The Bridge and Structural Engineer by IABSE

The Bridge & Structural Engineer Indian National Group of the International Association for Bridge and Structural Engineering ING - IABSE

Contents :

Volume 46, Number 3 : September 2016

Editorial ●

From the Desk of Chairman, Editorial Board : Mr. Alok Bhowmick

●

From the Desk of Guest Editor : Professor Mahesh Chandra Tandon

Highlights of ING-IABSE Event & IABSE Congress at Stockholm Highlights of the ING-IABSE Annual Day-2016 and Technical Presentations held on 21st May, 2016 at New Delhi

●

19th IABSE Congress held at Stockholm, Sweden from September 21-23, 2016 by Alok Bhowmick

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Special Topic : Tall Structures 1.

Design of Tall Piers for Railway Bridges in North-East India Sumantra Sengupta, Amitabha Ghoshal

High-Rise Buildings in India - Code Improvement Required Vipul Ahuja, B.M. Ahuja

Challenges Involved in Design and Construction of 275M Tall RCC Chimneys Vinay Gupta

Structural Design of Tall Buildings in India Satish Kumar SR

Experience with Medium-Tall Buildings in Germany - A Case Study Boris Reyher, Mike Schlaich

Some Basic Guidelines in Use of Steel Rebars in High Rise Buildings N.V. Nayak, Shantilal Jain

Tall and Handsome Natural Draft Cooling Tower (NDCT) Shells in India V.N. Heggade

Tall Buildings – A Review of Forms, Systems and Analysis Subhash Mehrotra

Contents

●

Research Paper 1.

Damage Identifications in a Through Type Steel Truss Bridge Model Suresh Kumar Walia, Hemant Kumar Vinayak, Ashok Kumar, Raman Parti

Panorma ●

Office Bearers and Managing Committee - 2016

The Bridge and Structural Engineer

Volume 46 │ Number 3 │ September 2016

The Bridge & Structural Engineer ING - IABSE

Journal of the Indian National Group of the International Association of Bridge & Structural Engineering

December 2016 Issue of the Journal will be a Special Issue with focus on Professional IssueS Confronting The Structural Engineering Profession in India Salient Topics to be covered are : 1.

Ethical and Professional Issues Confronting Indian Engineers

Need for Regulation of Engineering Profession in India

New Challenges Facing Engineers & Engineering Organizations

Role of Civil Engineers in Built Environment

Vision 2025 for Engineering Fraternity in India

The Bridge & Structural Engineer ING - IABSE

Journal of the Indian National Group of the International Association of Bridge & Structural Engineering

March 2017 Issue of the Journal will be a Special Issue with focus on Bridge Engineering Salient Topics to be covered are : 1. 2. 3. 4. 5. 6. 7. 8.

Aesthetics in Bridge Construction Composite Deck Systems Fatigue and Fracture Critical Bridge Inspection Floating Bridges Service Life Predictions for Reinforced Concrete Bridges Accelerated Bridge Construction to Rehabilitate Aging Highway Structures Efficient Methods for Upgrading or Reinforcing Existing Bridges Future of Bridge Designs

Those Interested to contribute Technical papers on above themes shall submit the abstract by 31st December 2016 and full paper latest by 31st January 2017 in a prescribed format, at e-mail id : ingiabse@bol.net.in, ingiabse@hotmail.com

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

The Bridge and Structural Engineer

September 2016

Disclaimer : All material published in this B&SE journal undergoes peer review to ensure fair balance, objectivity, independence and relevance. The Contents of this journal are however contributions of individual authors and reflect their independent opinions. Neither the members of the editorial board, nor its publishers will be liable for any direct, indirect, consequential, special, exemplary, or other damages arising from any misrepresentation in the papers. The advertisers & the advertisement in this Journal have no influence on editorial content or presentation. The posting of particular advertisement in this Journal does not imply endorsement of the product or the company selling them by ING-IABSE, the B&SE Journal or its Editors.

Front Cover : Reaching a height of over 160m, Latitude is a signature residential tower by the M3M group, located in Golf Course Extension road in Sector 65, Gurgaon. The height of the tower is further enhanced by strong vertical cladding lines topped by a dramatic cantilevered roof, soaring above the tower. The Tower has 4 basements, Ground and 43 Upper floors. The basements are meant for parking whereas the floors from Ground to 41st have 160 high end air conditioned residences. The uppermost 2 floors (i.e. 42nd and 43rd) are for state of the art sky Club with gym, Jacuzzi, spa, paddle pool, squash court, event space, cinema and Observation Deck with photovoltaic array…etc. Generous landscaping effectively screens the pool deck from the dropoff area, whilst hinting at the luxury within. The project has many modern amenities like sports hall, restaurant, jogging track, large pool area etc.

Editorial Board Chair: Alok Bhowmick, Managing Director, B&S Engineering Consultants Pvt. Ltd., Noida

Members: Mahesh Tandon, Managing Director, Tandon Consultants Pvt. Ltd., New Delhi A.K. Banerjee, Former Member (Tech) NHAI, New Delhi Harshavardhan Subbarao, Chairman & MD, Construma Consultancy Pvt. Ltd., Mumbai Nirmalya Bandyopadhyay, Director, STUP Consultants Pvt. Ltd., New Delhi Jose Kurian, Former Chief Engineer, DTTDC Ltd., New Delhi S.C. Mehrotra, Chief Executive, Mehro Consulants, New Delhi

Advisors: A.D. Narain, Former DG (RD) & Additional Secretary to the GOI N.K. Sinha, Former DG (RD) & Special Secretary to the GOI G. Sharan, Former DG (RD) & Special Secretary to the GOI A.V. Sinha, Former DG (RD) & Special Secretary to the GOI S.K. Puri, Former DG (RD) & Special Secretary to the GOI R.P. Indoria, Former DG (RD) & Special Secretary to the GOI S.S. Chakraborty, Former Chairman, CES (I) Pvt. Ltd., New Delhi B.C. Roy, Former Senior Executive Director, JACOBS-CES, Gurgaon Published: Quarterly: March, June, September and December Publisher: ING-IABSE C/o Secretary, Indian National Group of the IABSE IDA Building, Ground Floor (Room Nos. 11 and 12) Jamnagar House, Shahjahan Road New Delhi-110011, India Phone: 91+011+23388132 and 91+011+23386724 E-mail: ingiabse@bol.net.in, ingiabse@hotmail.com, secy.ingiabse@bol.net.in

The Bridge & Structural Engineer, September 2016

B&SE: The Bridge and Structural Engineer, is a Quarterly journal published by ING-IABSE. It is one of the oldest and the foremost structural engineering Journal of its kind and repute in India. It was founded way back in 1957 and since then the journal is relentlessly disseminating latest technological progress in the spheres of structural engineering and bridging the gap between professionals and academics. Articles in this journal are written by practicing engineers as well as academia from around the world.

Submission of Papers: All editorial communications should be addressed to Chairman, Editorial Board of Indian National Group of the IABSE, IDA Building, Ground Floor, Jamnagar House, Shahjahan Road, New Delhi-110011. Advertising: All enquiries and correspondence in connection with advertising and the Equipments/Materals and Industry News Sections, should be addressed to Shri R.K. Pandey, Secretary, Indian National Group of the IABSE, IDA Building, Ground Floor, Jamnagar House, Shahjahan Road, New Delhi-110011.

The Bridge and Structural Engineer

Volume 46 │ Number 3 │ September 2016

Journal of the Indian National Group of the International Association for Bridge & Structural Engineering

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From the Desk of Chairman, Editorial Board

It is with great pleasure that we bring out this special issue of the journal with the theme of “Tall Structures”. Tall structures are now a common features in the skyline of many cities throughout the world. India’s rapid economic growth led by industrialisation and urbanisation and escalating cost of land has led to critical need for building cooling towers, chimneys and creating ‘vertical cities’ in the form of ‘tall buildings’. With more than 40% of India’s population expected to live in urban areas by 2030, there is a growing demand for residential and commercial space. All stake holders are exploring the vertical space to overcome these challenges. Tall structures are becoming increasingly complex in plan geometry, shape, height, variation from floor to floor. The buildings have become much taller, and the importance of optimal structural design is becoming more significant in taller structures due to the premium for heights. The issues of vertical transportation, fire and life safety are also very critical in tall buildings. Because of their enormous scale, tall buildings are constructed with an abundant amount of resources and consume lots of operational energy during occupancy. Clients are becoming increasingly more demanding and expect the structural engineers to take all these factors and come out with innovative design solutions at optimised cost. It is an urban imperative to design tall structures for sustainability for counteracting the

Volume 46 │ Number 3 │ September 2016

effects of climate change. The design, detailing and construction of such tall structures requires detailed knowledge, experience and expertise to properly understand the material being used for construction as well as the sensitivity of such structures to static and dynamic loadings. The large difference in the environmental condition (temperature, humidity, wind pressures) that can occur between the top and bottom of tall building needs to be accounted for in the design. This special edition of the journal captures the snapshot of current ‘best practice’ in the design, detailing and construction of ‘Tall Structures’ and will hopefully be a useful insight to all readers connected with Tall Structures. We are once again privileged to have Prof. Mahesh Tandon as our Guest Editor for this issue of the journal. Prof. Mahesh Tandon hardly needs any introduction. He is a well known personality in the field of structural engineering. I am grateful to him for sparing his time generously in making this special issue of the journal possible. Happy reading !

(AloK BHowMICK)

The Bridge and Structural Engineer

From the Desk of Guest Editor

A large variety of Tall Structures are being constructed in India and the rest of the world. There have been many reasons why man wants to “reach for the sky” by building tall structures. One reason is certainly to demonstrate the technology possessed by the builder. Other reasons are based on necessity. Buildings going up vertically is because of the high cost of land in the urban environment and it may not be economical or feasible for a city to only grow horizontally. Another reason could be to discharge pollutants through tall chimneys. Yet another reason could be to cool the water (eg Cooling Towers) and re-cycle it by creating an upward draft while the hot water is on its downward journey. Tall structures have to cater to large wind and earthquake forces apart from gravity loads which require to be catered to in all structures anyway. Innovations in structural engineering are required so that the premium that has to be added over and above the requirements of gravity loads is kept to the minimum. The papers discussed in this issue of the Journal are briefly discussed below. Dr. Reyher and Prof. Schlaich have presented an interesting 90m tall building in the town of Bochum in Germany the elevation of which reminds one of a “slip disc”. The building is

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named Exzenterhaus, which literally Eccentric Building. The important challenge of the Architect and the structural Engineer was in the incorporation of a World War II bunker with the new 18 office floors constructed above. Mr. V.N. Heggade has presented brief analytical and design details of a 202m tall Natural Draft Cooling Tower at Kalisindh which is reportedly the tallest such structure in the world. The thin shell (160mm) combined with the large three dimensional structure required sophisticated construction systems. The structure is a hyperboloid of revolution giving a curvature in the meridional and circumferential directions. The membrane structure is predominantly in compression for symmetrical vertical loads. However, the analysis is also necessary for wind loading as well as perturbations caused at the base to design for flexural effects. Mr. Heggadediscusses the provisions of various codes of practice for some critical issues. Mr. Subhash Mehrotra has given a list of Tall Buildings (more than 100 stories) in his paper. He has enumerated the more common types of structural systems used for tall buildings and also mentioned about outrigger-braced structures which have been finding favour with designers in recent times. Mr. Mehrotra illustrates a type of Friction Damper which is activated at 130% of the design wind shear for

Volume 46 │ Number 3 │ September 2016

enhancing earthquake resistance. Other types of common dampers are also discussed in the paper. Mr. Vipul Ahuja and Prof. B.M. Ahuja discuss the shortcomings of the Indian codes in their paper. They have brought out the deficiencies in our codes by quoting from ACI 318. These include different response reduction factors to be taken in the two directions, shear wall design procedure and ductile detailing measures given in IS:13920. Mr. Vinay Gupta has summarized the present status of design and construction of tall chimneys in India. The usual height of chimneys for Thermal Power Plants is 275m. The Construction is carried out using slip forming or jump form. Special care is required during slip forming of tapering chimneys. Both single flue and multi flue chimneys have been given an exposure in the paper. The structural analysis and design has been discussed briefly for the two major loading. I.e., wind and earthquake. Mr. Sengupta and Mr. Ghoshal’s paper discuss tall piers for Railway Bridges located in the North East part of India, where there are requirements of crossing deep gorges. Various design alternatives considered have been discussed briefly and the selected alterative with simply supported through type steel trusses for spans of 106m c/c of piers has been elaborated. The substructure and foundations in the difficult terrain also find mention in the paper. Site specific seismic study was carried out at IIT Kharagpur for arriving at design parameters. Wind tunnel tests were carried out at IIT Kanpur. The construction is currently in progress.

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Prof Satish Kumar has presented a paper that summaries the basic considerations that must be taken into account for tall buildings 10 to 40 stories high which are most common. Three case studies have also been given. Dr. N.V. Nayak and Mr. S.H. Jain have presented some important considerations while using steel reinforcement bars in high rise buildings. They have highlighted that TMT bars are manufactured by a sophisticated process which is available with primary producers only. Minimising Sulphur and Phosphorus are essential for ductility and should be ensured. In vertical elements of the building (core walls and columns) in tall buildings the reinforcement is highly congested and use of Self Compacting Concrete (SCC) is essential. They have pointed out that many tall buildings have recently been constructed in Mumbai and none of them have used SCC despite density of reinforcement being as high as 590 kg/cum in one case. The authors have also recommended that higher strength bars than those being presently manufactured in India (Fe 550D) should be encouraged. They have mentioned that in Japan, reinforcement bars of the equivalent of Fe 980 are available, while in the USA Fe 690 is being used. All the above issues are important for tall buildings in particular and the Indian codes of practice need to be updated/ upgraded, now that provisions of a new code on tall buildings is under active discussion.

(PRoF. MAHESH CHANDRA TANDoN)

The Bridge and Structural Engineer

Brief Profile of Prof. Mahesh Chandra Tandon Prof. Tandon, Managing Director of Tandon Consultants Pvt. Ltd., is an international expert in the field of Structural Engineering. Many of the structures designed by Prof Tandon have been widely acclaimed and have received recognition in India as well as internationally. He is President, Indian Association of Structural Engineers (2015-16) and President, Indian Society of Wind Engineering (2015-16). Under the AICTE-INAE program he was appointed 2015). From the Institution of Engineers Tandon received Industry Excellence Awart (2015). He is Honorary Fellow of Indian Concrete Institute and Chartered Engineer and Fello of American Society of Civil Engineers. Prof. Tandon has made significant contributions in the development of a culture for innovation in structural engineering both within and outside his organization by sharing his expertise and experience. His special areas of interest also include motivating the next generation to adopt Civil Engineering as their profession and vocation in life.

FORTHCOMING EVENT OF THE ING-IABSE The Indian National Group of the International Association for Bridge and Structural Engineering (ING-IABSE), in co-operation with Government of Karnataka, PWD is organising a two day Workshop on “Inspection, Repair and Rehabilitation of Bridges and Flyovers” on 20th and 21st (Friday-Saturday), January 2017 at Bangaluru. Registration Registration is to be done by paying the required Registration Fees to the “Secretary, Indian National Group of the IABSE, New Delhi” through a cheque preferably by 16th January 2017. This will help in making advance arrangements.

Registration Fees Fee Rs.

Service Tax 15%

Total Rs.

Members, ING-IABSE

4000/-

600/-

4600/-

Non-Members, ING-IABSE

6000/-

900/-

6900/-

Young Engineers/ Students (under 35 years)

2000/-

300/-

2300/-

For any enquiry about the above Workshop, please address to the following:

Shri KB Sharma Under Secretary Indian National Group of the IABSE IDA Building (Ground Floor) Jamnagar House, Shahjahan Road New Delhi-110011 Mob: 9871481089 Tel:011-23388132, 23386724, E-mail: ingiabse@hotmail.com; ingiabse@bol.net.in

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Volume 46 │ Number 3 │ September 2016

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HIGHLIGHTS OF THE ING-IABSE ANNUAL DAY-2016 AND TECHNICAL PRESENTATIONS HELD ON 21ST MAY 2016 AT NEW DELHI The Indian National Group of the IABSE had organised its Annual Day-2016 along with technical presentations on “Proposed Tunnel in Shiradi Ghat in Karnataka” by Dr Florian Krenn, Managing Director, GEOCONSULT India Pvt. Ltd. and “Ganga Bridge - Patna” by

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Shri Atul D Bhobe, Managing Director, SN Bhobe & Associates Pvt. Ltd. at India Habitat Centre, Lodhi Road, New Delhi on 21st May 2016. The Annual Day 2016 and presentations was attended by about 75 delegates from various parts of India. The presentations was highly acclaimed.

A view of the Dais during the Inauguration

Dr. Florian Krenn during his Technical Presentation

Shri Atul D Bhobe during his Technical Presentation

A view of the Audience during the Technical Presentation

Shri D.O. Tawade, Chairman, ING-IABSE Delivering his Welcome Address during the Inauguration

Volume 46 │ Number 3 │ September 2016

The Bridge and Structural Engineer

In the 56th Annual General Body Meeting, elections under different rules were held for

Managing Committee Members. In the 106th Managing Committee meeting, the elections were held for Members of the Executive Committee. Shri D.O. Tawade was elected as Chairman, Shri B.N. Singh, Shri Alok Bhowmick and Shri A.K.S. Chauhan were elected as Vice-Chairmen of the Group. Shri R.K. Pandey and Shri Ashish Asati would continue to act as Secretary and Director of the Group.

A view of the Dais during the 105th Managing Committee Meeting

A view of the 56th Annual General Body Meeting

Besides the above, the following Annual Meetings of the Group were also held on the 21st May, 2016 at India Habitat Centre, New Delhi. □

105th Managing Committee

□

56th Annual General Body

□

106th Managing Committee

A view of the 106th Managing Committee Meeting

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Volume 46 │ Number 3 │ September 2016

A report of the 19th IABSE Congress held at Stockholm, Sweden from September 21-23, 2016 By Alok Bhowmick The recently concluded 19th IABSE Congress at Stockholm with its theme ‘Challenges in Design and Construction of an Innovative and Sustainable Built Environment’ was attended by four of our prominent members, Mr D O Tawade (ChairmanING-IABSE), Mr B.N. Singh (Vice Chairman, ING-IABSE) Dr. Harshavardhan Subbarao (Vice President, IABSE) & Mr. Alok Bhowmick (Vice Chairman, ING-IABSE). The congress report of 2871 pages is a wealth of information, comprising of 352 papers on seven sub-themes, covered in 62 sessions, with break-up as given below :

city centre (Pic 1). The congress was attended by more than 600 eminent practicing engineers from around the world. Indian participation was about 7-8 engineers, including officials from NHAI and Ministry of Road Transport and Highways.

A – Analysis (11 sessions / 67 papers) C – Construction & Production (6 sessions / 33 papers) F – Forensic Engineering (3 sessions / 15 papers) L – Loads (5 sessions / 29 papers) M – Materials (7 sessions / 36 papers) R – Repair & Maintenance (9 sessions / 55 papers) S – Structures (21 sessions / 117 papers)

Pic 1 : Venue for the Conference

Annual Meetings of IABSE were scheduled prior to the congress, on September 19–20, 2016 (Pics 2, 3 & 4).

In addition, there were following 6 keynote lectures delivered by experts in their respective fields: a)

“Sustainable Asset Management – A view from Asia” by Yozo Fuzino (JAPAN)

“A sustainability city is a city for people” by Helle Soholt (USA)

“Innovative Tunnelling in a Sustainable Built Environment” by Thomas Jesel, Switzerland

“Coastal Highway Route E39” by Borre Stensvold, Norway

“Trends Within Sustainable Bridge Operation and Maintenance” by Jens Sandager Jensen, Denmark

“Sustainable Concrete” by Karen Scrivener, Switzerland

The venue for the IABSE Congress 2016 was the ‘City Conference Centre’ situated in the heart of the

Volume 46 │ Number 3 │ September 2016

Pic 2 : Participants of one of the Annual Meetings of IABSE

A pre-congress one-day course on ‘Forensic Structural Engineering: Causes, Investigations and Prevention of Failures’was also organised by IABSE, on 20th September 2016, which was delivered by a team of experts on the subject led by Prof. Robert T. Ratay (USA) and supported by other team members

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Mr John Duntemann (USA), Dr Fabrizio Palmisano (Italy), Dr Karel Terwel (Netherlands) and Prof. Sven Thelandersson (Sweden). The scientific committee (SC) of this congress was chaired by Lennart Elfgren of Sweden. SC comprising of 70 members and ensures the quality of the papers presented at the conference. The conference format allowed for a variety of communication modes and encouraged interaction between speakers and delegates through active discussions and debates. Pic 5 shows a glimpse of some of the events during the conference.

Pic 3 : Interactions during Annual Meetings of IABSE

Pic 4 : Attendees of Working Committee WC-7 From Left to Right : Tatsuo Inada (Japan), Andrew Martin (Denmark), Guangli Du (Denmark), Ikuhide Shibata (Japan), Serge Montens (France), Ekasit Limsuwan (Thailand), Yue Liu (Germany), Martin Kirk (UK), B.N. Singh (India), Alok Bhowmick (India), D.O. Tawade (India)

Pic 5 : A Glimpse of some of the Events at the Congress

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Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

Recognition of Excellence :

Technical Visits :

In IABSE’s annual tradition of recognising excellence in structural engineering, The Shanghai Tower was given the Outstanding Structure Award (Pic 6), while Mr Hector Beade Pereda from spain received the coveted IABSE prize. Joining the circle of Honorary members are Professor Yozo Fujino of Japan and Ricardo Zandonini from Italy.

An informative technical visit to various important civil engineering projects was arranged by the organisers at the following places:

Pic 6 : Shanghai Tower

Pic 7 : Technical Visit to Stockholm Railway Bypass

Citybanan - Stockholm Railway Bypass :

Citybanan is a 6 kilometer long commuter train tunnel under central Stockholm. One part of the tunnel is designed and build as a submerged tunnel. This is a 400 meter long railway tunnel underneath the stream. Work with the tunnel was in its final stages at the time of Technical Visit (Pic 7).

Young Engineers Programme :

Slussen :

In conjunction with the growing demographic of Young Engineers and in line with the Young Engineers Programme (YEP) initiative of IABSE, the permanent committee approved the Student Membership (SIM) during the annual meetings which took place prior to the congress. The IABSE Student Membership (SIM) is open to all current students 27 years of age or younger, and includes all Individual Membership benefits with the quarterly journal Structural Engineering International (SEI) provided via online access only. The fee charged from students is less than 10% of the fee charged from members. This marks a great step for the association to encompass building structural engineers through an affordable membership fee and benefit from IABSE’s online resources and network.

