. 81 3 5470 8410; www.jterc.or.jp/tosho/pdf/Aesthe tics2016-10.pdf or https://www.govbook.or.jp/book/detail.php?product_ id=312626

Pictures on cover: top: Gare de Liège-Guillemins (2009, Belgium, arch.

Santiago Calatrava) Bottom: Asahikawa Station (2011, Hokkaido, arch. Hiroshi Naito-)

perimental programmes, with the close collaboration of key European hollow section manufacturers from different EU countries.

ECCS Design Guides and European Recommendations now available from Amazon ECCS has launched its technical publications in Amazon. The titles now available through this platform are the ECCS Design Guides and the European Recommendations covering several steel construction topics such as buckling of shells, sandwich panels, etc. The following titles are now available: – European Recommendations for the Determination of Loads and Actions on Sandwich Panels – European Recommendations on the Stabilization of Steel Structures by Sandwich Panels

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Steel Construction 10 (2017), No. 1

– Preliminary European Recommendations for the Design of Sandwich Panels with Openings – A state of the art report – Membrane Actions of Composite Structures in Case of Fire – Concepts and Methods for Steel Intensive Building Projects – Energy Efficiency of Lightweight Steel-framed Buildings – Guide to the CE Marking of Structural Steelwork – Buckling of Steel Shells – European Design Recommendations, 5th ed., revised 2nd impression The ECCS will continue to increase the publications offered via this online platform in the near future.

Readers of this eminent journal might possibly recognize the author of this book. Together with the co-author, the reviewer, she has contributed a few articles published in Stahlbau and its younger relative, Steel Construction – Design and Research, [1]–[5] and was also the interviewee in a Stahlbau Interview [6].

Recently, she compiled this book with the aim of presenting to the Japanese professional community the aesthetic status of Japanese railway stations versus that observed currently in Europe. In the past she has made several study visits to the leading countries of the European Union and has collected extensive material that, together with that gathered in Japan, became the foundation of the present book. She has also published (in English) various articles about the aesthetic issues of railway stations in Japan and Europe [7]–[9].

Book reviews / Discussion As mentioned, the book published in Japan is directed at the professional community in that country and therefore it was written in Japanese. However, a comprehensive English summary and commentary on the final 32 pages of the book can serve as a corresponding reference for those who do not speak Japanese. The chapter numbers (1–6) of the English summary correspond directly to those of the full Japanese version. The book is illustrated with coloured photographs of various stations in Europe and Japan. Apart from two, all the figures in the book are the author’s own. Unfortunately for English readers, the figures in the book have no English captions. However, a thorough study of the full English characteristics of the book, even by itself, will certainly assist in overcoming that deficiency, and the contents, as well as the message, of the book can be fairly well understood.

In order to express very briefly the formal contents of book, it is best to use the author’s own words: “Chapter 1 introduces railway stations in the context of the role of the stations, station design in the past in Japan and Europe, and the recent ‘station renaissance’. Chapter 2 describes the aesthetics of railways and factors of aesthetic station design, such as form of station building, the aspects of interior design, daylight and lighting, universal design, image-based elements and public art, commercial function, advertisements, railway tracks, the appearance of trains from outside and their interior design. The aesthetics of railways also includes the beauty of landscape, which can be seen from the trains. Chapter 3 and Chapter 4 respectively present European and Japanese railway stations and their aesthetic aspects. European stations have been discussed in the examples of the Deutsche Bahn (DB) in Germany, the Société Nationale des Chemins de fer Français (SNCF), the Régie Autonome des Transports Parisiens (Paris Metro), and also on the example of the LRT in France. The discussion of Japanese stations has been based on the examples of stations belonging to the East Japan Railway Company (JR East), Tokyo Metro and the Yokohama Minatomirai Railway Company. Chapter 5 presents the challenges of station design in Japan compared with the design of European stations. The total design has been shown here as the objective of aesthetic station design. Some aspects of design have been discussed in the example of the comparison of the stations on the Jubilee Line Extension – (JLE) in London with the Oedo Line in Tokyo. Chapter 6 formulates a proposal

towards better design, with aspects such as the promotion of design competitions, the importance of design that reflects the regional characteristics and the Context Sensitive Design.”

This book is, in Japan, most certainly the first of its kind. Japanese professionals will surely profit from it considerably. It could be relevant to architects, engineers, transport planners, railway companies and those involved in the planning and design of railway stations.

European readers, as stated, will get some benefits from the English commentary to the book in Chapter 7.

Therefore, the book can be recommended to the European professional community as well.