Slussen, located between the Old town and Södermalm, is one of the most important historical junctions in Stockholm`s traffic system but after 80 years the existing structure is in a very poor condition. The structures including the foundation needs to be replaced and the area will be adapted to face the future demands from the citizens of Stockholm. The project involves demolishing the existing plant and, on the same spot, build a new interchange including a new sluice.

Haga City Hospital :

In 2025, the Haga City will be completed – an inner city area where Stockholm and Solna meet. The area stretches from Vasastan

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Volume 46 │ Number 3 │ September 2016

and

Karolinska

University

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and Norra Stationsgatan over to Karolinska University Hospital and Karolinska Institute. With a combination of housing, parks and a business knowledge hub, Hagastaden, is being built in both Stockholm and Solna. The new Karolinska Hospital will be operational in 2016. The first homes will be ready to move in to, by early 2017. Social Programme : There were good opportunities provided in the conference to network and enjoy pleasant moments with colleagues, friends and at the same time

share your own experience with international community. President IABSE, hosted the reception party on 19th September for all members present during the annual meetings. Chair of the Swedish group of IABSE hosted a reception on 20th September 2016, for those attending the annual meetings. Conference Ice-Breaker party was hosted by the organisers on 21st September 2016. A gala dinner was also organised on 22nd September, as is the norm in all IABSE conferences. These gettogethers and social events are traditions of IABSE conferences. Pic 8 shows a glimpse of some of these events.

Pic 8 : A Glimpse of Gala Dinner, Reception Parties, Ice-Breaking parties held during the Conference

For more Information, log on to : https://www.iabse.org/IABSE/Events/Stockholm_2016

The Bridge and Structural Engineer

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

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LIST OF ING-IABSE PUBLICATIONS AVAILABLE FOR SALE Sl. No. Name of the Publications

Price Rs

Postage Rs

Technical Presentation on â&#x20AC;&#x153;Maintenance and Rehabilitation of Bridge Structuresâ&#x20AC;? held at New Delhi in November 1994

200/-

100/-

Seminar Report on â&#x20AC;&#x153;Elevated Transport Corridorsâ&#x20AC;?

250/-

100/-

Hard Copy Out of Stock â&#x20AC;&#x201C; CD is available

held at Maysore (Karnataka) on 27th and 28th June 2014 Themes of the Seminar are as under:7HFKQLFDO 6HVVLRQ , 3ODQQLQJ 3UHSDUDWLRQ (FRQRPLF EHQHÂżWV 9DOXH (QJLQHHULQJ Technical Session - II Design and Construction Technical Session - III Operation and Maintenance 7HFKQLFDO 6HVVLRQ ,9 &DVH 6WXGLHV 3

36th IABSE Symposium on â&#x20AC;&#x153;Long Span Bridges and Roofs â&#x20AC;&#x201C; Development, Design and Implementationâ&#x20AC;?â&#x20AC;&#x201C; held at Kolkata from 24th to 27th September 2013

250/-

100/-

9RO 1R 6HSWHPEHU Special Issue â&#x20AC;&#x201C; Urban Flyovers (Structure, Architecture, Sustainability)

250/-

100/-

9RO 1R 'HFHPEHU Hard Copy Out of Stock â&#x20AC;&#x201C; CD is available Special Issue â&#x20AC;&#x201C; Bearings, Expansion Joints & STUs for Bridges (Selection, Design, Testing Installation, Maintenance)

250/-

100/-

9RO 1R 0DUFK ÂąSpecial Issue â&#x20AC;&#x201C; Building Structures

500/-

100/-

9RO 1R -XQH Special Issue â&#x20AC;&#x201C; Codes & Standards in Structural Engineeringâ&#x20AC;&#x201C;(Developments & Need for Improvement)

500/-

100/-

9RO 1R 6HSWHPEHU ÂąSpecial Issue â&#x20AC;&#x201C; Reinforced Soil Wallsâ&#x20AC;&#x201C;(Current Practices & Future Directions)

500/-

100/-

9RO 1R 'HFHPEHU Âą Special Issue â&#x20AC;&#x201C; Structural Failuresâ&#x20AC;&#x201C;(and Lessons Learnt)

500/-

100/-

9RO 1R 0DUFK Âą Special Issue â&#x20AC;&#x201C; Earthquake Resistant Design of Structures

500/-

100/-

9RO 1R -XQH Hard Copy Out of Stock â&#x20AC;&#x201C; CD is available Special Issue â&#x20AC;&#x201C; Strengthening, Repair & Rehabilitation of Structures

250/-

100/-

9RO 1R 6HSWHPEHU Âą Special Issue â&#x20AC;&#x201C; Aesthetics of Structures

500/-

100/-

9RO 1R 'HFHPEHU â&#x20AC;&#x201C;Special Issue â&#x20AC;&#x201C; Geotechniques & Foundations Design of Structures

500/-

100/-

9RO 1R 0DUFK ÂąSpecial Issue â&#x20AC;&#x201C; Enabling works, Formworks & Scaffolding Systems

500/-

100/-

9RO 1R -XQH 6SHFLDO ,VVXH Âą Steel & Composite Bridges Hard Copy Out of Stock â&#x20AC;&#x201C; CD is available

250/-

100/-

Note: These Publications are available on cash payment or through cheque drawn in favour of the "Secretary, Indian National Group of the IABSE, New Delhi" at the following address:

The Secretary, ,QGLDQ 1DWLRQDO *URXS RI WKH ,$%6( ,'$ %XLOGLQJ *URXQG )ORRU 5RRP 1R -DPQDJDU +RXVH Shahjahan Road, New Delhi -110011 Tel: 91-11-23388132, 23386724 E-mail: LQJLDEVH#ERO QHW LQ; LQJLDEVH#KRWPDLO FRP xiv

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

The Bridge and Structural Engineer

Design of tall piers for Railway Bridges in North-East India

Sumantra Sengupta

Amitabha Ghoshal

Principal Manager (Design) STUP Consultant Pvt. Ltd., Kolkata, India sumantra.sengupta@stupmail.com

Chief Advisor to Board of Directors STUP Consultants Pvt. Ltd. Kolkata, India gamitabha@yahoo.com

Mr. Sumantra Sengupta graduated in Civil Engineering from Jadavpur University in the year 1990 and post graduated in Structural engineering from the same institute in 1992. Since then he is working in STUP Consultants Pvt. Ltd., Kolkata. His field of expertise is Bridges, Special structures and Seismic analysis.

Mr. Amitabha Ghoshal graduated in Civil Engineering from Calcutta University in 1957. He has been Director and Vice-President of STUP Consultants Pvt. Ltd., Kolkata.

Summary

The paper deliberates on the design of tall piers of railway bridges in North-East part of India where the railway line passes through steep rolling hills of Patkai region, eastern trail of Himalaya, and as a result large number of tunnels and bridges need to be designed. The deep gorges are being crossed by tall railway bridges of maximum pier heights varying from 50 m to 140 m. The region falls under highest seismic zone area of India. Due to the varying height and considerably high piers in seismic prone area, multi modal analysis is performed using peak ground acceleration (PGA) and acceleration response spectrum obtained from site specific seismic vulnerability study with the available data. Due to the pier locations on the steep slope of the hill, rigorous slope stability analysis has been performed. Wind analysis has been performed followed by wind tunnel analysis.

Indian Railway intends to connect the capitals of the four North-East states, Manipur, Mizoram, Nagaland and Arunachal Pradesh with Assam by railway link. The work of Manipur and Mizoram has been started and the designs of 5 tall bridges in Manipur are already complete and 6 tall bridges in Mizoram are in progress. Construction at site is progressing in full swing and presently piers are being constructed using slip form technology. The length of railway line in Manipur is about 125 km and that of in Mizoram is about 60km. The alignments of the railway lines pass through steep rolling hills of Patkai region, eastern trail of Himalaya, and as a result large number of tunnels and bridges need to be designed. While the high mountains are penetrated by tunnel, the deep gorges between the mountain ridges are connected by tall bridges. The tallest of such bridges spans over a gorge at about 140 m above its bed level with an overall length about 700 m at rail level. With extensive study and discussion on possible alternative

Keywords: Tall bridges, Site specific seismic vulnerability study, Multi modal analysis, Pushover analysis, Wind tunnel test.

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Introduction

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span arrangement of the bridges, considering the parameters like the length of span, type of span, location of the piers and constructability, it was finally decided that main superstructures will be steel open web through type girders of span up to 103.5 m (c./c bearing). The piers are RCC hollow type with the tallest piers of 140 m height. Other piers on the slope of the hills vary from 50 m to 90 m height. The foundations are being designed with 1.5 m diameter piles that penetrate into rock layers with maximum length of 30 m. The critical issues of analysis and design involve preparation of site specific spectrum for seismic design of the bridge, rigorous slope stability analysis of the hill slopes on which the tall piers are standing, wind tunnel analysis to ascertain the actual behaviour of the structure in wind, fatigue analysis of superstructure with the latest provision of fatigue. Apart from IRS (Indian Railway Standard), other codes like IS (Indian Standard), IRC (Indian Road Congress), AREMA (American Railway Engineering and Maintenance-of-way Association), UIC (International Union of Railway) and Euro code provisions have been taken into account. The paper presents the steps followed for making these bridges as sustainable structures in a highly seismic zone at optimum cost.

Configuration of Bridges

Type of the superstructure and substructure, their construction material, span length, shape etc was finalized after much study with different types of material, span length, shape etc. A Technical Advisory Group (TAG) comprised of a team of expert was formed by railway to discuss on the final configuration of the bridge. Configuration of Superstructure: As the region is under high seismic zone, light superstructure was preferred, so that less force is drawn to the foundation during seismic activity. Thus only steel superstructure has been considered for alternate bridge configuration study. RCC or PSC (Pre-Stressed Concrete) superstructures, which are comparatively heavier, have been eliminated from the alternate bridge configuration study for the same reason.

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Among the different configurations of the tallest bridge, different options, like, simply supported through type open web girder, continuous superstructure, steel cantilever arch, balanced cantilever arch and cable stayed options were considered. The alternate options are shown in Plate – I. In order to keep the lateral deflection of the track with in the permissible limit the span length vs width of the deck becomes the guiding factor and for single track bridge the optimum span length was worked out. From this criteria large span structure like arch or cable stayed bridge configuration was ruled out. 103.5 m span open web through type steel girder superstructure has been chosen as a final superstructure option which gives an optimal solution for the superstructure configuration. Simply supported option has been preferred to continuous one as a choice by Indian Railways from easier maintenance point of view. Configuration of Substructure: The substructure was considered of three different types for alternate study – Steel trestle type, RCC hollow cylindrical and RCC hollow tapered. As the heights of the piers are very high, the horizontal deflection at the top of the pier is of considerable importance in view of the stability of the superstructure and functional requirement. It was decided that the absolute deflection of the pier should be retained within H/500 subject to maximum value of 300 mm for normal and wind load case. In seismic case the limit has been relaxed, but the deflection was found well within H/250 which is the recommended value for building structure laid in IS 1893. In light with the above, the different types of substructure was analysed and their response to various load cases studied. It was found that the steel substructure is yielding much more than the RCC substructure under wind load case and the value of the deflection crosses beyond the allowable value mentioned earlier. Accordingly steel substructure option was discarded. Considering aesthetics and ease of construction, cylindrical hollow piers are adopted in the final design. From the lateral deflection

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point of view also, the cylindrical hollow piers are preferable choice as the corresponding deflection in wind load under loaded chord condition is least of all the options. In Plate – II different configuration of substructure is shown and in Plate – III the deflection response of various types of substructure under different load cases is furnished.

Plate - I

All the bridges are designed for Broad Gauge single track carrying 25T maximum axle load. Spherical bearings have been adopted for all the structures.

The seismic restraint blocks are provided at both longitudinal direction and transverse direction to contain the dislodgment effect.

Plate II (Cont...)

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Plate - II

Plate - III

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For foundation, the number of piles and the distance between them under each foundation is estimated in such a way that there should not be any residual tension in the pile after considering their self weight. 1.5 m diameter bored cast in situ pile with varying length between 20 m (at river bed) and 30 m (at hill slope) has been adopted. In the Plate – IV the artist’s impression of the tallest bridge is shown.

Artist’s Impression of the Tallest Bridge Plate - IV

Critical Design Issues

Due to the location of the bridge in severe most seismic zone and the considerably large height of the bridge in the hilly gorges, there are several critical issues in connection with the design aspect of the substructure of the bridge. The issues are discussed hereafter. Construction material: As the bridges are located in the remote area of the country, and in spite of the vicinity of the national highway in some of the bridge sites, the general road condition is extremely poor, and high quality construction material cannot be envisaged. Preferable grade of concrete for such tall bridges, as under consideration, should be as high as possible for best performance of the bridges. However due to the remoteness of the area, it has proposed that M40 grade of concrete shall be considered for design and construction. Due to very narrow and winding nature of the approaches, the individual chord lengths which are to be transported to the site are also limited and accordingly the joints in superstructure are determined. Slope stability: As the area falls under severe seismic zone and the foundations are located on the steep slope of the hill, the slope stability is also a critical design issue. Under existing condition the slopes are stable, as they are in the present configuration for a considerable period of time. When the construction is being done on the hills by disturbing the ground profile, and additional load is

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being applied on the hills in the form of foundation supporting the substructure and superstructure, the stability of the slope is disturbed. Necessary analysis is performed using modern software, and final slope of the hill, by trimming or excavating soil or depositing boulder in steel wire mesh (Sausage) at the downhill, is suggested so the slope can be ensured as stable even under design seismic condition. From the existing condition of the stable slope it is found that the angle of internal friction shall be 45 degree with some minimum value of cohesion, considering equivalent soil effect of the rock. For the design purpose however the shear parameter has been adopted as 36 degree. In a very recently occurred EQ in the region, the same assumption was corroborated. The epicenter of the EQ was only 5km from the tallest bridge site. No slope failure has been noticed in the vicinity. The magnitude of the EQ was 6.7 and the hypocenter was 55km from the ground. The PGA map given by USGS site, as shown in Plate - V, shows that near the epicenter the maximum acceleration of the ground was 0.14 g. With the above data in hand an analysis has been performed to estimate the shear parameter of the bed material of the ground at the hill slope. As the bridges are under the construction stage, steep excavation profile for construction of the pile foundation exists all over the site, which didn’t fail in the EQ, and this suggests that the bed material has high shear parameter. It has been found that an equivalent cohesion of about 20 t/m^2 or equivalent angle of internal friction value of 45 degree along with cohesion value of 5 t/m^2 should be the shear parameter value in order to keep the slopes stable. This suggests that the adopted shear parameter for the design imparts adequate safety margin to the structure against stability.

USGS Record of PGA for Recent EQ in Manipur Plate – V

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Site Specific seismic vulnerability study: As the overall size of all the bridges are considerably large, tall piers and large overall span with varying height of piers on the slope of the hills, it was a necessity to understand the actual behavior of the bridges under seismic loading. Due to the non-uniformity of the pier heights and their large dimensions, it is preferable, and also suggested by the codes, that site specific seismic vulnerability study is conducted. The study was assigned to IIT Kharagpur and a new set of seismic parameters has been suggested in the study. Peak Ground Acceleration (PGA) and acceleration spectrum was proposed in the study for consideration in the design, and while the PGA suggested in the study was found to be very close to the value mentioned in the code, the acceleration spectrum gives a considerably higher value of the forces, if compared to that the code (IS-1893:2002) suggests. Seismic analysis: As the bridge is located in severe most seismic zone, and the shape of the bridge is irregular due to varying height of the piers, seismic analysis is the most critical issue of the design. As the piers are of varying height, their stiffness is different, and thus their response is not identical during seismic event. As the top of these piers are connected by superstructure the response of one pier will affect the response of the adjacent pier. The phase difference phenomena has not been considered in the design as the length of bridges do not call for such analysis. If the bridge length is more than 700 m, there is chance of ground movement of one foundation opposite to the ground movement of the farthest one, thus spatially varying, time history analysis would have been necessary. Rigorous multimodal analysis has been done using the design spectrum suggested in the site specific seismic hazard analysis. Wind analysis: Wind force causes a large deflection in the tall substructures at their top and thus steel superstructure was discarded from substructure option. Wind force also creates large deflection on the superstructure in horizontal plane, and from the stability aspect in wind load case, the distance between the trusses has been suitably adjusted. Wind force is also a very critical force during the cantilever erection of the superstructure and the same is proposed to be monitored during the erection stage. Stability of the superstructure and their effects on details of the pier caps: Due to the large deflection of the piers the stability aspect of the superstructure is a critical issue. The deflection of the tallest pier to the order of 300 mm needs special attention so 6

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that the superstructures can remain stable and the riding comfort of the passengers do not get lost. The horizontal and vertical deflection of the superstructure is controlled by providing adequate stiffness in the corresponding direction. Suitable seismic arresters of adequate height and strength have been proposed on the pier and abutment caps. The size of the pier caps are also generously provided (The minimum size of the pier caps are 14 m wide and 5.5 m long) so that the possibility of dislodgment of superstructure can be averted under severe most earth quake. Inspection & Maintenance of the structures, Illumination and Ventilation of the Pier shafts: The inspection and maintenance of the structures is one of the most critical issues in design. For the inspection of the condition of the outside of the tall piers, rail arrangements at top of pier shafts for hanging inspection trolley have been envisaged. For inspection of the insides, steel ladder shall be provided which will be supported on steel platforms cantilevered out from the inner face of the pier walls and on the intermediate hollow diaphragms placed about 20 m interval along the height. Doors at top and bottom of the pier shafts shall be provided for entry to the pier shafts. For illumination, electric power supply arrangement shall be provided. Forced ventilation arrangement has been considered during inspection operation. Adequate instrumentation schemes are implemented for monitoring the deflection of the members of the structures.

Geotechnical Aspects of the Sites

Geotechnical study has been undertaken thoroughly and for that purpose soil exploration has been done under every foundation location. At the tallest pier locations where the pile cap size is almost 32m x 32m with 81 numbers of piles in square grid, four numbers of borehole locations had been proposed. In general, it has been found that for all the bridge locations the top 3 to 10 m depth is comprised of stiff clayey silty sand with mix of gravels and below this level there is moderately to highly weathered shale/ sandstone. Refusal has been encountered in the standard penetration test at the shale/ sandstone layer. From the RQD (Rock Quality Designation) value it was found that in most of the cases core recovery of up to 50% was available in the sample but almost zero RQD is encountered (Refer Plate - VI) . During the estimation of pile load capacity, both vertical and horizontal, and in the slope stability analysis, the soil has been considered as granular with angle of internal friction value maximum at 36 degree.

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Rock Sample Collected during Soil Exploration from Pier P1 of the Tallest Bridge Plate - VI

Site Specific Seismic Vulnerability Study

A site specific seismic vulnerability study has been undertaken by IIT Kgp on behalf of the consultant. In the study the estimates of peak horizontal ground acceleration (PGA) and peak horizontal ground velocity (PGV) for all the five bridges in Manipur has been undertaken. The study also recommends the design acceleration spectrum to be considered for the design. Although penetration resistance or shear wave velocities were generally unavailable, qualitative information from the geotechnical investigation report indicates that the sites are likely to be of NEHRP class A or B. The study is based on earthquakes of magnitude 3.5 or more, that occurred between 1762 and 2011, with epicenters within a rectangular area, bounded by latitudes and longitudes over a distance of 350 km from the bridge sites. The deterministic approach was followed in assessing the seismic hazard. The required attenuation

relationships were developed, by first relating felt earthquake intensity to the logarithm of hypocentral distance (in kilometer) and exponent of moment magnitude information, and subsequently converting the relationships for estimating PGA and PGV, using relationships between felt intensities and instrumental PGA and PGV records developed for data from two recent northeast Indian earthquakes. The results indicate that, for site classes A and B (NEHRP), the maximum PGA for the bridge sites is expected to be 0.31 g and the maximum PGV is expected to be 20.81 cm/s, without considering topographic amplification. For bridge sites with valley slopes steeper than 30º, without any likelihood of earthquake-related slope instability, a topographic amplification factor of 1.4 on the PGA estimate has been suggested. Correspondingly, for flatter slopes, topographic amplification factor of 1.2 has been suggested. As all the hill slopes under present bridge sites are within 30º the finally suggested PGA as per the study is 0.31 x 1.2 = 0.372 which is marginally above the code suggested value 0.36 in the seismic zone V. For estimating the acceleration spectrum, nine free-field accelerograms, obtained from www. srongmotioncenter.org and www. knet.bosai.go.jp recorded at hard rock and stiff soil sites during earthquakes with MW between 7.0 and 7.7, with max ground acceleration between 280 cm/sec2 to 315 cm/sec2, and max velocity between 20 m/ sec to 30 m/sec, which is the present site specific values, the response spectra were computed. The earthquakes used in this assessment were of the strike slip and thrust strike-slip type mechanisms, similar to what is expected in the geographic region under consideration.

Plate - VII : Site Specific Acceleration Spectrum

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In the Plate – VII, all the nine free field accelerograms and their mean and mean plus one standard deviation curve, has been plotted. The suggested design acceleration spectrum is the mean plus one standard deviation curve as shown in the red firm line. The above spectrum, when compared with the code specific spectrum as shown in the same Plates, shows that the values of the spectral acceleration are at some points more than two times higher.

Seismic Analysis

Multi modal analysis has been performed, using the acceleration spectrum suggested in the site specific seismic vulnerability study. The analysis is done for the whole model of the bridge including all the piers, pile foundation and superstructure. Stick model analysis of each pier has also been performed. The critical most results among the above two analyses is considered for design. Moreover as suggested by the code (IS:1893:2002), the gross seismic horizontal coefficient obtained from the modal analysis is compared with that for first fundamental mode of vibration, and the design forces are proportionately increased if the horizontal coefficient corresponding to first fundamental frequency is found higher. The analysis is performed for both longitudinal seismic and transverse seismic. The Peak Ground Acceleration (PGA) and the acceleration spectrum considered in the design are as furnished in the site specific seismic vulnerability study section. As the structure is comprised of uniform diameter pier with discretely varying thickness the mass distribution along the height of the piers are non uniform and due to this, in most of the cases the participation of higher modes are also predominant. It has been found in general that for tall piers the higher modes are predominant and for short piers first fundamental mode is predominant. As a result the

gross horizontal seismic coefficient from the modal analysis comes out to be higher than that of for first fundamental mode with 100% mass participation for the tall piers, where as for short piers it is the reverse case. During modeling of the structures for behavior in horizontal load (seismic and wind) the piles are also considered. The piles are introduced in the structure over a length up to their depth of fixity estimated as per pile code (IS 2911), and at the bottom of piles, fixed support has been considered. In the Plate – VIII the first two fundamental modes of vibration in transverse direction is shown. The structures are initially analysed without taking into consideration the effect of response reduction factor (R). After obtaining the moments at various sections of the structures, appropriate value of R is applied in order to get the design moments. As per the IITK-RDSO guideline the value of R has been considered as 1 at the junction of pier and pile cap, 1.5 for pile foundation and pier sections, 2.5 for pier cap. For stability of the structure in seismic case, ie while estimating the minimum vertical forces in pile in seismic, R has been considered as 1.5. The higher moments at the junction of pier and pile cap due to the consideration of value of R as 1 is catered by providing additional thickness at the bottom of pier by flaring the outer diameter over certain length depending on the height of the pier. While designing the piers and abutments in seismic, it was found that for the shorter piers and the abutments, due to their higher stiffness compared to the taller piers, the accelerations of the structures (Sa/g) obtained from the spectrum suggested in site specific study reach to their highest value which is 4.2 (The same maximum value in the spectrum given in IS 2911 is 2.5) and as a result the moments in the sections of these structures become considerably high. The horizontal seismic coefficient which is estimated from the formula

First Two Fundamental Mode Shapes of the Structure in Transverse Direction Plate - VIII

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Z/2.Sa/g.I becomes 0.372/2x4.2x1.5, ie 1.17 which means that the horizontal force for the shorter structures in seismic is more than one times g, ie more than their vertical effect. Thus for the shorter piers the diameters and thicknesses of the walls are comparatively higher than that of for taller piers. In case of abutments the detailing is done in such a way that except the dirt wall and the bearings, entire part of the structures remain below the ground level and thus the structures can be considered as part of the ground with no self flexibility. Under this condition the acceleration of the structure can be considered as same as that of the ground, ie the structures will accelerate with peak ground acceleration (PGA). Accordingly the horizontal acceleration coefficient for abutment structures has been considered as 0.372/2 x 1.5 = 0.279 which is 4.2 times less than that of for shorter piers. In designing the tall piers, second order theory for estimating additional moment due to slenderness is supposed to be considered. In the present cases due to the effect of high seismic force, as generated by the acceleration spectrum suggested in site specific study, none of the piers are having slenderness ratio more than 50, and thus the second order theory was not applied.

Pushover analysis

Pushover analysis has been performed on the designed RCC hollow pier section at the junction of pier and pile cap (Refer Plate – IX) in order to estimate the ductility performance of the section.

Plate - X : Moment vs Curvature Graph for the Tallest Pier Section

The Moment vs curvature (φ.D) graph has been plotted in Matlab (Refer Plate – X) using the stress-strain relationship of concrete section with specified stirrup as given in “Reinforced concrete structure” by Park and Paulay and the standard stress-strain relationship of reinforcement steel. From the Moment vs curvature graph, it is found that the curvature ductility of the section is 14. Deflection ductility of the section can be worked out from this graph and the value has been found 4.6. The over strength factor has been found 1.49. Corresponding product of displacement ductility and over strength factor is 6.9. From the above study it can be concluded that response reduction factor (R) may be considered as 6.9/2 = 3.45 (taking in to consideration the conversion factor 2 between Maximum Considered EQ and Design Basis EQ). However in the present

Plate - IX : RCC Hollow Pier Section of the Tallest Pier

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case the section has been designed with R = 1 as was recommended in the IITK-RDSO guideline, 2010.

8. Wind Analysis Wind force is estimated from the wind load code IS 875 (Part 3). The basic wind speed is considered in the zone as 50 m/sec. The probability factor or risk coefficient (k1) and the topography factor (k2) for all the bridges under consideration are taken as 1.08 and 1.0 respectively. The terrain, height and structure size factor (k3) varies with the height of the structures and the maximum value of the same in these bridges has been found to be 1.24. With the above factors it has been found that for the tallest bridge the maximum wind pressure comes out to be about 270 kg/sqm for the superstructure. For piers, depending on the heights, the pressure will reduce. For the smaller piers in any bridge the height factor k3 is estimated by considering the height of the piers measured from the minimum bed level.

and across-wind responses of the Piers of the bridge, under simulated flow conditions, with and without superstructures. Aero-elastic models of the prototype Piers were designed and fabricated at a model scale of 1:275. The superstructures have also been designed and fabricated at the same model scale (1:275).

Aero-elastic Model of Bridge No. 164 Plate - XII

In the Plate – XI the deflection response of the tallest bridge in transverse wind case is shown.

Deflection Response of the Tallest Bridge in Transverse Wind Plate - XI

From the analysis, while deflection of top of piers were checked against the allowable value, it was found that higher depths of pile caps were becoming necessary in order to make the structures stiffer and thus to keep the deflections within permissible value. It was found that instead of increasing the depth of pile caps in their entirety, if the central part of it containing the pier base is being increased the effect is becoming same. Based on this consideration, instead of increasing the pile cap depth, pedestals below the pier base at top of the pile cap has been introduced.

Instrumented Pier of Aero-elastic Model of bridge No. 164 Plate - XIII

9. Wind Tunnel Test The theoretical response as estimated above has been compared with the actual response of the structures in wind by undertaking the wind tunnel test in the National wind tunnel facility at IIT Kanpur. The wind tunnel studies carried out on a scaled down aeroelastic model (Refer Plate – XII). The main objective of the study was to estimate the maximum along-wind 10

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Accelerometer at the Top of the Pier at Two Orthogonal Directions Plate - XIV

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Table-1 : Table Showing Comparative Values of the Analytical Results and Test Results for the Tallest Pier BM BM (Model) Shear (Analytical) (Analytical)

Location

At base

98,373 tm

93,500 tm

985 t

830 t

At mid height

39,312 tm

36,000 tm

680 t

600 t

At top

156 mm

140 mm

The Pier models were instrumented with strain gages at four different levels (Refer Plate – XIII) to obtain the along-wind and across-wind bending moments and shear forces. In addition the pier models are also instrumented with accelerometers in two major orthogonal directions (Refer Plate – XIV) to obtain along wind and across wind tip accelerations and dynamic deflections. The bending moments and shear forces at various levels and for different wind incidence angles (varying between -90º to +90º @ 30º interval) have been obtained on the Pier models. Out of seven piers, the tallest pier P3 and the second tallest pier P2 were selected for the study. The test results and the analytical results have been compared in tabular form as shown in Table - 1. For Pier P3, the estimated first and second mode fundamental frequencies are 0.490Hz and 2.712Hz, respectively for ‘without live load case’ and they are 0.359Hz and 2.012Hz for ‘with live load case’.

Shear (Model)

Deflection Deflection (Analytical) (Model)

Sl. No.

The critical wind speed at which this pier could get excited because of vortex shedding is 39.2 m/s, for the first mode and 216.9 m/s for the second mode, for without live load cases and these speeds are 28.7 m/s and 160.9 m/s for the pier “with live load”. As the maximum wind speed is 69 m/sec there is no chance of second mode of vortex shedding to generate. The first mode vortex shedding occurs at 39.2 m/sec velocity, corresponding to which the transverse moment is maximum. Longitudinal direction maximum moment occurs at maximum wind speed, and thus, the two critical moments in the two orthogonal directions does not occur simultaneously. The maximum BM due to first mode vortex shedding is to the tune of 8000tm as obtained from the model test.

10. Design Details of the Bridges Some of the design details of the tall piers and their foundations are shown in Table – 2.

Table-2 : Table Showing Design Details of the Substructure and Foundation Height of pier (m)

Outer diameter of the hollow pier (m)

Wall thickness of pier at base (mm)

No. of pile below pier

Size of pile cap (m x m)

Thickness of pile cap (m)

141

1000

9x9

31.8 x 31.8

4.2

118

1000

9x9

31.8 x 31.8

4.0

100

750

7x7

24.3 x 24.3

3.5

750

7x7

24.3 x 24.3

3.0

500

6x6

20.6 x 20.6

2.5

500

5x5

16.8 x 16.8

2.25

7.5

500

6x6

20.6 x 20.6

2.25

700

5x5

16.8 x 16.8

2.25

11. Conclusion The design of the bridges is critical in many respects due to the undulated terrain, tall and varying height of

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the piers, large length of the bridges, position of the piers on the hill slopes, the presence of severe-most earthquake zone and the high response acceleration

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spectrum of the particular site, as obtained from the site specific seismic vulnerability study. Tension bearings for transferring high seismic forces generated from the superstructure to the substructure, and robust seismic arresters to prevent dislodgement of the superstructure during seismic, special features of the design. The construction methodology of the bridges, transportation of the materials particularly in view of the winding approaches, the limitation of the size of the fabricated steel chord of the superstructure and erection of the superstructure, affect the design aspect. All the above issues have been taken in to consideration with due importance to each issue and an effort has been made to achieve safe and sound structures with optimum cost.

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12. Acknowledgement The authors acknowledge the contribution made by various agencies that include ●

NF Railway Engineers

●

TAG formed with acknowledged experts

●

STUP Consultants Pvt. Ltd.

●

IIT, Guwahati who was responsible for proof checking

●

IIT, Kharagpur for site specific spectrum generation

●

IIT, Kanpur for Aerodynamic model analysis.

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High-Rise Buildings in India - Code Improvement Required

Vipul Ahuja

B.M. Ahuja

Director & CEO, Ahuja Consultants Pvt. Ltd., Lic Structural Engr. California

Managing Director, Ahuja Consultants Pvt. Ltd., & Ex. HOD Civil Engr. IIT Delhi

Vipul Ahuja received his M.S. (1982) degree from from university of Oklahoma USA, B. Tech. (1979), I.I.T. (Delhi), Licensed Structural & Civil Engineer Since 1988 (California, USA). He has vast experience in the field of structural/ earthquake engineering. His most recent experience include state-of-the-art technology implementation including base-isolation and damper application in high-rise buildings (Steel Composite & R.C.C.). He has designed hundreds of buildings of residential, commercial and retail spaces and long span roofs, bridges, shell, prestress concrete structures (Pretension and Post-tension) over his professional life both in USA (14 years) and India (25 years). He has experience using several current international codes including ASCE, IBC, AISC, ACI, FEMA, EN, Australian and Japanese standards and has several technical papers to his credit.

Prof. B.M. Ahuja, received his civil engineering Ph.D(1961). degree from the Imperial College, London, D.I.C. (London), B.Sc. (Engg.). Hons (London), MICE (UK) Chartered Engineer (UK). He began his career in Service contractors (one year), Consulting Engg (2 years), Central power & water commission (11 years), and he joined I.I.T. Delhi as a Head of Civil Engineer in 1961 and retired in 1986, He started consulting engineering consultancy in 1986 and his firm converted to a Pvt. Ltd. firm in 1991.He has done many challenging projects like irrigation design (Dams, barrages, hydraulic gates& valves), shell structures, Indoor sport complexes, high-rise buildings etc. His ex-students include many luminaries in the world of structural engineering today (including several leading consultants in Delhi-NCR and professors of I.I.Ts) and other walks of life.

Summary Currently IS 456 Code of practice for Concrete Structure design is vintage 2000 & of course IS 1893 (Loads for Earthquake Design) was last published in 2002, though being occasionally updated for minor corrections. The world today is moving at a rapid pace. For Indian industry to keep pace with world practice must incorporate at least those provisions which have acceptance in worldwide accepted codes such as those of ACI, Europe, China, Japan & New Zealand. Incorporating results of new research is a far cry. For example ASCE-07 Commentaries state that base isolation is a rapidly changing technology & therefore latest codes must be adopted as soon as possible.

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These codes mentioned above are quite advanced as compared with the IS Codes for high-rise building technology especially ACI, which publishes a new code every three years. The Code writing authorities in India may not realize but stalling this process with a 14 year plus cycle is causing the society great losses, directly or indirectly. Losses could be in terms of safety or financial. For example some provisions not in Indian codes may lead to unsafe situations & in future will result in being candidates for retrofit & cause loss to life & limb as well as financial loss. They may also cause extra expenses on account of being overly conservative. Topics such as floor diaphragms, shear walls, slender columns, flat slabs, base isolation, treatment of multiple Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

towers sharing a common podium or basement, leave much to be desired in Indian codes. Draft codes also are still not adequately covering these topics, though the final form is yet to be assessed. The paper covers several of these topics while recommending changes.

Introduction

High-rise building analysis & design procedures may sound similar to low or even mid-rise building but there are subtle differences. And if these differences are not recognized by the designers, major shortcomings may emerge which may lead to unsafe conditions or not achieve acceptable performance levels. Many countries have special sections within the code tailored to meet these special requirements or even special codes that address these issues. Currently BIS does have a sub-committee that is looking into this matter, however the code formulation time is required to be phenomenally speeded up as incorrect designs being done every day. Some day when these come to light in a design audit a retrofit may be required, or worse yet in an earthquake, will result in heavy losses to life & limb in addition of course to financial losses.

forces though the true behaviour is not that & leads to huge wastage. At the other end of the spectrum some engineers consider only the dimension of one tower and this could be a non-conservative assumption. Whereas the solution may lie somewhere in between. These assumptions require to be standardized. The transfer forces between tower & connecting diaphragm including the concept of chords & collectors also requires to be incorporated in the code. An alternate method of evaluating the time period of buildings with a common basement has been proposed in a draft of IS1893. However it is again an empirical procedure and without putting out a logic among code users and therefore does not evoke confidence. As a rule codes must be published along with commentaries to be readily accepted. Following approach is suggested: For the case where lateral stiffness of the tower is less than 10% of the podium then the ASCE approach shown in the later paragraphs of this paper should be followed else the following applies. a)

For a tower with podium attached to it, a comparative analysis shall be done of the tower with the podium modeled & without it. If the mass participation or time period of first three modes changes by more than 20% (open to discussion) of the higher value, the podium has significant effect on the behavior of the structure.

For significant interaction case bottom of the podium level (i.e. podium dimension) shall be used for ascertaining the controlling dimension of the building & the empirical period. The period so calculated shall be compared with the natural period of the whole model including the podium & the shorter of the two periods adopted. In either case the base shear shall be scaled at the base of the podium level (i.e. the level of determining the empirical period).

For insignificant interaction case, the tower height shall be considered from the base of the podium but width at the top of podium (i.e. tower dimension) for calculating empirical period. In this case too this period should be compared with the natural period & the shorter of the two periods adopted.

For non-uniform podiums along building height, such as reducing podium plan dimension with height, the height of the tower shall be

In this context provisions of ASCE 07-10 & ACI 318 have been compared with the IS 1893-2002 & IS 1893-2016 (draft), IS 13920 & IS 456, with selected references to the New Zealand, Japanese & Chinese codes.

Deficiencies in Current Code

2.1 Empirical Time Period Clause for Several Towers Connected with a Common Podium For the purpose of this discussion, a podium is the connecting structure among towers (or around a single tower) above the ground floor (not laterally restrained but moving with the tower) while basement is the common connecting structure below the towers (or surrounding a single tower). The basement is commonly bound by a retaining wall on all sides which restrains the lateral movement at the ground level. When a common podium connects various towers or surround a single tower above ground there is interaction among them (see Fig. 1) Code requires the maximum width of tower to be considered in its empirical formula to calculate minimum lateral force for shear wall buildings (& most highrise buildings being built today have shear walls). Though conservative engineers may consider the whole width of podium as tower width and subject to huge seismic 14

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considered above such level where the 20% of mass participation/time period does occur.

Where several Towers are connected with a common podium the empirical procedure suggested above for individual towers with podium shall be compared with a the natural period for a combined model. The combined model shall be examined for the mode with highest period of the single tower in question. This period shall be compared with the individual empirical period of tower & lower period adopted.

As an example a building with a flexible diaphragm having shear walls in some grids & moment frames in others. This kind of building is not a dual system. When this happens the most stringent applicable structural system limitations and its design requirements will apply. This is required & allowed as per ASCE-07.

Different Framing Systems (R values) along Vertical direction

Where the lower system has a lower Response Reduction Factor (R), R for the upper system may be used to calculate the forces and drifts of the upper system. For the design of the lower system, the R for the lower system shall be used. Forces transferred from the upper system to the lower system shall be increased by multiplying by the ratio of the higher R to the lower R.

ii)

Where the upper system has a lower R, this lower R shall be used for both systems.

Fig. 1 : Buildings having a Common Basement under Construction (no Expansion Joint below Ground)

Combinations of Framing Systems (R values) in the Different Direction

Different seismic force-resisting systems (with different “R” values) may be used to resist seismic forces along each of the two orthogonal axes of the structure. The respective R, coefficients shall apply to each system. In fact the retrofit codes like ASCE 41 have a member based reduction factor (“m” factors).

Combination of Framing Systems (R values) in the Same Direction

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iii) Two Stage Analysis Procedure

b) The period of the entire structure shall not be greater than 1.1 times the period of the upper portion considered as a separate structure supported at the transition from the upper to the lower portion.

d) The lower portion shall be designed as a separate structure using the appropriate values of R. The reaction from the upper portion shall be those determined from the analysis of the upper portion amplified by the ratio of the R of the upper portion over R of the lower portion. This ratio shall not be less than 1.0.

2.2 Combinations of Framing System in Different/ Same/Vertical Direction As per IS codes this is not allowed-the building must have the same “R” value in each direction. However the logic not to allow it, is weak.

A two-stage equivalent lateral force procedure is permitted to be used for structures having a flexible upper portion above a rigid lower portion, provided the design of the structure complies with all of the following: The stiffness of the lower portion shall be at least 10 times the stiffness of the upper portion.

The upper portion shall be designed as a separate structure using the appropriate values of R.

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The upper portion is analyzed with the equivalent lateral force or modal response spectrum procedure, and the lower portion is analyzed with the equivalent lateral force procedure.

Concrete Shear Wall Design Procedure to be Updated to Capacity Based & Boundary Element Design Revamp Required

Currently the approach for shear wall design is empirical & “wasteful”. Where vertical stresses are above a threshold (0.2fck) boundary element needs to be provided and discontinued when it falls below 0.15 fck see Fig. 2 for a case where boundary elements are being provided at great heights. In the new (ACI 31814) approach, a non-linear hinge is forced to be formed at the lowest story free to translate, such as ground floor. Normally if this hinge formation is enforced, ductile detailing at other places along the height is not necessary or greatly reduced. Having stated that it should be noted that the wall in the upper floors apart from the floors having the hinge must have adequate over strength so as to be emphatically stronger than the intended hinge location.

Fig. 2 : Shear Walls Continue to be Provided with Boundary Elements (in Circles) at Great Heights where no Possibility of Hinging Exists (with a Little Bit Extra Vertical Steel) Code Writing Authorities to Take Note of the Colossal Waste.

Also the detailing of boundary element is woefully inadequate. By any logic, the area of stirrups in the two directions should be different (see Fig. 3 & 4-ACI 318-2014 extract “Ash1” & “Ash2”). however IS code gives the same number each way leading to conservative/un-conservative design detail. Further the length of boundary element is not based on strain compatibility. These issues need to be sorted out. 16

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Fig. R18-7.5.2 Example of Transverse Reinforcement in Columns Fig. 3 : ACI 318 Extract of Hoop Reinforcement Requirement

Fig. 4 : Table 18.7.5.4 Extracted from ACI 318-2014 for Hoop Area- Directional Bias

Coupling Beams Simplified

Detailing

Coupling beams detailing as shown in IS 13920 is a nightmare to implement - especially for narrow width beams (less than 300 mm wide). Having diagonals in both directions and stirrups around each diagonal it becomes impractical to accommodate bars & pour concrete. ACI now has option not to provide stirrups around diagonals, making the placement much simpler. See Fig. 4.

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Fig. 4 : Shows New ACI 318-2014 Coupling Beam without Stirrups Around Diagonal Bars

Provisions for Large Lightly Reinforced walls Required

As per EN 1998-1 Sec 5.4.2.5 there are special provisions for large light reinforced walls. Incorporating these or similar provisions will lead to significantly less steel consumption & save national waste, not to mention the green footprint of the building.

Relative Depth of Beams to Column Size

A new requirement based on lessons Earthquakes & experimentation of ACI is that the column lateral dimension shall be more than half the size of beam depth (See Fig. 5). This ensures proper transfer of forces in the joint based on the strut and tie approach. With smaller ratios stresses within the joint zones become excessive & brittle failure results. If not addressed this can result in countless retrofits in future. Further the strength of the columns (above & below a joint) must be at least 20% more than beams at the same joint.

zonation studies for densely populated metropolitan cities & suburbs many of which have already been done, but lying waste. Each structure would be assigned a “design category” based on several parameters including expected peak ground acceleration based on short & long periods as well as soil type, importance factor etc. For example currently a building with high importance factor in Zone 2 need not be designed for ductile detailing, but then depending on the design category the rule may dictate ductile detailing. So rules would apply to the design category rather than the zone.

Phasing Out of Zone Concept

Mapped Response spectrum parameters already approved by NDMA should be incorporated in code. This should replace the antiquated “Zone” based approach & pave the way for a more deterministic approach. This will also logically require microThe Bridge and Structural Engineer

Fig. 5 : Steep Strut in Joint as Column Dimension Reduces (Illustration from Structure Magazine May 2016, Dr. S.K. Ghosh)

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Reverse Shear (Back-Stay Effect)

The concept of reverse shear can be best illustrated with a cantilever; the fixity being provided by a couple of supports spaced much closer than the overhang. A high rise building with a stiff basement around it may be viewed through this prism. When the overturning moment is high the resisting couple forces (also called reverse shear or backstay force) can be several times greater than the base shear (due to wind or EQ forces). So if one would imagine designing shear walls below ground level to several times the base shear (five to six times amplification is common). However if the in-plane stiffness of the ground floor slab is a finite number (and not infinite as would be provided by an idealized lateral support) then the reverse shear is significantly reduced. This assumption of lateral stiffness magnitude or its process of evaluation is currently unregulated by the codes. This can result in large variations in design of shear walls from being utterly wasteful to being under-designed. Further when this reverse shear goes through the ground floor diaphragm the same must be designed for those forces. Currently there are no flags in the code for this and often diaphragms are ignored for these nature of forces. The same is true for peripheral retaining walls which need to be designed as shear walls depending on the level of fixity assumed at the ground floor as one credible load path must exist. The issue gets even more complex with multiple towers sharing a common basement. Some codes insist that all towers be modeled together with the common basement, unless an expansion joint separates the tower from the basement entirely.