[1] Kido, E. M.; Cywin´ski, Z.: The new steel-glass architecture of buildings in Japan. Steel Construction – Design and Research 6(2013), 3, pp. 229–237.

DOI: 10.1002/stco.201320012 [2] Kido, E. M.; Cywin´ski, Z.: The new steel-glass architecture of railway stations in Japan. Steel Construction – Design and Research 7(2014), 3, pp. 208– 214. DOI: 10.1002/stco.201420022 [3] Kido, E. M.; Cywin´ski, Z.: The new steel-glass architecture of air terminals in Japan. Steel Construction – Design

and Research 7(2014), 4, pp. 246–251.

DOI: 10.1002/stco.201420034 [4] Kido, E. M.; Cywin´ski, Z.: The new steel-glass architecture of passenger service stations on expressways in Japan. Steel Construction – Design and Research 8(2015), 3, pp. 210–215.

DOI: 10.1002/stco.201520024 [5] Kido, E. M.; Cywin´ski, Z.: The colours of steel bridges in Japan – principles and examples. Stahlbau 85(2016), 3, pp. 181–194. DOI: 10.1002/ stab.201610355 [6] Stahlbau Interview: Arbeiten in Japan – Harmonie und Vermeidung offener Konflikte sind wichtig. Stahlbau 84(2015), 3.

[7] Kido, E. M.: Aesthetic issues of railway stations in Japan and Europe.

IABSE Symp., Weimar, 2007, Proc., pp.

260–261; full text CD A-0041.

[8] Kido, E. M.: Stations for people – Important factors in station design.

IABSE Symp., Madrid, 2014, Reports, pp. 850–851; full text CD-ROM. [9] Kido, E. M.: New stations in Japan reflecting new age. World Engineering Conf. & Conv., Kyoto, 2015, Abstract Book PS 5-1-5; full text CD-ROM.

Zbigniew Cywin´ski, Gdan´sk

Discussion Discussion on the paper “Experimental and numerical assessment of RHS T-joints subjected to brace and chord axial forces”, by Nizer et al., Steel Construction 9 (2016), No. 4, pages 315–322.

The authors have made an experimental and numerical study on T-joints with the brace axially loaded incrementally until failure, while the chord was subjected to a constant level of axial force (either zero, compression or tension).

It is important to note that a T- or Y-joint achieves equilibrium by developing shear forces in the chord member, which in turn create a bending moment distribution in the chord. These bending moments cause normal stresses at the chord connecting face, which significantly influence the joint behaviour. The authors, like some other researchers, have unfortunately made a fundamental mistake of not including the influence of this bending stress in the chord connecting face, when the brace is subjected to increments of load until the ultimate value. Table 1 shows how the chord stress factor “n” is to be correctly calcu-

lated, based on both the combined axial and bending normal stresses at the joint in the chord connecting face. This is adopted by all current international design provisions. This is vastly different to the calculations by the authors, as evident by their Table 4 values. Table 1 herein shows the expected trend: as the compression “pre-stress” level in the chord connecting face decreases, the connection strength (as measured by the axial force in the brace) increases. At no time did the experiments by the authors achieve tensile normal stress in the chord connecting face at the joint (positive n). The interpretation of their results is therefore incorrect and their conclusions – and their comments on codes and international design recommendations [1], [2] – are therefore misleading and inappropriate.

It is therefore considered important to inform investigators of how to correctly carry out and to interpret their research on isolated joints in the future.

Investigations on T-joints are usually based upon simple supports at the chord ends equidistant from the brace, where the distance between the supports influ-

Steel Construction 10 (2017), No. 1

89

Discussion ences the bending moment at the brace.

Although experiments and numerical models can be performed on joints with any chord span (perhaps for the purpose of validating finite element (FE) models), it has been shown [3], [4], [5] that chord end boundary conditions have an effect on joint strength unless a minimum distance between the chord end supports is achieved. Thus, design recommendations for T-joints ultimately need to be based on experiments or numerical models employing this minimum chord length. Chord stress functions for rectangular hollow section Tand X-joints are given in [6], [7], which illustrate the background research to the influence functions adopted in current international design guidelines [2].

In FE modelling of T-joints, the influence of normal stress due to chord bending on the joint strength can be eliminated by applying compensating inplane bending moments at both chord ends, which cancel out the bending stress in the chord connecting face on either side of the brace, as shown in Fig. 1. Demonstrations of this procedure are available in the literature [4], [5]. In this manner, by removing the chord bending effect, the influence of just chord axial forces upon joint strength can be studied. In experimental work, this can be achieved by cantilevering the chord ends beyond the supports and applying counteracting loads incrementally, to give the compensating in-plane bending moments at both the chord ends, as for the FE modelling.