10. Piled Rafts Frequently the structural engineer is faced with low soil bearing capacities for highrise buildings. The obvious, though expensive option available is piles. Piling is also a time taking activity that requires a specialized contractor. An option that has emerged for keeping foundations in the economical range is the piled raft see Fig. 6. The piles are allowed to yield (geotechnically) & then the raft & piles act together in tandem. This also reduces the settlements, brings down costs & construction time (since far fewer & shallower piles are required). However the issue of Soil FEM modeling is even more important in piled raft situations. The method of analysis & assumptions made throws up different results with different set of assumptions. There is no regulation on this and must be included in the code.

Modulus of Sub-Grade Reaction for Raft Foundations

These days the modulus of subgrade reaction is being supplied by the geotechnical consultant based on the premise that under total gravity load 75 mm settlement is experienced by the raft. If structural engineers see something other than 75 mm settlement in analysis they begin to question the geotechnical engineer. They are not aware that this settlement calculation includes the factor of safety. As a result this value varies across the spectrum of consultants. The higher the modulus the greater the economy. This reflects the state of current practice. 18

In any case for tall buildings the method of modeling the soil in FEM is advisable & soil stiffness at various strata should be evaluated & used for analysis. These settlements give a far more realistic picture than the archaic modulus of sub-grade reaction. Sometimes this becomes an iterative procedure between the geotechnical & structural engineers as the solution must converge to a common settlement by both. Code should mandate beyond what height of building, FEM of soil is mandatory & outline important benchmarks in the study.

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Fig. 6 : Piled raft Showing Sparsely Spaced Piles (Red Circles)

11. Slender Columns/Shear Walls ACI has in 2011 made a radical change in design of slender columns has limited magnified moment to

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1.4 times the moments on account of other sources (ACI-318, 2014, Sec 6.2.6). This severely impacts sizes of columns & virtually requires thicker columns than used earlier. This has obviously been included in ACI after significant research. Ignoring such scientific information is putting the whole society at risk.

The definition of wall & column is made clear based on aspect ratios & height to thickness ratios. IS code has no guidelines to the effect. High-rise buildings are being built on 200 mm thick walls with plenty of honeycombing due to rebar congestion. Some day the society will pay for retrofitting of such structures at a huge price. See Fig. 7 for ACI 318 requirements.

Fig. 7 : Extract of Table r18.10.1 of ACI 318-14 Showing aspect Ratios Defining Walls & Columns

12. Precast Building Provisions

14. Conclusion

Precast concrete is increasingly finding its hold in the Indian high-rise landscape. There are many concepts from ACI 318 that are very useful to incorporate for indian conditions. The definition of strong connection, extra wider beam as compared with column & location of even mechanical splices outside the hinging zone for ductile frame to name a few. The strain being concentrated in the connection zone is a worry. The mechanical couplers must be type 2 and be able to withstand a force equal to the tensile strength of the bar and splices must resist 1.5 times the yield strength. Precast walls also require special attention.

Much work remains to be done in upgrading the current IS codes to make them ready for practitioners who are at this time following one or the other international codes or none at all for topics not adequately covered. As such there is no licensing system for structural engineers today, even codes are woefully wanting. This change is required to be done on a war footing. That includes licensing of structural engineers as well.

References 1.

A Comparative Study of Code Provisions for Ductile RC Frame Buildings Yogendra Singh & Vijay Namdev Khose Indian Institute of Technology Roorkee, Roorkee-247667, India Dominik H. Lang Norsar, P.O. Box 53, 2027 Kjeller, Norway.

ACI 318M-014. (2014). Building Code Requirements for Structural Concrete (ACI 318M-08) and Commentary, American Concrete Institute, Farmington Hills, MI 48331, U.S.A.

ASCE 7-10. (2010). Minimum Design Loads for Buildings and other Structures, American Society of Civil Engineers, Virginia, USA.

EN1992. (2004). Eurocode 2: Design of Concrete Structures, European Committee for Standardization (CEN), Brussels, Belgium.

13. Redundancy Redundancy is expressly mentioned in various international codes. This has translated into several specific codal requirements. Additional bottom reinforcement to be provided at column drop - this prevents sudden collapse as bottom bars start acting in a catenary form. Minimum continuous steel to be provided in all beams so even if a column fails (like in a bomb blast) the structure does not collapse. Push down analysis to be performed in some cases (similar to pushover analysis-non-linear static procedure). There are several others not possible to be listed here but the point is this should be introduced in the IS code as well.

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EN1998-1. (2004). Eurocode 8: Design for Structures for Earthquake Resistance, Part 1: General Rules, Seismic Actions and Rules for Buildings, European Committee for Standardization (CEN), Brussels, Belgium).

IS 456. (2000). Plain Reinforced ConcreteCode of Practice, Bureau of Indian Standards, New Delhi, India.

IS 1893 Part-1. (2002), Part 1: Criteria for Earthquake Resistant Design of Structures General Provisions and Buildings, Bureau of Indian Standards, New Delhi, India.

IS 13920. (2016). Ductile Detailing of Reinforced Concrete Structures Subjected to Seismic Forces-Code of Practice, Bureau of Indian Standards, New Delhi, India.

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NZS 1170.5. (2004). Structural Design Action Part 5: Earthquake Actions New Zealand, Standards New Zealand, Wellington 6020.

10. National Standard of the People’s Republic of China (NSPRC) (2010), Chinese Code for Seismic Design of Buildings (GB50011-2010), Ministry of Construction of Peoples Republic of China, Beijing, China. (In Chinese). 11. Architectural Institute of Japan Published Codes (Various). 12. Development of Probabilistic Seismic Hazard Map of India, Technical Report of the Working Committee of Experts (WCE) Constituted by the National Disaster Management Authority Govt. of India, New Delhi. Approved by NDMA on 28th August, 2008 (Vide No. 3-7/2008/PMU).

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CHALLENGES INVOLVED IN DESIGN AND CONSTRUCTION OF 275M TALL RCC CHIMNEYS Er. Gupta, is an active member of various Codes and Standards Committees of BIS and IRC. He has been honored with awards for the best papers titled “Seismic Design and Construction of Radisson Hotel, New Delhi” and “Launching Systems for Segmental Bridges” by IBC and Prestressed Concrete Design Award 2013 by IEI. He has been lecturing as Guest Faculty in NITHE, CRRI, CIDC, ISDA, DPC etc. Er. Gupta is the Vice President of ICI.

Vinay GUPTA Chief Executive Officer, Tandon Consultants Pvt. Ltd., New Delhi 110014, India Vinay.Gupta@tcpl.com

Summary

Height of chimneys has been increasing with time, present usual height being 275 m, for safety against existing harmful gases from Thermal Power Plants. Tall chimneys pose the problems of designs & constructions. Additional considerations of dynamic wind, longitudinal & transverse wind effects, oval ling etc. need to be accounted for, adequately in the design. On the other hands, construction demands specialized equipments, such as slip forming or jump form. Extreme care is required to ensure straightness and verticality of the chimney during slip forming process. At time, it has been found in the case of tapered chimneys that it is not feasible to reduce the dia further after reaching certain height, say 2/3rd height. Such aspects need to be adequately planned in advance to avoid design difficulties and construction problems. These days, multi-flue chimneys are more common to serve multiple boilers of power plants. The flues may be in brick or steel (more commonly used). Insulation is provided around the flues to reduce the temperature between chimney shell and flues for human comfort and to reduce thermal gradient in chimney shell.

India has been striving to alleviate the electric power crisis, recently aggravated due to the economic boom in the country. Out of the two major sources of power, i.e. Hydel Power and Thermal Power, the latter has become more popular due to its adaptability towards larger production capability. Thermal power is obtained through burning coal, which is required to operate the steam boilers. When burnt, the coal produces polluting gases that need to be discharged at an elevation high enough to dilute the pollution and to keep it within acceptable limits at ground level. An adequately designed tall chimney serves this purpose. As the pollution norms have become stringent with time, the chimney heights have gone up progressively from 100m to 150m to 220m to 275m. This is represented well in the chimneys designed over the years by the authors (If this sentence is added, Prof. Mahesh Tandon should also become one of the authors), Fig. 1 In most thermal power plants, 275 m tall concrete chimneys have now become the standard norm. It may be worthwhile mentioning here that a bi-product of burning of coal is fly ash, which is produced in the process line between boiler and chimney. This fly ash is extracted using electrostatic precipitators, which incidentally can be used in blended cement as mineral admixture in concrete.

Keywords: Tall Stacks, Flue, Strakes, Dynamic Analysis, Across Wind, Thermal Stresses.

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Introduction

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Fig. 1 : History of Chimneys

Single V/s Multi Flue Chimneys

There are broadly two types of chimneys, namely (i) Single Flue Chimney and (ii) Multi Flue Chimney. Single flue chimney serves one boiler of a thermal power plant. The 100 m tall chimney for ATV at Mathura and the 130 m tall Sika Thermal

Fig. 2 : Picture of ATV Chimney at Mathura

Power Plant are good examples of the same, Figs. 2 & 3. Single flue chimneys are also used for pollution control in Cement Plants and in Oil Refineries. Single flue chimney has one brick lining flue housed inside a reinforced concrete shell. Figs. 4 & 5 depict a typical single flue chimney of 100 m height.

Fig. 3 : Picture of GSEB Chimney at Sika, Gujrat

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Fig. 4 : Single Flue Chimney : Elevation

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Multi Flue chimney, on the other hand, has more than one flue (2, 3, 4 or even 5 nos.) housed inside the concrete shell. The captive Power Plant for Vedanta Aluminum Ltd, Jharsuguda, Orissa has 275 tall 4 flue and 5 flue chimneys, Fig. 1. The flues may be in brick or steel. Multi flue chimney serves more than one boiler, provided for multi unit power plants, which is common in case of high capacity power plants, often in the range of 250MW to 1500 MW. Figs. 6 & 7 depict a multi flue chimney. A chimney shell can either be flat, vertical or tapered (more common) or parabolic. The chimney shell is often referred to as the “Wind Shield”.

Fig. 5 : Single Flue Chimney: Plan at Duct Entry Level

Fig. 6 : Multi Flue Chimney: Elevation

Fig. 7 : Multi Flue Chimney: Plan at Duct Entry Level

Structural Analysis of Chimney

3.1 Global and Local Loads The structural analysis of the chimney involves the determination of: (a) Global Loads (b) Local Loads The Global Loads are catered to by considering the chimney structure as a whole as a vertical cantilever fixed at the base. The main affects are (i) bending moment and shear due to wind/earthquake acting in the horizontal direction and (ii) the vertical load due to the self weight of the structure and superimposed loads supported by it.

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Local effects on the chimney shell are caused in both vertical and horizontal planes by (i) the temperature difference of the gases passing through the chimney with respect to the ambient, (ii) local moments caused by the eccentric support to the lining affected by projecting brackets inside the shell, and, (iii) ovalling moments caused by non-uniform pressure of wind around the circular section. In the case of the reinforced concrete chimneys consisting of an annular circular cross section, the reinforcement is provided in vertical and horizontal directions on both faces. Global Loads: The Global Loads due to simple bending and vertical compression generate stresses in the annular Volume 46 │ Number 3 │ September 2016

cross-section of the chimney structure. Since the ratio of the thickness of shell to diameter of section is small, the general assumptions of thin shell theory hold good. The vertical reinforcement in the chimney shell can be considered as concentrated in the mean circumference. Computation are made at around 10m intervals over the full height for “Shell Alone” conditions as well as “Completed Chimney”. Local Loads: Coming to the Local Loads, a thermal gradient is setup across the concrete shell thickness with the higher temperature on the inner face in service conditions, thus creating compressive stresses on the inner face and tensile stresses on the outer face. Local moments (longitudinal) created by lining loads on internal corbels are shown in Fig 8 while the transverse ovalling created due to uneven distribution of wind pressures around the circular periphery are depicted in Fig. 9.

3.2 Wind Effects Global Effects The Global Wind effects invariably govern the dimensioning of the structure. For estimation of the wind speed/pressure at the location, IS:875 (Part 3), Ref 1, is used. Both the “3-sec Gust Wind” as well as “Mean-Hourly wind” is determined for arriving at the design forces. Wind Forces: The 3-sec gust wind is determined by the familiar formula (Ref1) for evaluating wind speed at height: Z : VZ = Vb K1 K2 K3

... (1)

Here, K1 = Probability factor. The mean probable design life of the structure is usually taken as:

●

25 years for shell alone condition;

●

100 years condition

for

completed

chimney

K2 = Terrain category factor depending on the roughness of the terrain which is in turn dependent on the height of obstacles (buildings, trees, etc.) of the surrounds K3 = Topography factor which depends upon features like existence of valleys, ridges, cliffs, etc. in the vicinity.

Fig. 8 : Corbel Effect

To ensure the safety and serviceability of the chimney structure, load combinations involving both Global and Local Loads are combined in appropriate manner. Direction of Wind

Direction of Wind

The mean hourly wind is again arrived at using Ref 1 as follows:Vz = Vb x K1 K2’ K3

... (2)

Here,

K1 and K3 shall be as defined equation (1)

K2’ depends on the terrain category as indicated in the code (Ref 1) for various terrain categories.

The wind pressure at any height Z can be calculated from the following equation with the dimensions N/mm2; using equation (1) or (2) as the case may be:

(a) Distribution of the Wind Pressure on the Shell

(b) Dstribution of shears Along Shell Periphery

Fig. 9 : Evaluation of Produced by Ovaling Moments

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Pz = 0.60 V2z

... (3)

The peculiarity of the chimney structure is that it has a response which can be categorized in two lateral directions and evaluated separately. The first is called “Along Wind” and the second as “Across Wind” (perpendicular to the wind direction). They correspond

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to the so-called “drag” and “lift” force coefficients as understood in fluid dynamics. Along Wind: The “Along Wind” effects are first determined from static wind force (N/mm) using the 3 sec wind at any height Z as follows: Fz = 0.60 CDDz V2Z

... (4)

Where,

CD = Drag coefficient, taken conservatively as 0.8 for circular structures

Dz = Diameter of chimney at height Z

Vz = as defined in equation (1)

Fz = Fzm + Fzf

... (5)

Where,

In the design of the chimney first the critical wind speed Vcri is determined at which the Vortex Shedding can “lock-into” the across wind oscillations. This critical wind speed is given by Vcri = fi d/Sn

... (6)

Where,

The “Along Wind” effects are additionally calculated by the Gust Factor Approach using the mean hourly wind which takes account of the turbulent characteristics of the wind as well as the natural frequency of the structure, as explained well in IS:4998 (Part 1), Ref 3, by the following formulation:

number (Re), which, as we know is proportional to the wind velocity multiplied by the diameter. At subcritical Reynold’s numbers, when wind speed is not very high, frequency of Vortex Shedding can come in resonance with that of the chimney, resulting in high amplitude across wind vibrations that could reach high proportions.

Fzm is the wind load in N/mm due to the meanhourly wind pressure, while Fzf is due to the fluctuating component of the wind force at the same height.

The “Along Wind” effects are taken as the more severe of those determined from equations (4) and (5). Across Wind: The “Across Wind” response of chimney corresponding to the “lift” force coefficient is caused by the phenomenon termed as “Vortex Shedding”, Fig. 10.

fi is the frequency of the chimney for the ith mode of vibration,

d = effective outer diameter of chimney at 5/6 height from base Sn = Strouhal number, taken as 0.2 Generally, the first two modes of vibration are adequate for such computations. If the value of Vcri is such that it cannot occur at the site, i.e., it is higher than the steady wind speed calculated by equation (2), then no further considerations are required for Across Wind response. The lateral force Fz on the chimney for Across Wind response follow the procedure given in Ref 3, which look complex but in fact are fairly simple, being based on the work of Vickery and Basu, Ref 7. Aerodynamic Interference: Additional Considerations In case there is a tall object (chimney, silo etc.) in the vicinity of the chimney in question, it may cause “buffeting” leading to additional disturbance or magnification called Interference Effects,

Fig. 10 : Across Wind Effects

Depending on the dimensions, surface roughness and dynamic characteristics of the chimney, “vortices” are formed as the wind flows past it. The air flow pattern around circular objects result in force coefficients which vary with the Reynold’s

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Fig. 11 : Buffeting/Aerodynamic Interference

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Both Across Wind effects as well as Interference effects can be substantially alleviated by providing suitably designed “strakes” (usually in steel) as depicted in Fig. 12. The objective of providing “strakes” is to break the wind pattern so that resonance conditions can be eliminated.

Thermal effects, Corbel moments etc. as per the provision of IS:4998 (Part 1) – 1975 & 1992, Refs 2 and 3. Permissible concrete and steel stresses are defined in the code for various load combinations. The design is essentially performed for vertical and circumferential effects, separately. Since, IS:4998 (Part 1) - 1975 was prepared at the time when working stress method of design was predominantly used, the shell and other parts of the structure are till recently designed by working stress method, though the BIS has come out with the latest IS:4998:2015 combining IS:4998 (Part I) – 1975 and 1992 where limit state design is specified for wind shield.

The most commonly used type of foundations are open spread foundation (raft foundation) or pile foundation. In either case the foundation can be a continuum or sometimes designed as annular circular plate. When the ratio of vertical loads to bending moments is comparatively larger, annular foundation works better, so that P/A can be increased and M/Z can be reduced in order to prevent foundation uplift conditions.

Fig. 12 : Provision of Strakes (Ref 3)

3.3 Earthquake Effects India is divided into four Seismic Zones as elaborated in IS:1893 Part 1,Ref 4. The horizontal seismic coefficient for Design Basis Earthquake (Ref 6) is given by

Ah =

... (7)

Here Z = Zone Factor depending on the location of project in India I = Importance Factor, taken as 1.5 for Reinforced Concrete Chimneys R = Response Reduction Factor taken as 3, Ref 5 Sa/g = Spectral acceleration for 5% damping and as a function of period of vibration and sub-strata below the foundation. Dynamic analysis using response spectrum method is generally recommended for tall structures in consonance with equation (7). The no. of modes to be considered in the analysis should be such that about 90% of the modal mass is excited. The modes are then combined for response (shear, moment, etc.) as suggested in Ref. 5.

Design of Chimney Shell

The shell is designed for various loading combinations involving Dead load, Wind load, Earthquake load, 26

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Design of Foundation

6. Optimisation of Design A careful consideration is required while deciding the shell diameter and foundation size. Since the foundation cost is usually 25% to 50% of the total structural cost of chimney, the foundation design necessarily needs to be optimized. In case the safe bearing capacity of substrata is low, resulting into large foundation size, the shell diameter must be increased to control the foundation thickness. However, a check needs to be exercised to see the corresponding increase in the wind load and earthquake moments and increase of dead weight of shell and related components. A true optimization is determined through combined cost of foundation, shell, platforms, brick liner, insulation (in case of single flue chimney), etc.

Construction Aspects

Chimney shell is either constant in diameter or as is more common tapered. Construction is either done using Slipforming (which is more common) or Jump Forms. Slipforming is illustrated in Figs 13, 14 & 15. Slipforming requires placement of jack rods at 1.0 m to 2.0 m spacing along the shell to support and raise the shuttering and working platforms. The concreting in this case is a continuous process, without construction joints. During slipforming the

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projecting reinforcement dowels for corbels and platforms need to be tackled. A better solution is to provide reinforcement couplers at these locations, with internally threaded sleeves left inside the shell during slipforming.

Fig. 16 : Flue Erection

Fig. 13 : Slip Forming Equipment

Fig. 17 : Strand Jacks for Lifting Flues Fig. 14 : Slip Forming in Progress

Fig. 15 : Slip Forming Equipment

In the case of multi-flue chimney incorporating steel flues, flue erection needs to be planned in a careful manner because each flue segment is to be prefabricated and brought into position at ground level before being hoisted up. The hosting is carried out using strand jacks placed at the respective platform. Members of these platforms need to be strengthened to cater to the condition of flue erection. See Figs. 16 & 17 for illustration. The Bridge and Structural Engineer

References

__, IS:875 (Part 3, Wind Loads) – 1987, “Code of Practice for Design Loads (Other than Earthquake) for Buildings and Structures”.

__, IS:4998 (Part 1, Design Criteria) – 1975, “Criteria for Design of Reinforced Concrete Chimneys”.

__, IS:4998 (Part 1, Assessment of Loads) – 1992, “Criteria for Design of Reinforced Concrete Chimneys”.

__,IS:1893 (Part 1, General Provisions and Buildings) – 2002, “Criteria for Earthquake Resistant Design of Structures”.

__, IS:1893 (Part 4, Industrial Structures including Stack-like Structures) – 2005, “Criteria for Earthquake Resistant Design of Structures”.

Pinfold, G.M., “Reinforced Concrete Chimneys and Towers. Viewpoint Publications, 1975.

Vickery, B.J. and Basu, R.L. “Simplified Approaches to the Evaluation of the Across Wind Response of Chimneys”. Journal of Wind Engineering and Industrial Aerodynamics, Dec 1983. Volume 46 │ Number 3 │ September 2016

STRUCTURAL DESIGN OF TALL BUILDINGS IN INDIA Dr. Satish Kumar obtained his Doctorate from Nagoya University, Japan in 1996. He then worked for NKK Corp. (now JFE) Japan before becoming facutly at IIT Madras. His main area of research are Steel Structures and Earthquake Resistant Design and Testing.

Satish Kumar SR

Professor Structural Engineering Lab, Dept. of Civil Engineering, IIT Madras, Chennai 600036 E-mail: sr.satishkumar@gmail.com

Introduction

The number and distribution of tall buildings has been growing at an exponential pace in India in recent years. This phenomenon is triggered by several factors such as increased demand for housing in urban areas, increased per capita incomes, reduced supply of land, inadequate infrastructure outside major city limits etc. Even in cities with scope for lateral expansion like Chennai and National Capital Region, the number of high-rises being constructed has seen a dramatic increase. Consequently, new designs and technologies are being tried to evaluate their suitability. In this paper, the Focus is on highrises (10 to 40 storeys) rather than skyscrapers (> 40 storeys) as these are much more common and are designed by local architects and structural designers. The paper takes a critical look at the architectural and structural designs employed in India and keeping the developments in the international level as a reference tries to identify the issues that need to be better addressed in the Indian context. A brief review of the common structural systems used for high-rises is given keeping aside the systems typical for skyscrapers. The design issues related to wind and earthquake loads are then addressed along with a reference to current code requirements. The requirements for non-structural components are then briefly reviewed. Fire disasters have plagued the country much more in recent years than in the past and fire design requirements are also reviewed. The motivating factor for the choice of the construction material such as reinforced concrete, steel-concrete composite and mixed systems is the cost and speed of construction. This is also addressed and the need for all concerned to understand the life-cycle cost

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analysis is emphasized. Finally, a few case studies are taken up with which the author had some familiarity to illustrate the possible concerns. Keywords: tall buildings, wind, earthquake, fire, planning, construction

Common Structural Systems

It is well known that design of short buildings (< 10 storeys) is primarily governed by gravity loads while the design of high-rises (10 to 40 stories) is governed by both gravity and lateral loads and the design of skyscrapers (> 40 stories) is governed predominantly by lateral loads. Specific structural systems have been found to be suitable depending upon relative the magnitude of gravity and lateral loads. As the height of the building increases, it becomes important to reduce the weight of the structure and to reduce the cross-sectional dimensions of the structural members. The simplest way of achieving this is to use materials of higher strength to weight ratio such as high strength concrete and structural steel. Steel-concrete composite structures also help in reducing the section sizes. In particular, the slab and wall weights need to be reduced. Profiled deck slabs or hollow core or bubble deck slabs are methods of reducing the slab weight (Fig. 1). Brick Walls are replaced with hollow or light weight concrete blocks or glass curtain walls. As the lateral load dominance increases with height, the moment frame becomes inadequate to limit lateral drifts and additional lateral load resisting systems are needed. The most common lateral load resisting system adopted in steel is the braced frame while the shear wall is the common system adopted in

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reinforced concrete and mixed systems made of steel frame and shear wall (Fig. 2). The location of the shear wall or braced frame in the plan of the structure is also important. Centrally placed cores are often

used but such buildings lack torsional resistance. Peripheral systems on the other hand provide good torsional resistance but interfere with lighting and ventilation.

Fig. 1 : Light-Weight Slabs

Fig. 2 : Common Lateral Load Resisting Systems for High-Rise Buildings

Shear walls can be with uniformly distributed reinforcement or with boundary elements where more reinforcement is concentrated at the ends. Flanged walls are often used to accommodate this extra reinforcement (Fig. 3). Another possibility is to connect two adjacent shear walls by beams at every floor level to get a coupled shear wall. (Fig. 4).The link beams are specially detailed with cross-reinforcement to transfer large shear forces arising out of axial tension and compression in the walls.

Fig. 3 : Shear Walls with Boundary Elements and Flanges

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Fig. 4 : Coupled-Shear Wall

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The moment frames undergo shear type deformation while the shear walls bend like cantilevers. Hence when the two systems are combined, significant wallframe interaction results as shown in Fig. 5. Shear wall frame interaction has been recognized to be an

important phenomena way back in 1964 by Khan and Sharounis. Due to the complete dependence on the computer analysis, this is often not realized by current designers. Soil-structure interaction effects may also warrant consideration in some cases.

Fig. 5 : Shear Wall-Frame Interaction Effect

Another important factor influencing the choice of the structural systems for tall buildings is the speed of construction. Often Indian architects and engineers work out only the material cost and come to a conclusion that reinforced concrete buildings are cheaper than steel buildings. However, if one takes into account the cost of formwork, labour and early returns as well as the salvage value, collectively known as the life-cycle cost, it will be realized that steel buildings will be cheaper for tall buildings. This fact seems to have recently dawned on them and the number of steel buildings have started increasing considerably. Fig. 6 shows that the typical weight of steel required per square metre increases exponentially with the number of stories. This is largely due to the dominating effect of the lateral forces since the steel required to support a floor of certain span will remain same over all the floors under gravity load. The effect has been termed the ‘Premium for height’ (Ali, 2001).

Often dual systems comprising of a steel frame to take gravity loads and an RC shear wall to take lateral loads is used. The shear wall construction is relatively faster due to the fact that it does not have to take gravity loads and techniques like slip form or climbing forms can be used to speed up the construction. Steel concrete composite construction is also helpful in speeding up the construction process without undue increase in the cost of the steel work. Concrete-encased steel columns also enable faster construction and provide better fire-resistance than concrete-filled or pure steel columns. Whatever system is chosen, it is important to give a thought to the construction sequence and scheduling along with the requirements of resisting the various loads acting on the building.

Design for Wind and Earthquake Loads

Fig. 6 : Increase in Structural Steel Weight with Number of Stories

The relative magnitude of design wind and design earthquake loads will depend primarily on the location of the building. A careful observation of the wind and seismic zone map of India reveals that some places are prone for cyclones such as the east coast whereas the earthquake loads are relatively small. On the other hand, places like the north-east, and sub-Himalayan belt are earthquake prone but have low design wind speeds. Thus, buildings located in different places are likely to be affected by wind and earthquake loads to different degrees. Although the lateral load resisting system for wind can cater to resist earthquakes and vice versa, the design requirements can be quite

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different due to the different nature of these two loads.

a cumulative addition of the pressure times area over the entire height of the building.

While wind loads depend on the frontal area in elevation, earthquake loads depend on the overall mass of the structure. Earthquake loads are dominant for buildings with natural time periods up to 2 sec while wind loads are dominant for buildings with natural time periods greater than 2 sec. However, this is a rough estimate and the number of stories at which the transition occurs may vary with shape, size, stiffness, weight and location of the structure.

Wind loads are likely to be critical for pure steel and steel-concrete composite construction especially without shear walls. For buildings with adequate shear walls, the resulting stiffness could be adequate to resist wind and control lateral drifts.

If we take the often used expressions for a quick estimate of time periods, namely T = 0.1 N, where N is the number of stories, this means that time periods will be less than 2 sec for buildings up to 20 stories. The earthquake spectral acceleration coefficient is given as (IS1893:2002).

Sa/g = 1/T

for

T > 0.4

... (1)

When plotted against the height, it can be seen in Fig. 7 that the coefficient decreases rapidly with number of stories and becomes one-fifth of its maximum value for a twenty storied building.

Fig. 7 : Variation of Wind Height Factor and Earthquake acceleration Coefficient for Buildings

On the other hand, wind velocity increases with height by the well-known power law (Holmes, 2001). The Indian Code for wind load IS 875 (part 3): 1987 gives wind velocities for various heights with the velocity up to 10 m height as the reference. From this, the variation of wind height factor defined as the ratio of velocity at height h to that up to 10 m can be found to vary with height as Wind Height factor k2 = 0.77 h 0.1

... (2)

Again assuming a storey height of 3.5 m, the variation of k2 with number of stories can be obtained as shown in Fig. 7. The actual wind load on a building will be The Bridge and Structural Engineer

At the high-rise range of less than 40 storeys (@140 m), dynamic effects of wind are not critical for most buildings. As per IS875 (part 3) gust factors need to be considered only if Time period exceeds 1 sec. However, when wind load is a governing design load and efforts must be made to reduce this load by suitable aerodynamic shapes. Thus, sharp corners should be avoided and either chamfering or rounding of the corners should be done to reduce the flow separation. As per Holmes (2001) Chamfers of the order of 10% of the building width produce up to 40% reduction in the along wind response and 30% reduction in the cross-wind response. Also adequate stiffness must be provided to limit lateral drifts as well as to reduce vibration to imperceptible levels. The design for earthquake load involves three major performance criteria as follows: 1.

Ability of non-structural elements to tolerate a certain amount of drift without damage

Ability of the structural elements to resist a design basis earthquake with essentially elastic response and to limit storey drifts.

Ability of the structural elements to deform and dissipate energy without collapse

The regularity in the disposition of structural elements in plan and elevation has been found to be a deciding factor in the earthquake performance of buildings (Naeim 2001). Irregularities in plan tend to induce twisting of the frames and will require large torsional stiffness of the building as a whole which a centrally placed core would be inadequate to resist. Irregularities in elevation tend to concentrate damage over a single or few stories there by giving inadequate performance. Design for elastic response and drift limitations is carried out by the usual procedures of structural design. For tall buildings, modal analysis along with the response spectrum method is usually adopted. Design for inelastic behaviour and damage limitation is done by following the capacity design procedures wherein the structural elements are detailed to achieve Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

a ductile failure mode along with considerable energy dissipation capacity. Provisions for ductile detailing of RC members and joints are given in the Indian code IS 13920: 1993 and for steel structures are given in Section 12 of IS 800:2007. A desirable development at present and particularly after the Gujarat earthquake of 2001 is that most reinforced concrete and even composite buildings are designed with shear walls as their lateral load resisting system. However, the number and distribution of such walls and the capacity of the diaphragm to transmit lateral loads to these shear walls is questionable. Most walls are designed with uniformly distributed reinforcement and do not have boundary elements or flanges. Even the code for ductile detailing IS13920:1993 is silent about walls such walls. Also no stipulations on the adequacy of diaphragms are available and it is a concern that many tall buildings may not act integrally with their shear walls. These issues need urgent attention. The performance of non-structural systems is important under wind and earthquake. At present, no specific guidelines exist for adopting a particular nonstructural system in a high-rise building. Common wall materials used in residential and commercial buildings are hollow blocks, light weight blocks and glass. In cyclone prone areas, block masonry may be preferable over glass curtain walls. This fact was borne out of the recent cyclone HudHud which struck the city of Visakhapatnam. Block masonry also provides better thermal insulation which is an important factor in a hot country like India. The design of glass curtain walls need to be addressed from several angles such as their resistance to normal wind pressure, impact from flying debris, ability to accommodate lateral drifts, ability to resist fires etc. At present no codal provisions or even guidelines

exist for the design of glass facades or curtain walls for each of the above effects and failure of glass panels has increased as witnessed in Chennai Airport. A variety of Vibration control systems have been used across the world to reduce the building vibration due to both wind and earthquakes. Although feasible,the relatively higher cost does not warrant their use in India in most cases.

Design for Fire

It is very important to design high-rise buildings for the possibility of a fire accident. The fire service in most Indian cities are neither trained nor equipped to handle fires in high-rises. Both reinforced concrete and steel buildings are prone for damage under a major fire although the concrete buildings have somewhat better resistance than steel buildings. Fire design involves three major parts namely 1) Architectural planning; 2) Structural design and 3) fire fighting. In architectural planning the location of fire barriers, escape routes etc are addressed while in structural design, the ability of the members and the structure as a whole to withstand a certain fire load for a certain period of time is assessed. Several codes are available to guide architects and designers on the issue of fire design (IS1642:1989). However, test data and design experience are yet meager and need to be improved. Fire resistance also depends on the minor detailing followed. For example, the projection of the diaphragm beyond the wall line is important to prevent fire ‘jumping’ from floor to floor (Fig. 8). The provision of double-curtain walls can also be beneficial towards this goal (Fig. 9) (O’Connor 2008). Similarly, the beam-slab system can be protected by using proper refractory boards in the false ceiling.

Fig. 8 : Curtain Wall Floor Connection Details to Contain Fire

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Fig. 9 : Double-Skinned Curtain Wall detail used in JSW Building, Mumbai

Case Studies

A few case studies are taken up to illustrate the above design requirements and the possible deviations in assumed and actual behaviour is emphasized to identify the need for further research and codification. The first case study is a 40 storied building in Mumbai with a small plan of 21 x 27 m.The building is divided into two parts – 1) the lift and staircase part and 2) the office space. For reasons of fire safety, it was decided to make part 1 with a large number of RC shear walls while the office space was made with structural steel to get better lighting and outside view (Fig. 10). This highly un-symmetric disposition of structural elements invited high torsional loads and so the steel frame had to be braced with large braces. To overcome the problem, bracing members spanning over six stories were adopted along with belt trusses as shown in Fig. 9. Also square concrete-filled tubular (CFT) columns were adopted which made beam-to-column connections complicated as in Fig. 10 even though these connections had to transfer only shear. The beams were protected with intumescent coatings for fire resistance but the columns were left unprotected as the concrete filling inside carried considerable part of the load. The columns along the shear walls were I-sections encased in concrete (CES). The speed of The Bridge and Structural Engineer

construction made possible by use of steel framing could not be used to advantage as the pace was controlled by the concrete part.

Fig. 10 : Structural Plan

The second case study is that of a 40 storied building made of slim-deck secondary beams and concreteencased steel section columns. It is a ‘Plaza’ type building with a large plan for the first few stories Volume 46 │ Number 3 │ September 2016

and two tall towers rising above. The towers are sharp cornered rectangles with a large width and a much smaller depth. Typical tower was of plan size 25 x 41 m and height 154 m. This makes them very flexible in the depth direction and so shear walls are introduced to get adequate stiffness. The entire building is founded on RC Raft foundation which makes soil-structure interaction a possible issue. The primary reason for adopting steel frame for gravity loads was to ensure fast-track construction and also to get reduced member sizes. Also semi-rigid end-plate connections were adopted for the beams to reduce span bending moments. Thus, the issue of wall-frame interaction arises which again cannot be solved since the rotational stiffness of the connection itself is an unknown. This also has an impact on the estimated time period as the building would be stiffer than the computer model. However, due allowance was given to this based on ‘engineering judgement’.

that there needs to be considerable research to clarify the discrepancies between the assumed against the exact behavior of these buildings. The building has set a new record in construction speed in the country.

The paper summarizes the various issues to be addressed in the planning and design of tall buildings. The focus was limited to high-rises of up to 40 stories rather than skyscrapers of over 40 stories height. The typical structural systems for reducing gravity loads and for resisting lateral loads are mentioned and their advantages and disadvantages are discussed. The effect of wind and earthquake loading and fire on design of both structural and non-structural systems are also briefly explained. By means of a few case studies, the need for better analysis procedures and guidelines for design are emphasized.

Fig. 12 : A Plaza-Type Building using Shear Walls and Composite Construction

Fig. 11 : Photoes Showing Bracing and Connection Details

Further, steel of grade 345 MPa was used. This poses two problems –one with the weld ability which depends on the metallurgy and the other with quality of weld electrodes. The former was sorted out with the steel maker while the latter was solved by importing the weld electrodes. Another major problem with high strength steel is the susceptibility for local bucking thereby giving more stringent width-to-thickness ratios than mild steel. This was taken care by encasing the sections in reinforced concrete (Fig. 10). However, there is no data available to quantify the encasement effect on the local buckling strength. It may be realized

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Summary and Conclusions

References

ALI M.M., and MOON K.S., “Structural Developments in Tall Buildings”, Architectural Science Review, Univ. of Sydney, Vol. 50, 2007,pp 205-223.

IS 1893:2002 “Criteria for Earthquake Resistant Design of Structures- General Provisions and Buildings”, Bureau of Indian Standards, New Delhi, India.

IS 875 (Part 3):1987, “Code of practice for Design loads (other than Earthquake) for

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Buildings and Structures”, Bureau of Indian Standards, New Delhi, India.

NAEIM F., “The Seismic Design Handbook”, Springer, 2001.

IS13920:1993 “Code of Practice for Ductile Detailing of Reinforced Concrete Structures Subjected to Seismic Forces”, Bureau of Indian Standards, New Delhi, India.

PEER, “Guidelines for Performance-Based Seismic Design of Tall Buildings”, Version 1.0, Report No.2010/05, College of Engineering, University of California, Berkeley, 2010.

IS 800:2007, “Code of Practice for General Construction in Steel”, Bureau of Indian Standards, New Delhi, India.

HOLMES J.D., “Wind Loading of Structures”, Spon Press, 2001.

10. IS 1642:1989, “Fire Safety of Buildings (General): Details of Construction-Code of Practice”, Bureau of Indian Standards, New Delhi, India.

SARKISIAN M.P., “Designing Tall BuildingsStructure as Architecture”, Routledge, New York, 2012.

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11. O’CONNOR D.J., “Building Façade or Fire Safety Façade”, Council on Tall Buildings and Urban Habitat Journal, Issue II, Chicago, 2008.

Volume 46 │ Number 3 │ September 2016

Experience With Medium-Tall Buildings in Germany – A Case Study

Boris REYHER Associate schlaich bergermann partner Berlin, Germany b.reyher@sbp.de

Mike SCHLAICH Professor Berlin University of Technology Berlin, Germany Mike.schlaich@tu-berlin.de

Dr.-Ing. Boris Reyher, born 1970, received his engineering degree from Technische Universität Berlin and University of Michigan. With more than ten years experience his executed projects range from office and residential highrises to stadia roofs and long-span bridges.

Prof. Dr. SC. Techn. Mike Schlaich, born 1960, received his civil engineering degree from the ETH Zurich. He is managing director of schlaich bergermann partner and certified German proof checking engineer. In 2015 he has been awarded the Gold Medal by The Institution of Structural Engineers

Summary This paper deals with the authors’ experience in the structural design of medium tall buildings in the 100m class, which are common and popular in Germany. A case study featuring interesting structural solutions is provided. Keywords: medium-tall buildings, office towers, post-tensioned floor plates, ......

with slenderness are normally not severe. This leads to a number of interesting architectural approaches, which call for creative structural solutions. Some interesting aspects and solutions for the structural design of an office building of 90m height, Exzenterhaus in Bochum, are presented by the authors in this paper.

Introduction

The term ‘tall building’ is a relative one. Germany is not known for record-breaking building heights. In fact, according to German building regulations, any building exceeding 22 m in height is regarded a tall building. While there is only a little more than a dozen of buildings in Germany exceeding the 150 m mark, towers up to 100 m height are widely popular as investment projects or for company headquarters. schlaich bergermann partner have been responsible for a number of tall buildings, including several between 60 m and the 100 m mark. At this scale, leasable floor space efficiency is generally very good while at the same time structural issues connected 36

Volume 46 │ Number 3 │ September 2016

Fig. 1 : Exzenterhaus, Bochum, Germany

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vertical axis pointing in different directions to connect with urban landmarks of the town.

Figs. 4, 5 : Architectural Model

Fig. 2 : LVM Office Tower, Münster, Germany

The architectural design and the location of the existing bunker posed a number of challenges to the structural design, which were solved in close coordination with the architect: ●

The existing bomb shelter featuring thick concrete external walls and roof, had been constructed in the 1940’s to withstand air raids, but was not meant to resist the wind forces acting on a 90 m tall building, especially with respect to the foundations.

●

A subway tunnel and station built in the 1980’s lies immediately adjacent to one side of the site complicating the foundation situation, as no changes in the soil pressure acting on the subway structure was allowed.

Fig. 3 : Festo Automation Center, Esslingen, Germany

The Architectural Design of Exzenterhaus

The Berlin-based architect Gerhard Spangenberg’s design concept for the Exzenterhaus (lit. Eccentric Building) in the western German town of Bochum is based on the idea of placing an 18-storey office building on top of an existing bomb shelter from World War II times. The bunker had been sitting in the centre of Bochum for some 60 years awaiting a new civilian usage as well as a connection to the surrounding urban context. The architect achieved these goals by structuring the additional building body in similar proportions to the existing cylindrical bunker and by arranging three packages of floor levels eccentrically to the common

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Fig. 6 : Typical Floor Plan

Volume 46 │ Number 3 │ September 2016

●

The bunker’s interior staircase walls were not fit to resist the vertical forces from 18 additional floor levels.

●

The eccentric floor plates featuring large cantilevers had to be constructed lightly so as to limit the vertical dead load acting on the foundations.

●

Fair-face concrete surfaces of highest quality were required architecturally for all core walls and slab soffits.

Structural Concepts and Realisation

outward eccentric moment is resisted in the floor slabs at the top and bottom of the inclined columns through in-plane normal forces.

3.1 Lateral Stability System The tower depends on the central cylindrical core shaft of reinforced concrete for lateral stability. At the bunker levels, the 2 m thick external bunker wall, which proved to possess surprisingly high compressive strength values on the order of 50 MPa and more, also acts as horizontal bracing. Lateral load transfer from the central core to the external bunker wall takes place at the bunker top through a newly installed strong transfer slab.

Fig. 9 : Column Transfer at Bunker Top

3.3 Foundation The design of the piled foundation required to transfer the building loads safely into the underlying soil was done in cooperation with GuD Geotechnik und Dynamik Consult engineers, Berlin. The main challenge was how to design and construct a new deep foundation consisting of large diameter bored piles within the confines of the existing outer bunker walls after partially dismantling the 2.0 m thick base slab of the structure. The requirements posed by the owner of the adjacent subway line, BOGESTRA, precluded any changes in the horizontal and vertical soil pressures acting on the existing concrete tube and subgrade station walls. An extensive monitoring programme was set up by BOGESTRA before commencement of construction in order to detect any cracks forming which would immediately have stopped the construction works on the tower.

Figs. 7, 8 : Vertical and Lateral Load Paths

3.2 Vertical Load Paths Vertical floor loads are resisted by the central core and a total of 15 RC columns arranged concentrically around the core axis. The columns of the additional storeys transfer their vertical forces to the external bunker wall. Due to the greater radius of the column ring, an offset between columns and bunker wall is managed through a system of inclined columns at an intermediate plant level above the bunker top. The

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The foundation design resulted in a total of 12 no. bored RC piles of 1200 mm diameter. After the decision had been taken by the architect and structural engineers to remove the roof and all internal walls and slabs of the bunker, the use of large diameter piling rigs became feasible. It was still a major logistical challenge to insert the 75-ton rig through the open top of the remaining cylindrical bunker wall. A 500-ton mobile crane was used to this end. Due to the requirements posed by BOGESTRA, the pile shaft friction had to be made ineffective over the first six metres of pile length measured from the underside of the new pile cap. This was achieved by placing permanent steel tube sheathing down to this level. Furthermore, any bearing pressure was

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to be precluded at the underside of the pilecap, which is normally disregarded in the design of deep foundations and not considered harmful. This was achieved by pouring the 2.0 m thick new pilecap on special mats with a defined time-dependent stiffness, which are normally used for drainage at the outside of basement walls. Due to BOGESTRA’s concerns regarding the resulting global lateral wind force acting on the foundation, additional inclined prestressed ground anchors had to be installed at the pile cap to prevent any potential increase in the stress level at the subway tube.

●

Maintaining only 65 centimetres of total structural build-up while allowing for dense ducts and installations within the floor space above the slab.

●

Keeping dead load to a minimum so as not to overload the foundations and existing bunker wall.

In addition, the slabs were to be fitted with groundwater tempering hoses for cooling in the summer.

Fig. 10 : Piling Rig Being Lifted into Bunker

3.4 Floor Plate Design The structural design of the new eccentric floor plates with cantilevers measuring up to 4.0 m beyond the ring of columns posed a great challenge. Several options of slab systems were studied and discussed. The governing criteria were: ●

Having a flat underside surface which rises up towards the edge of the floor plate in order to allow more light entering the floor space.

●

Having highest-quality fair-face concrete surface at the underside.

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Figs. 10, 11 : Hamburg TV Tower

The final choice for the structural system was found in the analogy of a perfectly rotationally symmetrical television tower capsule structure. In the 1960’s, as TV towers were largely built across Germany, a common approach was to build conical shell shafts to support the ring slabs of the capsules. To deal with the large tensile stresses concentrating at the top edge of such a conical shell, circumferential prestressing tendons were used.