[1] EN 1993-1-8:2005, Eurocode 3: Design of steel structures – Part 1-8: Design of joints, incorporating Corrigenda

Table 1. Corrected values of “n”, for the authors’ experimental tests (excluding TN02N0 which had different chord end restraint), based on experimental data and assuming EN 10219 section properties

Test

n " (M0/Wpl,0fy0) (N0/A0fy0) (non-dimensional normal stress ratio in chord connecting face at the joint)

N1,EXP [kN] in brace (Table 4 by authors)

TN05N70-

–(52.9/2)(0.350/18.57) – 429.6/597.14

" –1.22

–52.9

TN06N50-

–(68.6/2)(0.350/18.57) – 306.9/597.14

" –1.16

–68.6

TN01N0

–(84.5/2)(0.350/18.57) – 0

" –0.80

–84.5

TN03N50

–(109.2/2)(0.350/18.57) 306.9/597.14 " –0.52

–109.2

TN04N70

–(109.0/2)(0.350/18.57) 429.6/597.14 " –0.31

–109.0

December 2005, September 2006, July 2009 and August 2010, European Committee for Standardization, Brussels, 2005–2010.

[2] ISO 14346: Static design procedure for welded hollow-section joints – Recommendations, International Organization for Standardization, Geneva, Switzerland, 2013.

[3] van der Vegte, G.J.; Makino, Y.: Further research on chord length and boundary conditions of CHS T and X joints. Advanced Steel Construction 6(2), 2010, pp. 879–890.

[4] van der Vegte, G.J.; Makino, Y.: Ultimate strength formulation for axially loaded CHS uniplanar T-joints. International Journal of Offshore and Polar Engineering 16(4), 2006, pp. 305–312.

[5] Voth, A.P.; Packer, J.A.: Numerical study and design of T-type branch plateto-circular hollow section connections.

Engineering Structures 41, 2012, pp.

477–489.

[6] Wardenier, J.; van der Vegte, G.J.; Liu, D.K.: Chord stress function for rectangular hollow section X and T joints.

Proceedings of the Seventeenth International Offshore and Polar Engineering Conference, Lisbon, Portugal, 2007, pp. 3363–3370.

[7] Packer, J.A.; Wardenier, J.; Zhao, X.L.; van der Vegte, G.J.; Kurobane, Y.: Design guide for rectangular hollow section (RHS) joints under predominantly static loading. CIDECT Design Guide No. 3, 2nd ed., Comité International pour le Développement et l’Étude de la Construction Tubulaire, Geneva, Switzerland, 2009.

Discussers Prof. dr. Jeffrey Packer Department of Civil Engineering University of Toronto 35 St. George Street, Toronto Ontario M5S 1A4, Canada jeffrey.packer@utoronto.ca Prof. dr-ing. Ram Puthli (Corresponding author) Karlsruhe Institute of Technology (KIT) Steel & Lightweight Structures Research Centre for Steel, Timber & Masonry Otto-Ammann-Platz 1 76131 Karlsruhe, Germany puthli@kit.edu Dr. G.J. van der Vegte Van der Vegte Consultancy Zwolle, The Netherlands gjvdv2001@yahoo.com

Fig. 1. Bending moment distributions in the chord member caused by transverse forces, compensating moments at the chord ends, and the combined effect, which results in zero normal stress in the chord connecting face beside the brace (n " 0)

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Steel Construction 10 (2017), No. 1

Prof. dr. ir. Jaap Wardenier Faculty of Civil Engineering & Geosciences Delft University of Technology P.O. Box 5048, 2600GA Delft, The Netherlands j.wardenier@tudelft.nl also: Department of Civil & Environmental Engineering National University of Singapore #E1A-07-03, 1 Engineering Drive 2, Kent Ridge, Singapore 117576

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Steel Construction 2/2017 Ulrike Kuhlmann, Jakob Ruopp Steel-to-concrete joints with large anchor plates under shear loading Dennis Lam Recent research on composite beams with demountable shear connectors Marko Pavlovic, Milan Veljkovic FE validation of push-out tests using bolts as the shear connectors Milan Spremic, Zlatko Markovic, Milan Veljkovic Recommendation for design of grouped shear studs

Demountable shear connectors are of growing interest to researchers and practitioners as a way of facilitating reuseable composite structures

Wojciech Lorenc The model for general composite section resulting from introduction of composite dowels Mark Lawson, Francois Hanus Design of composite beams with large web openings (subject to change without notice)

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