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For the floor slabs of Exzenterhaus, a very similar concept was developed with the main difference of not having rotational symmetry and also of having a very low inclination in the ring slabs. Nevertheless, it could be proven by 3D structural analysis that the desired shell action had a significant beneficial effect and resulted in only 25 centimetre thick concrete slabs. Circumferential unbonded prestressing tendons were used around parts of the perimeter, where the cantilevers are substantial. The anchorages of the 4- to 5-strand bundles were placed within the inclined slab edges carefully detailed to prevent any compromise to the high-grade concrete surfaces at the soffit. Vertical deflections of the slab edges could effectively be controlled by the degree of unbounded posttensioning. Still, much coordination was required with the design and construction of the curtain wall facade to make sure that differential vertical movements of 30 mm or less between two successive floor plates were acceptable to the facade structure under all circumstances.

Concern was raised about the long-term stability of the existing bunker wall considering extensive vertical cracks and in lieu of detailed knowledge of the existing rebar in the concrete. Therefore, it was decided to tie the bunker wall together with external unbounded PT tendons running horizontally around the perimeter of the bunker. These tendons would later be covered within the necessary thermal insulation layer for the bunker levels. 3.6 Structural Analysis and Design The structural analysis and member design was carried out using a global finite element model for the analysis of load take-down and core wall design. The global model was programmed parametrically in SOFiSTiK FEA software and featured only a rather coarse element mesh which was chosen to yield sufficient accuracy but keeping numerical effort manageable. For the design of the shell-type floor slabs, refined local models were used to represent the bending behaviour of the prestressed floor slabs in greater detail. The global effects of differential column and core wall settlements were included in special constraint load cases. This included consideration of the time-dependent effects in the RC members in compression and the cancelling of tolerances and vertical strains during floor-to-floor erection. A detailed semi-local model of the external bunker wall was used to model the effects of loading from the additional floors and of modifications to the existing concrete structure. Furthermore, strut-and-tie and other simplified local models were used throughout to analyse and design details.

Fig. 12 : Layout of PT Tendons

3.5 Modification and Reinforcement of Existing Bunker Structure The existing bunker structure was completely stripped of all internal walls and floor plates. The existing base slab was removed to make room for the new deep foundation. Furthermore, a large main entrance opening was cut over two levels. Additional door openings were made at the lowest level for service entry. At the connection with the new pilecap, radial corbel fingers needed to protrude into the existing wall to transfer vertical wall loads. For this reason, a number of cavities had to be cut into the foot of the bunker wall. 40

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Figs. 13, 14 : Global and Local FE Models

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3.7 Construction Process First, the demolition of the internal bunker structure was carried out. Then, the construction of the bored piles took places inside the remaining bunker wall. The new office floors were erected level-by-level with two sets of reuseable steel formwork sets for the fair-faced concrete slabs. Each floor took an average of two weeks to place formwork and reinforcement, concrete and apply the post-tensioning to the hardened

slabs. In order not to waste time waiting for the minimum required concrete strength for prestressing, subsequent floors were already constructed prior to prestressing of a finished floor with the use of temporary props through several levels. The erection of the curtain wall facade followed after the concrete construction with a lag of approximately five levels. Construction of structure and facade was completed in July 2012.

Figs. 15, 16, 17, 18 : Construction Photos

Conclusion

With the Exzenterhaus, the town of Bochum has gained a new urban landmark which has been recognized

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throughout the region. In close cooperation between architect, structural engineer and geotechnical engineer, a challenging set of problems could be Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

solved with a convincing and comprehensive design approach including many innovative details. The

existing bunker structure was thus given a new purpose and a new life as part of the urban context.

Figs. 19, 20 : Completed Building

AG, 2015, p. 208-211, ISBN 978-3-03768189-3.

5. Literature 1.

Cardno, Catherine A.: German High-Rise Built Atop Historic Bomb Shelter, ASCE’s Civil Engineering Magazine, September 2012. Van Uffelen, Chris: CONCRETE - Pure. Strong. Surprising., Braun Publishing

Volume 46 │ Number 3 │ September 2016

Holgate, Alan: The Art of Structural Engineering, Edition Axel Menges, Stuttgart/ London, 1997.

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Some Basic Guidelines in Use of Steel Rebars in High Rise Buildings

Dr. N.V. Nayak Principal Advisor, Gammon India Limited, Mumbai â&#x20AC;&#x201C; 400 025 nvn@gammonindia.com

Shantilal Jain Chartered Engineer

Narayan V. Nayak is born in the year 1936 and graduated in Civil Engineering from the University of Bombay in the year 1959. He secured his M.Tech. in Civil Engineering from the Indian Institute of Technology, Bombay and his Ph.D from the University of Wisconsin, U.S.A. in the year 1970 (GPA, 4.0/4.0). He has 8 years of teaching, 8 years of consultancy & 35 years of practical experience.

Shantilal H Jain received his Bachelor degree in 1981 and Master degree in 2006 from VJTI. He has 27 years experience in structural design of high rise building.

Preamble

In India particularly in Metros, many high rise buildings are coming up. Our Indian codes are not geared up for such buildings. Based on past experience, it can be said that it may take long time before codes are updated to meet this requirement. Often developers appoint Architects, Structural Consultants from Advanced Countries. According to authors, that does not solve the problem. As per the authors, there are competent persons and organisations to meet our requirement within India, though we may have some of these organisations/persons from advanced countries as our advisors. With respect to high rise building reinforcement and concrete, the following points need to be given due attention. While selecting re-bars, particularly for high rise structures, the following four parameters need to be given due attention. Unfortunately, due attention is not given by many concerned people/organisations including foreign consultants from the developed countries. Hence, the authors felt it is very desirable to publish this paper giving some basic guidelines.

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Re-bars from Primary and Secondary Players

In India re-bars are manufactured by both the Primary and Secondary players. Even for high rise buildings some developers use steel re-bars manufactured by secondary players, though it is not permitted by Indian Code (IS 1786). Many experts from developing countries are not aware of steel manufactured by secondary players as such players are absent in their countries. The consistency of various properties of a rebar to a great extent depends on the consistency of the rim formation which in turn depends on a very sophisticated process of Thermo-Mechanical Treatment process (TMT). A non-uniform ring formation, especially if there are points of non-martensitic area like horse shoe, may attract fatigue crack nucleation. The manufacturing (quenching) processes adopted by primary players are through specially designed water boxes. Water spray, at high speed, is utilized while quenching the steel from high to low temperatures. This process is critical for dependable and consistent Martensite Rim for steel rebars. This process is not adopted by secondary players. Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

With inconsistent martensite rim, the corrosion starts where rim is minimal as shown in Fig. 2 (secondary player process). In Fig. 1, rim is uniform (primary player process) and hence corrosion possibility is relatively low w.r.t. secondary player process. Even if physical properties are consistently met with, the chemical composition is of prime importance in selection of rebars particularly for high rise buildings, as noted in next para.

Fig. 1 : Typical Macrostructure of Rebar of (Primary Player) Primary Players

Even developers who use steel manufacturered by primary players, some use rebars of Fe 500, some use Fe 500D, some use Fe550D and so on. Subsequent presentation in this annexure will indicate that for high risers, our steel rebars needs to have impurities and properties better than even Fe500D etc.

Impurities in Steel Rebars and their Influence on Steel Behaviour

Amongst the various impurities, the crucial impurities which have a significant impact on the behavior of rebars are Sulphur and Phosphorus. If the percentage of sulphur and phosphorus increases in steel, the ductility of rebars/structures get adversely affected as noted in Table-1 and may even lead to failure of buildings (Adeleke/and JK Odusute Fig. 2 : Typical Macrostructure of Rebar (Secondary Player) (2013)). Table 1 : Effect of Sulphur and Phosphorus on Rebar Properties CAUSE EFFECT Sulphur in Rebar due to Iron (Fe) forms Iron sulphide (FeS) in Raw Material presence of sulphur which forms galvanic cells with Fe stimulating corrosion. Impact strength of steel is found to gradually decreases with increase in sulphur. Impact strength exhibits toughness of rebar under suddenly applied loading condition FeS itself and a eutectic of Fe and FeS having a very low melting point (988°C) segregates to grain boundary in steel.

RESULT Enhances Rebar Corrosion

Failure of structure during unforeseen loading in case of earthquakes or strong winds Severely reduces ductility at high temperatures. Hence, failure can occur during events like fire. Phosphorus in Rebar due Iron(Fe) in presence of Phosphorus forms Reduces the toughness and to raw material Iron phosphide - a highly brittle material ductility of steel which segregates to grain boundary in steel .It increases stress and causes various forms of embrittlement which severely reduces ductility. Phosphorus content of over 0.04% makes Poor Weldability weld brittle and increases the tendency to crack. TMT Process without Non uniform tempered Martensite rim Reduces fatigue strength uniform Water spray with Ferrite/Pearlite Core. of Rebar and also enhances quenching Box. corrosion

Volume 46 │ Number 3 │ September 2016

IMPLICATION Negative (not desired) Negative (not desired) Negative (not desired) Negative (not desired)

Negative (not desired) Negative (not desired)

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For above reasons, various national and international standards have restricted the limit

of sulphur Table 2.

and

phosphorous

noted

Table 2 : Limit of Sulphur and Phosphorus in Different Standards of Rebar (Indian/International) Product

Standards

Sulphur (max)(%)

Phosphorus (max) (%)

Sulphur+ Phosphorus (max)(%)

Rebars-India

IS-1786-Fe 500

0.055

0.105

IS-1786-Fe 500 D

0.040

0.075

Tata Tiscon Fe 500D

0.035

0.070

Rebars – USA

ASTM A 706

0.045

0.035

Rebars -Japan

JIS-G3112 (SD-490 grade)

0.040

Not mentioned

by introduction of new grades with lower sulphur and phosphorus limit. Permissible upper limit of these values have been consistently reduced over the years to get more dependable product which will yield more durable structures. However, even these lower limits of sulphur and phosphorous were found to be inadequate for high rise buildings.

Fig. 3

In Fig. 3 is the graph on the trend of maximum limit of sulphur and phosphorus in Indian standards evident

The Ministry of Land, Infrastructure Transportation and Tourism, Japan in collaboration with fifteen construction companies have developed guideline of High strength materials to be used for high rise buildings as given in the Table 3.

Table 3 : Limit of Sulphur and Phosphorus in High Rise Building Projects in Japan Tensile Strength/ MPa

Yield Point (YP)/MPa

Chemical Component (%)

USD685A

≥ YP/0.85 = 1.176 YP

685-785

≤ 0.50

≤ 1.50

≤ 1.80

≤ 0.030

≤ 0.050

USD685B

≥ YP/0.80 = 1.250 YP

685-755

≤ 0.50

≤ 1.50

≤ 1.80

≤ 0.030

≤ 0.050

USD980

≥ YP/0.80 = 1.250 YP

≥ 980

≤ 0.80

≤ 1.20

≤ 2.00

≤ 0.030

≤ 0.050

C = Carbon, Si = Silica, Mn = Magnesium, P = phosperous, S = Sulphur, Cu = Copper

At present in India none of the even Primary players are manufacturing rebars meeting the requirements of chemical components noted in Table 3. But some of these players like Tata Steel have conveyed their confidence of manufacturing the steel rebars meeting above requirements once it is specified.

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Steel Rebar Congestion

It is observed based on some of the projects constructed/ being constructed in Mumbai, steel rebars congestion is very high. Some of these projects are listed in Table 4. Intentionally names of the projects have been withheld and these are listed as (A), (B)...... (E). Volume 46 │ Number 3 │ September 2016

From the Table it can be noted that in some cases, steel reinforcement in core wall is as high as 0.59t (of Fe500 grade) per cum of concrete. Fig. 4, illustrate one of such congested steel reinforcement.

Spacing of Bars

When congestion is high, bar spacing gets reduced (Fig. 4) below the minimum requirement. The minimum bar spacing is governed basically by two factors. (1) spacing should be such as to allow concrete to flow easily and allow proper compaction. (2) spacing is also essential for proper bond to develop.

Table 4 : Denisity of Rebars in Concrete in Corewall of Tall Buildings in Mumbai (Height more than 150 m) Name of the Project (A) (B) (C ) (D) (E )

Height (m) Steel (kg/cum) (Approx.) 153 320 203 220 280 462 284 515 429 590

When spacing of bars is closer and congestion is preventing proper flow, vibration and compaction, the minimum precaution one should take is to specify Self Compacting Concrete (SCC). None of the above projects noted in Table 4 , to our understanding, had/ has SCC. For spacing of bars, one may refer Indian Code IS 456, Eurocode EN 1922, American Code ACI 318. Broadly speaaking, the clear spacing between parallel bars shall be maximum of the following.

Fig. 4

Such dense reinforcement have many limitations. Some of the limitations are highlighted below:

Not less than diameter of largest bar (EN 1992 -1-1; IS 456, ACI 318)

Not less than maximum size of the aggregate (MSA) + 5mm (EN 1992-1-1; IS 456)

Not less than 20mm (EN 1992-1-1)

Not less than 25mm (ACI 318)

Many consultants/developer violate this requirement in high rise buildings.

Fig. 5

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

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IS 456 clearly states that diameters of the bars noted above are effective diameters and not nominal diameters. Bars are provided with ribs and hence effective diameters are more than nominal diamters. In Fig. 5, Table provides “nominal diameter, minimum and maximum outside (OD) diameter of bars” as per one manufacturer. For e.g. as per this table, for 40 mm dia (nominal) bar, OD considering ribs could be 43.2 mm (minimum) or 45.6 mm (maximum). Hence, to work out clear spacing we may consider effective bar diameter as nominal 40 mm diameter.

= 44.4 mm against

Congestion of steel bars can be reduced using higher strength bars. Generally in India, we use Fe500 steel, though many primary manufacturer can supply Fe550D bars and Tata Steel can supply Fe600 steel. But in developed countries much higher strength bars are used. In Japan they produce steel bars equivalent to Fe980 (USD 980) which is 1.96 times stronger than Fe500, hence spacing of these bars will be roughly 2 times the Fe500 bars. Similarly, in USA they produce rebars equivalent to Fe690. Hence, India should embark on manufacturing and encouraging use of such high strength rebars, if we have to construct high rise buildings meeting the codal requirements w.r.t. spacing etc. which are very reasonable for producing durable structure.

6. Lateral Reinforcement : Bar Diameters It is observed that in some of the high rise buildings, lateral reinforcements of diameter 25 mm and above are also used, though Indian Code IS 456 (2000), permits use of lateral reinforcement upto and including 20 mm. However, higher diameters are used by some consultants particularly from advanced countries. Requirement of BIS is reasonable. If main bars say vertical , need to be effective without buckling, there has to effective contact between lateral and vertical reinforcements. Such perfect contacts are difficult to achieve particularly when steel rebars congestion is high and hence, quite often lateral reinforcements need to be deformed by pushing in position as shown in Fig. 6. Such deformation of rebars is possible when diameter of the lateral bar is 20 mm or less. In fact the authors prefer 16mm or less from the ease of convenience of deforming the bars in position at site., If theoretical analysis requires use of lateral reinforcement of bars of diameter more than 16/20mm, then it is suggested to replace bigger diameter lateral bars by 2 bars of smaller diameter such as 8 mm, 10 mm, 12 mm , 16 mm etc. to get the required amount of reinforcement in lateral direction”.

Fig. 6

Summary

It is essential we use steel rebars manufacturered by primary players and not to use steel rebars manufacturered by secondary players unless it is an unimportant or temporary structure.

It is very desirable to restrict contents of

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sulphur and phosporous each to less than 0.03% particularly for high rise buildings. For other buildings we may use steel rebars meeting the codal requirements w.r.t. these impurities. However, steel rebars with lower percentages of impurities of 0.03% or less for each sulphur and phosphorous are always preferable.

Volume 46 │ Number 3 │ September 2016

When designing the RCC structures it is very desirable to meet the codal requirements w.r.t. spacing of bars etc.

Use of higher strength steel bars is always preferable to reduce congestion of steel paritcularly in high rise buildings. At present in India upto and including Fe600 rebars are manufacturered. It is desirable that higher strength bars are manufactuered for high rise buildings to reduce congestion.

In high rise buildings, if it is difficult to a certain extent to avoid congestion of rebars, it is very desirable to specify self compacting concrete to avoid honey combing and to achieve proper compaction.

References

Adeleke, AA and JK Odusute, (2013) “Evaluation of the Mechanical Properties of Reinforcing Steel Bars from Collapsed Building Sites”, Journal of Failure Analysis and prevention, pp.737-743.

ASTM A-706/A-706M, Standard Specification for Deformed and Plain Low Alloy Steel Bar for Concrete Reinforcement. Basu, PC, P, Shylamoni, and AD, Roshan, (2014) “Characterization of Steel Reinforcement for RC Structures : An Overview and Related Issues” The Indian Concrete Journal, Jan issue, pp.19-30.

BIS IS-456 : (2000) Plain and Reinforced Concrete Code of Practice.

BIS IS-1786 : (2008) High Strength Deformed Steel Bars and Wires for Concrete Reinforcement Specification.

Castro, DBV, JM, Venttura, COFT, Ruckert, D, Spinelli and BWW, Filho (2010) “Infleunce of Phosphorous Content and Quenching/ Tempering Temperatures on Fracture Toughness and Fatigue Life of SAE 5160 steel” Materials Research pp. 445-455.

Cyril, N, A, Fatemi and B, Cryderman, (2008) “Effects of Sulphur Level and Anisotropy of Sulfide Inclusions on Tensile, Impact and Fatigue properties of SAE 4140 Steel” Paper 2008—1-0434, SAE international report.

Volume 46 │ Number 3 │ September 2016

Ebrahimi, Z and G, Krauss, (1984) “Mechanism of Tempered Maertensite Embrittlement in Medium Carbon Steel”, Acta Metallurgica Vol.32, No.10, pp.1767-1777.

Hippsley, CA and NP, Haworth, (1988) “Hydrogen and Temper Embrittlement in 9Cr-I Mo Steel” Materials Science and Technology, Vol.4, pp.791-802.

10. Horn, RM and RO, Ritchie (1978), “Mechanism of Tempered Embrittlement of Low Alloy Steel”, Metallurgical Transaction A Vol.9A, pp.10401053. 11. Miyajima, M, “The Japanese Experience in Design and Application of Seismic Grade Rebar” 12-Proceedings of International Seminar on Production and Application of High Strength Seismic Grade Rebar Containing Vanadium, Beijing China. 12. Rudyuk, SI, EI, Feldman, EI, Chernov, and VF, Korobeinik, (1974), “Effect of Sulphur and Phosphorous on the Properties of Steel 18B” Metal Science and Heat Treatment, Dec.1974, Vol.16.16 Issue 12. Pp. 1056-1059. 13. Spaeder, GJ, (1975) “Effect of Sulphur on Through-Thickness Properties of C-Mn Steel”, Welding research supplement, June 1975, pp.196s-200s. 14. Prof. Subramanyan (Retd. IIT Bombay) Article (2015) “Mumbai is at Risk of Earthquake, IIt Prof. warns against highrises” in Times of India January 7th pp.4. 15. Tata Steel Brochure (2015) : “Adding Structural Integrity to High-Rise Buildings by Using High Strength Grade of Rebars”. 16. Viswanath, S, Prasada, LN, Radhakrisna, HS, Natraja, (2004), “Substandard Rebar in the Indian Market”, The Indian Concrete Journal, Jan.issue Vol.78, Nol, pp.52-55. 17. Wranglen G, (1969) “Review Article on the Effect of Sulphide Inclusion on the Corrodibility of Fe and Steel”, Corrosion Science, Vol.9, pp.585-602. 18. Zabilskii VV, (1987) “Temper Embrittlement of Structural Alloy Steel” Metal Science and Heat Treatment, Vol.29, Issue 1, pp.32-42.

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TALL AND HANDSOME NATURAL DRAFT COOLING TOWER (NDCT) SHELLS IN INDIA V.N. Heggade Member Board of Management of Gammon India Limited Mumbai venkat.heggade@gammonindia.com

Summary The column based hyperbolic reinforced concrete cooling towers are ecstatic in aesthetics, gigantic in nature and complex in geometry. The evolution of ‘Design and Constructions’ of these towers have come long way in India and as on date, 202 meters height Kalisindh Natural Draught Cooling Tower is the tallest standing tower in the country as well as world. In the past, before the computer revolution, the analysis used to be done only for membrane forces in shell elements. However, as the towers grew in stature beyond 100.0 meters height clubbed with infamous episode of Ferry Bridge cooling tower in UK emphasized the relevance of bending analysis coupled with membrane forces. Invariably it is found that the structural criticality in NDCTs is due to the wind induced dynamic forces, which compounds the complexity of the above analysis. In the paper followed, the complexity of wind induced forces and rationale behind the applications, the progression from membrane analysis to bending analysis, the evolution of different design parameters for NDCTs are discussed in length.

Mr. V.N. Heggade has around 31 years’ experience in Designing & Constructing bridges, aqueducts, industrial structures and marine structures, Mr. Heggade is Member board of management of Gammon India Limited and heads its Special Bridges vertical in addition to Engineering design management. He is a member of various national and international institutions. He is a recipient of IABSE-PRIZE-2002 & is a Fellow of National Academy of Engineers (India)

environment/sustainability consciousness and demand in power plants in India is likely to witness sudden spurt of cooling towers in the imminent future. Among the cooling towers, Natural Draft Cooling Towers do not need any external cooling devices driven by prime movers, and hence are able to balance environmental factors and operating costs of the power plants. The hyperbolic profile provides the most beneficial heat exchange efficiency besides structural stability. The very first such construction in India in 1934 measuring 34.0 m dia and 38.0 m height has now crystallized in structures 202 m high - huge ones by any standard (Fig. 1). The structural design of this highly complicated hyperbolic shell has been perfected over the last 70 years along with the sophisticated constructional systems for realizing in space the three dimensional accuracy of the thin shell of minimum thickness 160 mm. The internal pre stressed louvers for the heat exchange stack are the only ones of its type in the world and have been adopted in India to a large extent for NDCTs.

Keywords: NDCT, Hyperbolic geometry, Raker columns, Wind Loading, Enhancement factor, Quasistatic, Frequency, Dynamic.

Introduction

The cooling towers are extensively employed as recycled and reused water resource for condensing steam in order to create effective heat sink behind the turbines in power plants. The present alarming trends of depleting water resource, ever accentuating

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Fig. 1 : Evolution of NDCTs in India

Volume 46 │ Number 3 │ September 2016

Structural Arrangement

The tower consists essentially of an outside hyperbolic shell of reinforced concrete, the principal function of which is to create draught of air in a similar way to a chimney. The other water-cooling and collecting components of the tower such as internal grillage structure, pond floor, fill; water distribution system is structurally independent of hyperbolic shell and is housed inside the tower. Shell diameter at the base and throat, shell height and air inlet-height is governed by thermic design consideration. Fig. 2 shows typical cross section of the cooling tower.

appreciably less in proportion than that of an egg shell, and sensitiveness to horizontal forces. These towers are one of the largest civil engineering structures where wind forms the major applied loading.

Analysis of the Tower

The structural analysis and design of these peculiar structures should cater for: ●

Unconventional geometry/form.

●

Real boundary conditions

●

Stress, temperature and time dependent linear/ non linear behaviour of constituent materials including composites.

●

Space-time dependent loads including random excitation.

●

Different types of behaviour - linear, nonlinear dynamic stability - of structures to be examined.

The shape of the tower structure represents a hyperboloid of revolution (Fig. 3) and the element in the shell assumes the curvatures in two way of Negative Guaussian value {1/r1*1/r2}.

Fig. 2 : General Arrangement of NDCT

The hyperbolic tower in elevation stands tall above the circular basin sill in plan with the shell thickness reducing from soffit to throat from where again increases to top cornice. The shell is supported on numbers of pairs of circular raker (Inclined) columns that traverses the air inlet height. In India ‘V’ type of circular raker columns are in vogue among the prevailing variations elsewhere viz ‘X’ type, ‘A’ type and meridonial columns. The raker columns are housed in the pedestals which are integrated by RCC diaphragms to make a continuous pond wall of basin to retain the water. Ultimately, the load is transferred through pedestal supporting footing or pile cap to the founding stratum. The natural draught cooling towers are special structures, in view of the hyperbolic shape and large size combined with very small shell thickness, 50

Volume 46 │ Number 3 │ September 2016

Fig. 3 Hyperboloid of Revolution with 2 Way Curvatures

Up to eighties it was assumed that for doubly curved shells, except for the edge zones, the bending stresses developed are negligible because of the large vertical loads. Thus the membrane analysis, which does not

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involve complex analytical process with bending, was considered to be adequate. As shown in the Fig. 4, the loads operate in 3 coordinate directions and the following general equations of equilibrium are considered to solve for the co-planar stresses.

Fig. 5 : Stress Resultants & Loads on Shall Element

The following general equations of equilibrium in terms of the above six quantities can be written: 1. 2. Fig. 4 : Force and Loads on shall Element

1. 2. 3. After having applied the necessary boundary conditions, the ‘Stress Function’ is induced to bring down the three variables , θ & , θ into one. Though the tower behaves predominantly as a Membrane structure, the wind tunnel studies carried out at various international laboratories revealed that the corrections are required near boundaries especially at the cornice and ring beam locations (where raker columns meet the shell) due to the edge perturbations. As the tower sizes increased beyond 120.0 m height, it was imperative to predict the bending behavior of the structure. If the shell element is subjected to the external loadings in three directions as shown in the Fig 5, in the shells of revolutions, the internal forces developed will be:

Where, However, since there are six unknowns, additional three equations are required to make the problem determinate which are obtained by compatibility relations between the strain and curvature parameters of the middle surface of the shell. 4. 5. 6.

Where,

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Volume 46 │ Number 3 │ September 2016

The above classical bending analysis by numerical method, apart from being very complex, does not take into consideration the soil structure interaction at the foundation levels, which adds further complexities.

The base of the natural draft cooling tower is supported on equi-spaced pairs of inclined columns to facilitate air intake. The raker of the shell supporting columns (Fig. 7) and also the angle of foundations is matched with the base of hyperbola, to subject the column and foundation maximum to axial forces, avoiding the kinks and consequential flexural stress formations. However, the column supports produce concentrated reactions along the bottom edge of the shell, while between the columns shell edge remains stress free.

Since there are many commercial soft wares available and the method can be used to any structure with complicated boundary conditions, the Finite Element Method is extensively used in the analysis of NDCTs, beyond eighties. Hill and Collin used this method for the first time for cooling towers. In this method, it is possible to include variation in thickness, arbitrary loading, the pre stressing effect at the edge beam and the idealization of edge beam (ring beam) as a part of the shell, which are difficult to include in other methods. Also for an accurate determination of frequencies of free vibration of cooling towers, the flexibility of shell supports and foundations should be included in the analysis. The bottom and top stiffening will also have a significant effect on the frequency of free vibration. Perhaps Finite Element Method is the only convenient tool when allowance must be made for such distinct structural features. The standard Finite element Model, which has been evolved over a period of time with the maximum experience of designing NDCTs in India is shown in Fig. 6. The discretization should encompass and provide for the following structural configuration and responses.

Fig. 7 : Schematic View of Raker Columns

These edge reactions in turn, produce stress concentrations and give rise to bending moments and transverse shears in the shell, which is otherwise predominantly membrane shell. Thus the accurate determination of stresses at the junction of columns and shell elements is of significant structural importance. The presence of columns adds further complications by destroying axi-symmetric nature of shell geometry, which would have been simple, idealization wise. Though the ‘X’ and ‘A’ types of raker columns were attempted in India in initial stages, ‘V’ columns are found to be more popular in resisting forces in circumferential direction as well as in facilitating ease of construction. The angle at the base lintel is generally fixed between 16 and 18 degrees in various cooling towers from structural optimization and construct-ability view.

Fig. 6 : Typical FEM Model for NDCT

Volume 46 │ Number 3 │ September 2016

The top thickness, i.e. the cornice of the tower and the portion of the soffit between the raker columns are specified to be free which shall yield zero stress resultants. It has been shown by ‘Abasing and Martin’

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by considering four different idealized boundary conditions that base conditions do not significantly affect stress resultant, except near the base and the fixed conditions give an upper bound solution. Also, since the base edge beam (ring beam) is nearly 1/8 to 1/10 of overall dimension of the tower, it is realistic to consider the base as fixed.

4. Loadings and Design of Elements The various load combinations studied before inferring the critical stress resultants/forces for the design are as below: ●

Dead load + seismic load for shell and fill structures

●

Dead load + construction load for shell structure only

●

Dead load + wind load

●

Dead load + seismic load + temp. Gradient + sun’s radiation

●

Dead load + wind load + temp. Gradient + sun’s radiation

In India, the NDCTs built up to 1987 have been designed for peak wind pressures of short duration by static method. The wind loading forms the major part of the external loading in the design of the shell and supporting components and generally governs the design. This has a large steady component and a significant random component because of the air turbulence. The latter dynamic components can be calculated in frequency domain by natural frequency analysis, which is established to be contributing 50% at the total peak responses. However, this involves very complex analysis both in meridonial and circumferential directions at different level, for tensile, compressive forces and bending moments as separate combinations. The above has necessitated the codal provisions to translate the external forces and structural responses to ‘quasi-static’ analysis by application of ‘Gust-factor’ (G). The gust factor depends on the natural frequency in the fundamental mode, with speed, terrain and size of structure and the peak response occurring in a time interval of 1-hour duration. Normally ‘G’ value fluctuates between 1.6 to 2.2 being in ascendancy with smaller tower height and rough terrain. This values more or less tally with IASS recommendations. However, the gust factors given in German VGB guidelines varies between 1.0 to 1.15 and in ACI-ASCE report no 334, the same is 1.0, though these are for peak wind pressures rather

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than mean hourly wind pressures considered in earlier cases. The design wind pressures in the meridonial direction depend upon factors related to probable life of structure, terrain, and local topography, size of the structure for both peak factor and gust factor methods. The combined effect of these factors deduced by multiplication is shown in Fig. 8. It may be interesting to note that up to 40.0 m height, the peak factor method attains criticality.

Fig. 8 : Wind Effect on Cooling Tower

The variation across a particular elevation is ascertained by normalizing the values of equal angle increments from the windward direction and is represented by Fourier Series, H =Summation (An cosNθ) The pressure coefficients ‘An’ as per IS:11504-1985 and BS: 4485-1996 which are used for NDCT is given in Table 1. Table 1 : Fourler Coefficients

The pressure distribution suggested by ‘IS’ is around the throat and generally the same is also used for all other elevations in India. However, the study at SERC Ghaziabad reveals that Indian standards estimate is at least conservative by 50% using actual pressure distribution at different levels. It shall be noted that the coefficients suggested by Zerna, Neimann, and

Volume 46 │ Number 3 │ September 2016

Sollenberger are based on full-scale measurements on towers with ribs. Further the coefficients recommended by Zerna and BS/IS include internal pressure of –0.5 and –0.4 respectively. The effect of internal negative pressure of 0.4 increases the circumferential compressive forces by around 40% and correspondingly decreases the circumferential tensile forces. However, it does not have any impact on meridonial forces. The Fig. 9 depicts the effect of Fourier series coefficients due to wind at foundation nodal points.

●

2.0 for towers 1.5 dia apart.

Generally in India a clear spacing of 0.5 times the base diameter is provided between cooling towers and the group effects are accounted for by enhancing the wind pressures between 10 to 40%. In some cases findings of the Indian Institute of Science, Bangalore through wind tunnel testing on a group of four and six cooling towers as given in Table 3 are also used for guidance in specifying enhancement factors. Table 3 : Enhancement Factors as per Indian Institute of Science, Bangalore

Fig. 9 : Circumferential Wind Distribution

When the NDCTs are grouped, the wind pressures are augmented because of the aerodynamic interference effects and generally on the basis of wind tunnel testing, the enhancement factors are considered in the design. The enhancement factors deduced on the basis of wind tunnel testing for some specific jobs are tabulated in Table 2. Table 2 : Enhancement Factors after Wind Tunnel

Niemann (Germany) has reported interference factors up to 2.8 depending upon the distance between boiler house building and wind direction, while the factor of 1.60 was recommended for the interaction between cooling towers. After extensive investigations Armit (UK) had suggested the following group effects. ●

1.0 for isolated tower

●

1.25 for typical station building

●

1.25 for towers, 7.0 m apart

Volume 46 │ Number 3 │ September 2016

In the latest BS:4485 the amplification factor includes dynamic resonant stress resultant N,r. This is combined with meridonial stress resultants by applying group correction factors to arrive at final amplification factors. It is observed that the magnitude of resonant stress resultant at the top of the tower not approaching to zero gives abnormal values of amplification values at top. The code does not give the variation of resonant stress resultant along the height of the tower but gives the different resonant stress resultant for bottom third, middle third and top third portion of the shell. This problem is overcome by by line fitting for resonant stress resultant for one of the NDCT recently. In fact Niemann concludes that the resonance response is small even for very high tower with small eigen frequencies below 1 Hz indicating that cooling towers are not generally sensitive to vibration. In their paper Abu-Sita, S.H et al recommend to neglect resonant component as the contribution of the same is less than 10% of total strains based on the aero-elastic model for 126.8m tall NDCT. Both these papers show that the total meridonial force including resonance component at top is zero. With the understanding that resonant stress resultant reduces along the height of the tower and using empirical constants (i.e. B factor) as given in BS 4485 a line can be fitted (Curve 1) to arrive

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at resonant stress resultant to work out amplification factors along the height of the tower.

Curve 1 : Resonant Force along the Ht for Amplification Factors

On the basis of wind tunnel experiments conducted by Dev and Fiddler, the IS and BS give simplified empirical formula for the critical wind pressure to cater for global buckling behavior of the tower. Since the formula does not account for the influences of bottom and top edge stiffening, vertical variation of wind pressure, curvature effect and aspect ratio of tower, the factor of safety used is as high as 5. However, the latest BS:4485-96 recommends local buckling criteria, which is a function of both meridonial and circumferential stress resultants at a particular elevation. This is called Buckling Stress State (BSS) approach and developed by Mungun of Ruhr University Bochum. After having deduced the critical stress resultants, the elements are designed for membrane forces and bending moments both meridonially and circumferentially, restricting the tensile forces in concrete to modulus of rupture, as has been the assumption in the state of the art analysis. The following observations highlight the general non-consensus among the international community pertaining to design for temperature loading. The German VGB guidelines specifies Ieff = ½ (I gross + I crack), while BS & IS only state that consideration shall be given in the analysis for the strain resulting from temperature gradient, without mentioning anything about stiffness criteria. IASS and also ACI are silent about the stiffness factors to be considered for temperature stresses. However, IS: 4998, which deals with calculations of thermal stresses, recommends the use of working stress method, using cracked section stiffness. Thus it is inevitable to take recourse into companion Indian Code IS: 4998 and use ‘Superposition’ method for calculating temperature stresses, though the ‘Combined Simultaneous’ method gives slightly larger stresses. However the latest

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BS:4485 deems to cater for thermal gradient loading by provision of minimum 0.3% steel in the shell. In towers with ‘V’ and ‘X’ type supports tangential to the shell thermal loadings caused by support restraints may be ignored. However, where supports are stiff in the direction normal to the shell, like in meridonial columns thermal loading effects should be critically investigated. The cracking is always considered to be adequately covered by the provision of minimum % age of steel as per codal specifications and in the eventuality crack widths need not be investigated. Unlike other stack like structures, in NDCTs, the maximum meridonial tension occurs at windward 0º and compression occurs at around 72º instead of leeward 180º. Thus the raker columns and pedestals have to be checked for maximum axial tension and compressive forces with the corresponding bending moments. Also it has to be ensured that the foundation is not uplifted under maximum axial tensile forces to limit the significant changes in the shell stresses that may occur under the lift off conditions.

Conclusion

NDCTs are very peculiar structure in built environment and call for in-depth knowledge of structural behaviour for design and construction aspects for realisation. The structural analysis has a significant influence on the accuracy, reliability and economy with which the results can be achieved. Therefore, while in-depth understanding of the various aspects of modelling and their effects is imperative for the designer, the realisation of this three dimensional structure in the space to the accuracy with in the tolerance limits is a challenge in it self for the constructor.

References 1.

Hill D.W., Collin G.K., “Stress and Deflection in Cooling Tower Shells due to Wind Loading” Bulletin of IASS, No. 35. Sept. 1968, pp 43-51.

Albasing, E.L., Martin D.W., “Bending and Membrane Equilibrium in Cooling Towers” Journal of Engineering Mechanics Division of ASCE, Vol. 93, June: 1967, pp 1-19.

IS:11504-1985, “Criteria for Structural Design of Reinforced Concrete Natural Draught Cooling Towers” - Indian Standards Institute, 1986, pp 15-18.

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V.N.Heggade, “Raichur Natural Draught Cooling Towers - Units 5 & 6” - Journal of Indian Concrete Institute, vol.1, July-Sept 2000, No.2 pp 7-18.

BS: 4485 Part 4: 1996 “Code of practice for Structural Design and Construction.

H.J.Neimann & M.Kasperski, The Assessment of Wind Load on Cooling Towers, Proc. International Symposium on Wind Loads on Structures, 1990, India.

Sollenberger.N.J, Scanlan.R.H and Billington .D.P, Wind Loading and Response on Cooling Towers, Jr of Structural Engineering, ASCE, 106,1980.

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Zerna. W, Impulses on the Research on Development of Large Cooling Towers, 2nd International Symposium, Ruhr University, Bochum, Sept 1994. 9. Prabhakar.N, Structural design aspects of hyperbolic cooling towers, New Delhi, 18-20, January 1990. 10. V.N. Heggade, ‘Proceedings of the Fifth International Symposium on Natural Draught Cooling Towers,20-22 May 2004, Istanbul, Turkey. 11. Heggade V N, ‘The Evolution of Design and Construction Aspects of Natural Draught Cooling Towers in India’, SEWC-2007, 2-7th November, Bangalore, India.

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TALL BUILDINGS – A REVIEW OF FORMS, SYSTEMS AND ANALYSIS Subhash Mehrotra, born in 1949 received his M. Tech. (Structural Engineering) degree from I.I.T. Delhi. He is Governing Council Member of Structural Engineers World Congress (SEWC) India and Engineering Council of India (ECI). He is Chairperson of Association of Consulting Civil Engineers (India) (ACCE(I)) Delhi Center. He is former Executive Committee Member and Chairman of Membership Committee International Federation of Consulting Engineers (FIDIC), Former President of Indian Association of Structural Engineers (IAStructE), former President of Consulting Engineers Association of India (CEAI).

Subhash Mehrotra

Managing Director, Mehro Consultants New Delhi, India scmehrotra@mehroconsultants.com

Introduction

Though definitions for tallness varies, from the point of view of structural engineer, it is the one predominately affected by lateral forces such as wind and earthquake actions. The rapid growth of urban population, pressure on limited space in cities for business and residential

development and the need to avoid urban sprawl, tall buildings provide a feasible solution. The large number of tall buildings construction during last few decades has been due to development of high strength materials, new design concepts, new structural systems and improved construction methods.

Table 1 : Tall Buildings More Than 100 Storeys S. No. 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17

Building Name Burj Dubai Chicago Spire Pentominium Russia Tower China 117 Tower Doha Convention Center Tower Sears Tower Burj Al Alam Busan Lotte Tower International Commerce Centre Pearl River New City West Tower Empire State Building Taipei 101 Shanghai World Financial Center Princess Tower Marina 101 John Hancock Center

City

Year

Stories

Dubai Chicago Dubai Moscow Tianjin Doha Chicago Dubai Busan Hong Kong Guangzhou New York Taipei Shanghai Dubai Dubai Chicago

2009 2011 2012 2012 2013 2012 1974 2011 2013 2010 2009 1931 2004 2008 2009 2010 1969

160+ 150 120 118 117 112 110 108 107 106 104 102 101 101 101 101 100

Structure

The architectural features and functional requirement of commercial and residential buildings differ considerably and the requirement to provide

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Height M 800+ 609 618 612 570 551 442 501 510 484 438 381 509 492 414 412 344

adequately stiff and strong structures accommodating the various features lead to radical development in structural framing involving braced frames, frame tube and wallframe systems at the same time ensuring speed of erection as a vital factor. Volume 46 │ Number 3 │ September 2016

Apart from function of all structural elements to resist gravity loads, the Structural elements are further required to resist the highly probable second function due to loads caused by wind and earthquake. Taller the building the escalation in cost for system to resist wind and earthquake will be very large. Provision of adequate lateral stiffness is a major consideration in the design of tall buildings not only for critical load combination but also to include augmented moments due to P – Δ effect and stresses caused by creep, shrinkage or temperature in elements and materials.

Eccentrically Braced Frame

At least one end of each brace must be eccentrically connected to the member.

Structural Form

Major consideration affecting the structural form is the function of the building. Large column free floor spaces in offices may require arrangement of vertical components around the perimeter of the plan and around elevators stairs and service shafts, at the same time providing adequate lateral stiffness against wind and earthquake forces. Various systems so evolved are frame- tubes, tube – in - tube, bundled – tube with or without frames having diagonal bracing. In the case of residential building or hotels, where accommodation is subdivided permanently and repetitively from floor to floor, interacting shear wall or core and frames can be a suitable system. a)

Braced Frames

Braced frames develop their resistance to lateral forces by the bracing action of diagonal members. The braces induce forces in the associated beams and columns so that all work together like a truss, with all members subjected to stresses that are primarily axial.

Fig. 1.

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Fig. 2

Large- Scale Braced Frame

Frames shall be braced diagonally to provide enhanced lateral stiffness, so that together with the girders they form the web of a vertical truss and column acting as chords. Such bracing can be storey wise or large-scale by coupling 2 or 3 storeys.

Fig. 3. : Large-Scale Braced Frame

Infilled – Frame Structures

Framing of reinforced concrete or steel, having infilled panels of brickwork behaves similarly as braced frames but is ineffective under transverse vibration, as the panels can be dislodged.

Fig. 4. : Infilled Frame

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Wall- Frame Structure

Bundled- Tube

Coupled wall structures with beams effectively connected in the planes, act as a composite cantilever bending about the common centroidal axis of walls resulting in mobilising much greater lateral stiffness.

In bundled tube, structural form consists of parallel rigid frames in each orthogonal direction, to form bundled tubes.

Fig. 5.

The system is appropriate for buildings in the 40 to 60 storeys. Frame- Tube structure consisting of very stiff moment resisting frames that form a tube around the perimeter of the building with closely spaced columns connected by deep spandrel girders carries all the lateral loads, while gravity loading is shared between the tube and interior columns and walls. This system has been used for buildings ranging from 40 to 100 storeys.

Tube – in- Tube

A variation of the framed tube is Tube – in – Tube, in which the external tube and internal service core structure jointly resist both gravity and lateral loading.

Fig. 6.

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Fig. 7. : (A-B) Bundled - Tube for 50To 100 Storeys

Steel Braced Tube & Concrete Braced Tube

In Braced – Tube structures efficiency of the framed tube is increased further by adding diagonals bracing in the exterior tube to allow greater spacing between columns and improve its efficiency.

Fig. 8.

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(A) Outrigger- Braced Structure & (B) Outrigger – Braced under Load

Another efficient structural form consists of a central core consisting of braced frames or shear walls, with horizontal cantilever outrigger trusses or girders connecting the core to outer columns. For horizontal loads the vertical plane rotation of the core are Restrained by the outriggers, through development of compression and tension in outrigger columns.

These structural arrangements are suitable for a prismatically shaped towers. However, a judicious combination of such single identifiable systems will be required in the same building, which has to cater to different functional requirements.

Comfort Criteria

The most common causes of vibration in structures are wind, earthquakes, machinery, nearby industrial plants, and the various types of transportation. The motions resulting from these causes can vary greatly in duration and intensity, and there are a variety of mechanisms by which the apparent motion may be exaggerated. The perception of building movement depends largely on the degree of stimulation of the body’s central nervous system. The sensitive balance sensors within the inner ears play a crucial role in allowing both linear and angular accelerations to be sensed. Human response to building vibration is influenced by many factors, such as the movement of suspended objects, and the noise due to turbulent wind or fretting between building components.

Fig. 9

Space Structure

In space structure the primary load – resisting system consists essentially of 3 dimensional triangulated frames. However, they are geometrically complex in assembly.

Fig. 10. : Space Structure

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Though there are no accepted International standards for comfort criteria, it is generally agreed that acceleration is the predominant parameter in determining the nature of human response to vibration. Table 2 : Showing Human Perception Levels Range

Acceleration (m/sec2)

Effect

< 0.05

0.05 – 0.10

Sensitive people can perceive motions; hanging objects may move slightly.

0.1 – 0.25

Majority of people will perceive motion; level of motion may affect desk work; long – tern exposure may produce motion sickness.

0.25 – 0.4

Desk work becomes difficult or almost impossible; ambulation still possible.

0.4 – 0.5

People strongly perceive motion; difficult to walk naturally; standing people may lose balance.

0.5 – 0.6

Most People cannot tolerate motion and are unable to walk naturally.

0.6 – 0.7

People cannot walk or tolerate motion.

> 0.85

Objects begin to fall and people may be injured.

Humans cannot perceive motion

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Braced Structure with Friction Damper

In modern building, avoidance of structural collapse alone is not enough. The cost of nonstructural components is much higher than the cost of the structure itself and must be protected. The alternate solution is to establish performance based design criteria and dissipate seismic energy mechanically. In typical structure without dampers, the inherent damping is merely 2-5% of critical. With the introduction of supplemental damping of 10-15% of critical, the forces and deformations on the structure can be significantly reduced. Multiplying Factors for obtaining Values for Other Damping (IS 1893 (Part 1) : 2002) Table No. 3 Table 3 Damping Percent Factors

Fig. 12. : Friction Damper in Chevron Bracing 10

3.20 1.40 1.00 0.90 0.80 0.70 0.60 0.55 0.50

Similar to automobiles, the motion of vibrating building can be slowed down by dissipating seismic energy in friction. One of the Friction type dampers is Pall Friction damper. The friction dampers are designed not to slip during wind. During a major earthquake, they slip prior to yielding of structural members. In general, the lower bound is about 130% of wind shear.

Fig. 13. : Unitech Garden Galleria, Gurgaon

Fig. 11. : Typical Bay with Friction Damper in Diagonal Bracing

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Fig. 14. : Unitech Garden Galleria, Gurgaon

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effective for beam and columns respectively. Major structural analysis programs offer a variety of finite elements for structural modelling of slabs, walls, coupling beam etc. However, choice of the elements should be appropriate to reflect the principal action of the structural elements.

Fig. 15. : Central Plates for In-Line Friction Dampers

6. Loading Unlike low rise buildings, cumulative effect of loading from large number of floors produces high order of gravity loading on columns and walls. In addition wind loading and inertial loading due to earthquake will considerably influence in the choice of structural form, cost and construction methodology in the case of tall buildings, as the buildings dynamic response plays a large part in the choice of the system. Assessment of earthquake and wind loading may at times require experimental procedures such as shake table and wind tunnel tests. Construction load, though, temporary has also to be duly considered, as they are greater in intensity than normal loading.

Modelling

A preliminarily model should represent fairly well the principal modes of actions and interaction of major structural elements. With the advent of computers a 3-D dimensional analysis of fully detailed model of the structure is now possible. However, a judicial choice and combination of different structural systems satisfying the functional requirement will be required for detailed modelling. Floor slabs are assumed to be rigid in plane, so that at all floor levels horizontal displacements of vertical elements can be defined in terms of a rigid body rotation and two horizontal translations. By reducing gross moment of inertia, effect of cracking in concrete members can be considered. Some International Codes suggest 35% and 70% of gross moment of inertia as 62

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Integrated design of Super Tall Buildings

The basic difference in design for earthquake and wind forces on buildings is that the earthquake generally governs the design of building having lower time periods i.e. approximately 0.1 sec-1.5 sec, as maximum response of a structure to a specified ground motions is strongly influenced by its fundamental mode of vibration. Earthquakes are of short duration and impart high energy at lower time periods. When the structure becomes more flexible its time period is increased and design will be governed by wind forces, as wind carries higher energy at higher altitudes, have longer time periods of 1.5 secs to 10 secs and induces oscillations both along and across the direction of flow. Buildings have been designed for 100 storeys i.e. approx. 300 m high having structural arrangement consisting of shear wall, rigid frame, frame tubes, tube-in-tube, braced tube and outrigger structural system. But when the structure is very tall, 200 storeys and above, the structural response due to flexibility of the system has a greater impact on comfort level of occupants in addition to its influence on structural integrity. Average human perception threshold to translational vibration ranges from 0.6% of g at 0.1Hz to 0.3%of g at 0.25Hz. The vibration level above 1.2% of g is annoying and above 4% causes difficulty in walking. The Council on Tall Buildings and Urban Habitat (CTBUH) recommends 10 year peak resultant acceleration of 1.0-1.5% g for residential building, 1.5-2.0% g for hotels and 2.0-2.5% g for office building. Generally more stringent requirements are suggested for residential buildings, which would have continuous occupancy in comparison to office buildings usually occupied only part of the time and whose occupants have the option of leaving the building in advance of a storm. Therefore, the structural response at high wind speed is be controlled specially the acceleration at various frequencies of oscillation of structures.

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Different type of studies which can be done to obtain the dynamic response of the structure subject to wind loading. a.

Wind Tunnel Test:

In the design of super tall buildings, when response due to wind load is an issue, it must be based on the results of wind tunnel test. In wind tunnel the scaled model of structures are subjected to scaled atmospheric wind in a control laboratory setup. The sensors can measure various quantities of interest such as pressure, shear, moment etc.

Analytical Study: With analytical studies the structures are modeled in structural dynamic sense. For any analytical study a few parameters are determined through wind tunnel tests.

Computational Fluid Dynamics: The structures are modeled in structural dynamic sense and the wind flow is modeled using basic fluid dynamic equations such as continuity energy etc. Thereafter a specific turbulence model is used and the equations are solved using numerical techniques and the responses are obtained.

Increasing the damping

By increasing the damping the amplitude of motion and the duration of vibration are reduced. Therefore the floor acceleration is reduced and this results in increase in human comfort level.

There are various types of dampers: a)

Passive damper

Dissipate the wind or earthquake induced energy to the structural system by movement of the building and their mechanical properties are pre-defined. One benefit of these passive dampers is that they do not need a source of power to operate and their cost is relatively low since they are not accompanied by electronic devices or mechanical actuators. Example is Friction Damper.

Active Dampers

Have structural components.

Semi Active Damper

In semi active dampers, the adjustment in the mechanical properties of the device can be achieved. The most commonly used semi-active damper is a passive viscous damper with the external path for the fluid with a control valve.

The overhead water tanks in tall buildings also acts as a damper during wind. Since the displacement at top level of the building is high and also the duration of displacement is much higher in wind load case in comparison to earthquake load case, the water tanks acts in the opposite direction and reduces the displacement.

Following are the remedial measure by which the response to wind forces can be controlled:a)

Aerodynamic modifications

Typically the geometry can be modified to reduce the wind induced response. For instance rounded corners or step corners on sharp edge cross sections can reduce the intensity of vortex shedding and cross wind loading. Since vortex shedding and galloping are caused by the shape of the structure, modifications to the shape have the potential to eliminate the root cause of aerodynamic stability problems.

Structural Modifications

Increase natural frequency of the structure by increasing the stiffness. This will reduce the response of the structure due to the winds of low frequency.

control

with

several

Acknowledgement 1.

“Tall Building Structures: Analysis and Design” By – Bryan Stafford Smith& Alex Coull.

Hand Book of “Concrete Engineering” Edited by Mark Fintel.

Increasing mass

It results in decrease in floor acceleration and thereby reduces the susceptibility to aerodynamic instability and motion.

“Analysis of Skeletal Structures” By – Seetharamulu Kaveti.

“Best Tall Buildings - 2008” –By –Council on Tall Buildings and Urban Habitat.

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Volume 46 │ Number 3 │ September 2016

Damage Identifications in a Through Type Steel Truss Bridge Model

Suresh Kumar WALIA Research Scholar Dept of Civil Engg, NIT Hamirpur sureshkumarwalia@ gmail.com Suresh Kumar Walia, born 1967, received his civil engineering degree from the NIT, Hamirpur. He has been working in HPPWD since 1993, and is presently Executive Engineer. His main area of research involves damage detection in bridges.

Hemant Kumar VINAYAK Assistant Professor Dept. of Civil Engg., NIT Hamirpur hemant.vinayak@ gmail.com Hemant Kumar Vinayak, born 1975, received his civil engineering degree from the PTU, Jalandhar. He was a research scholar at IIT Roorkee before becoming Assistant Professor at NIT Hamirpur. His main area of research is related to Damage detection, Structural Health Monitoring and Retrofitting of Structures.

Summary This paper presents a study of damage detections and localization methodologies on a through type truss bridge model at laboratory-scale. The parameters worked upon are frequency, mode shape, mode shape curvature and modal strain energy change ratio. The objective of this work is to understand behavioural concept of damage identification methodology for a bridge using different parametric variations. The Finite Element model was created to generate baseline information for damage detection. The comparison of the modal parameter and its derivatives with respect to both the trusses before and after damage has been presented. The damage has been introduced by removing the member stiffness between the nodal points. It has been concluded that upstream and downstream truss of the bridge can behave in different manner even if designed for the same loading due to constructional variations. It is observed that modal derivatives especially Modal strain energy change ratio is an efficient approach useful for damage localization. The identification and location of the

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

Ashok KUMAR

Raman PARTI

Professor Dept. of Earthquake Engg., IIT Roorkee akmeqfeq@iitr.ac.in

Professor Dept. of Civil Engg., NIT Hamirpur ramanp@yahoo.com

Ashok Kumar, born 1951, received his civil engineering degree from BITS Pilani. Presently as professor in IIT Roorkee, his main area of research is related to Earthquake Early warning system, signal processing techniques.

Raman Parti, born 1961, received his civil engineering degree from PEC Chandigarh. He has been working in NIT Hamirpur since 1989. His main area of research is Transportation Engineering.

damage through lower modes can be determined with higher level of confidence than with respect to higher modes of vibration. Keywords: steel truss bridge model; frequency; mode shape; mode shape curvature; modal strain energy change ratio.

Introduction

To assess and localize the damage in the civil engineering structures using dynamic parameters between the damage and intact states fast and inexpensive method is required [1]. The developments in the sensor technologies and improvement in the computing and networking capabilities have led to in-situ sensor networks for modern bridge monitoring practices [2]. Winds and traffic being natural excitations methods have led to ambient vibration testing methodologies thus surpassing challenge of performing dynamic tests on bridges [3], [4]. Ambient vibration testing and modal parameter identification have been successfully applied to various types of bridges [5][3]. However for the short span bridges

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an initial impetuous is required so that the bridge is set into vibration for recording and the initiation of vibration with vehicle has been carried out under various researches [6]. Since the dynamic parameters such as natural frequencies, damping ratio and mode shapes are directly associated with stiffness of the structure and any decrease in the natural frequency and change in mode shape points out to the loss of stiffness [7], [8]. The accuracy of modal parameter is always in question because these parameters depend on various factors of experimentation leading to further evaluation of their derivatives [9]. The change in the structural system are characterized by changes in the Eigen parameters that are natural frequency, damping values and the associated mode shapes [10][11]. The signal to noise ratio is associated with accuracy of extraction of frequency and mode shape [12]. Hence for an accurate and complete extraction of modal parameters the excitation of the structure to appreciable amplitude is important [13]. Moreover improper measurements lead to incomplete modal extraction even with the algorithm of established reliability [14]. Initially Cawley and Adams [5] used an incomplete set of measured natural frequencies to identify the location and provide a rough damage estimates. Early researches focused on indicating the existence and location of damage [15]. A number of structural identification techniques based on modal data have been proposed in the literature. Most of them are formulated as an optimality criterion, where the stiffness distribution of a chosen reference configuration of the system is iteratively updated so that the differences between the analytical and measured values of the modal parameters of the first vibrating modes are minimized [16]. Modal strain energy change method is simple and robust in locating single or multiple damages in a structure and is sensitive to damage [17] and helps to indicate the damage from the measured mode shape [18]. For the location of damage in the structure, the strain energy method appears to be more accurate as compared to other methods [19]. The primary objective of the present paper is to develop a method that can effectively localize the damaged members, as well as accurately estimate their severities. Since frequency measurements can be easily acquired and are more reliable, the approach could provide an inexpensive structural assessment technique. The mode shapes and modal derivatives are able to effectively identify the geometric locations of the damaged members requiring only a small number of modal frequencies identified from the damaged structure. To demonstrate

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the effectiveness of the method, modal analysis was carried out on a through type steel bridge model structure.

Description of the Steel Truss Bridge Model

The model for the through type steel bridge was constructed in the structural laboratory. The bridge model represents the bridges being constructed in the field as through type steel truss bridge and type of members are as in field. This bridge model is not to the scale and has been constructed in the manner to facilitate experimental and modal parameter extraction. This single span 6m steel truss bridge is divided into 10 equidistant zones (Fig. 1). The deck of the bridge consists of 10 mm thick steel plate. This Bridge model is a simple support bridge with a hinge support at one end and roller support at other end. The vertical and diagonal members of the bridge model are made up of 20 mm x 3mm steel plate member. The top and bottom horizontal members are made of angle section 30 x 30 x 3 mm. The members of the trusses have been joined with welded connections. The carriageway is 300 mm. The analytical model of the bridge was generated in SAP2000[20].

Fig. 1: Lab Model for through type Steel Bridge

The members are assigned specifically as built crosssection to achieve the analytical model as accurate and close to the actual bridge (Fig. 2). The results obtained from analytical model are included as base line for considering different modes.

Fig. 2 : Analytical Model on SAP2000 of the Bridge

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

Experimentation

The experiment on the steel truss bridge was carried out to detect the damage by sequentially cutting the truss members. As any variation in the expected behavior of the trusses can be due to a particular steel member being ineffective in taking the stress and is transferring the load to other members with partial rigidity. Impact testing utilizes a sudden power by an object such as hammer to excite the bridge model [21]. The weights of the impact hammer can be adjusted to produce different forcing levels to be applied to the structure. In the present work the forced and ambient vibrations were measured with hammering with hammer of 5 kg. In the present study, the bridge is of single lane and the hammer was struck at three points at 2/3, 1/2 and 1/3 of the span. The impact has not been applied simultaneously and the impact force was not measured as purpose of the impact was to excite the structure for free vibration. In the present study the mass of the hammer has been kept same and bridge was excited by hammering with same intensity so that the comparative values will not be affected. The modal parameters such as the natural frequencies and mode shapes have been extracted through the recorded response. The variation in modal parameters with damages since the response has been obtained through multiple number of sets of the excitation for a particular speed. This multiple number of excitations has included the variation due to the bridge and vehicle interaction. The parameters for the sensor location, input motion, sampling frequency, are taken as same for the various setups so that the records obtained are not affected by the other parameters. In the present study it has been considered that the various factors such as mass of hammer, temperature, surface roughness etc. have similar effect on whole of bridge model and not specific to any particular mode. Though these factors will have an effect on the individual values of the results but get nullified when the relative variations are considered. The details of the test specifications are as below. 3.1 Sensor and Data Acquisition System The data acquisition system and laptop was installed at one end of the model bridge. The sensors used were dual channel MEMS accelerometers having natural frequency of 150 Hz. The results obtained from cabled sensors are comparable with wireless sensors and thus cabled sensors were used in the present study [22]. Accelerometers provide voltage output, which is proportional to the acceleration of

Volume 46 │ Number 3 │ September 2016

the points of placement. This analog acceleration time history is then fed to data acquisition system. The data acquisition system first conditions the analog signal to its requirement through signal conditioning amplifier and then the conditioned analog signal is fed to AD Converter where digitization at prescribed sampling rate takes place. This digital data is then stored in hard disks of the laptop in ASCII format. The basic functions of data acquisition system are achieved with the in built interactive software and hardware system. These systems can be standalone or coupled to a computer and have the facility of acquiring simultaneously multiple channels of data from various sensors. The data acquisition system acquired for this particular case study for generating output of the response of the structure had the desired specification to generate the output which has sufficient information to extract the modal frequencies after processing accurately. The data acquisition system DAQ used for this study is KI-4100-A-8-500 of KAPTL instrumentation make is an advanced Micro controller Based system which is designed for high speed precision simultaneous measurement of physical parameters. This 12V battery operated DAQ of eight channel with analog to digital converter of 16 bit, precision of 2.5 to 5 volts, with short circuit protection, zero balancing and simultaneous sampling up to 500sps is suitable for MEMS based accelerometers. The Data acquisition system has programmable filter of up to 500 hz, Drift - 2µV/°C, Accuracy - 0.05%, Signal Conditioner Skew rate - 2.4 x 10-6 V/sec, resolution - 0.001 g, noise level on high speed - ± 0.002 g. The accelerometer and data acquisition system interface is through RS – 232C/ USB port. The self-calibrated MEMS technology based accelerometer used is of make Freescale, Type – Triaixial, Frequency response upto 150 hz, sensitivity - 1volt/g. The schematic diagram of Setup of Sensor, Data Acquisition System with Data Storage is as shown in Fig. 3. 3.2 Sensor location To obtain the distributed response of the structure, sensors were placed at the intersection points of the vertical and diagonal members in the vertical direction of motion on both the trusses of the bridge. In a single set, four sensors were used for measurement. Thus, vertical acceleration time histories at 18 locations were measured in six sets. The sensors were not placed at supports (four nodal points) as complete fixity is considered at the supports. Each sensor location has been identified with unique nodal coordinate number.

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Fig. 3 : Block Diagram of Setup of Sensor, Data Acquisition System with Data Storage

Fig. 4 Shows the placement of the sensors position on the bridge with the typical green coloured arrows

indicating movable sensor location and blue colored arrow as reference sensor location.

Fig. 4 : Typical Sensor Location of One of Setup on the Bridge

The nodes have been designated as 1,2…11 of one truss and 12,13…,22 on the other truss of the model bridge. The accelerometer at node 6has been used as reference sensor. The vibration measurements of the nodes in the model bridge were carried out along the vertical direction. The direction of the installed acceleration pick up was accounted for during the analysis of the data. 3.3 Processing of Recorded Data In case of analytical modal analysis the determination of the modal parameter is independent of the applied dynamic force as a computational parameter depends on stiffness, mass and damping. However in case of experimental modal analysis carried out through Frequency domain decomposition algorithm, the modal parameters are dependent only on the signature output of the nodes of the structural model. The digitized output signal generated by the

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data acquisition system is given as an input to the algorithm. Throughout the experiment, the sampling rate for acquiring the signal was fixed at 200 samples per second. This sampling rate is sufficient to provide information regarding modal frequencies up to 100 Hz (Nyquist frequency) and four modes can be covered in this range. The Frequency domain decomposition algorithm extracted frequencies and the associated normalized mode shapes for each set of the time history. The typical input in the ARTeMIS software is to assign nodal coordinates in the structure, members connecting the nodes, details of the files containing recorded data, which in this case is acceleration time history, direction of the installed sensor during testing which give the information to the software that whether record generated by the sensor is in phase or 180 degree out of phase with the direction of the motion of the structure and master slave node condition considered in the model. The master slave Volume 46 │ Number 3 │ September 2016

condition involves the dependency of the movement of the non-measured node on the measured nodes.

experimental cases as the different damage cases has been evaluated with respect to undamaged state.

4.1 Frequency Change The damage from dynamic analysis can be almost precisely predicted and detected using natural frequencies and mode shapes [23]. Table 1 and Table 2 shows the comparative frequency of the bridge and the effect of damage on the frequency of the bridge for the four experimental cases (Fig. 4) generated in the model bridge.

Experimental Studies

The damage in different components of the structure leads to corresponding change in the dynamic behavior. Although steel is a homogenous material but various component of the structure i.e. structural member, connection plates, rivets, all made of steel undergo different mode of vibration when subjected to dynamic forces. Taking these aspects into consideration work was directed in the direction such that introduction of damages is studied on the structure by evaluating various modal parameter. The study has been carried out considering first four modes as higher modes were not feasible to be extracted and all modes are vertical modes. The modal parameters that are the frequency and mode shape have evaluated for analytical as well as experimental cases while mode shape curvature and modal strain energy change ratio only for different

Case 1 (E0) - No damage, Case 2 (E1) - One damage, inclined member between nodes 13 and 14 removed Case 3 (E2) - Double Damage, inclined member Between nodes 13 and 14, 2 and 3 ineffective. Case 4 (E3) -Triple Damage, inclined member between nodes 13 and 14, 2 and 3, 8 and 9 ineffective.

Fig. 4 : Damaged Member in the Mode Bridge (a) Single Damage (b) Double Damage (c) Third Damage

Table 1 : Modal Frequencies of Bridge Model for Different Cases Freq. (Hz)

1st

8.80

8.88

8.71

7.97

7.23

2nd

16.55

16.42

15.89

15.60

15.48

3rd

23.53

21.25

20.89

20.40

19.29

4th

39.38

39.17

38.95

36.26

35.59

Table 2 : Modal Frequencies Change with Damage of Bridge Model for Different Experimental Cases Mode

Change from

Effect Change from

E0 to E1(%)

E1 to E2 (%)

E2 to E3(%)

E0 to E1

E1 to E2

E2 to E3

1st

1.91

8.50

9.28

Intermediate

Most

2nd

3.23

1.83

0.77

Most

Least

3rd

1.69

2.35

5.44

Intermediate

4th

0.56

6.91

1.84

Least

Intermediate

The frequency decreased with the increase in the number of members being damaged. With the increase in damage from single to double and further to triple

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

damage, decrease in the frequency increased in all the modes. The change in the extracted frequency for single, double damage and triple varied from 0.56% to

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3.23%, 1.83% to 8.5%, 0.77% to 9.28% respectively, with respect to previous state of the bridge indicating change in the state of the bridge. For each state of the bridge one of the four modes turned out to be least affected while the other mode is most affected. The lowest modes turned out to be the most affected.

of linear as the plot has been generated with the experimental values and interpolation is not carried out with any algorithm. The difference in the mode shape of the downstream and upstream trusses of the bridge is studied for different cases generated due to induced member damage (Fig. 5). The variations of the amplification in downstream truss and upstream truss are dissimilar at different nodes with respect to the analytical model and undamaged state indicating variable behavior of both upstream and down trusses with respect to each other due to constructional variations.

4.2 Mode Shapes The mode shape is a function of the physical properties of the structure and changes in the physical properties causes noticeable changes in the mode shape. The mode shapes are not joined by cubic spline instead

(a)

(b)

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Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

(c)

(d) Fig. 5 : Mode Shape of Downstream and Upstream Truss of Model Bridge (a) 1st (b) 2nd (c) 3rd (d) 4th

The variation of mode shapes at nodes 13 and 14, 2 and 3, 8 and 9 indicate undesired behavior reflecting damages at these points. Although the modal parameter can detected the damage however unclear damage detection can be attributed to the imperfection in the model fabrication leading to some visible variations in the other nodes of the bridge. Further the higher modes have been highly un-predictive in the locating the damage. 4.3 Mode Shape Curvature Amid various techniques for structural damage detection comprising modal parameters using forced, free or ambient vibrations, mode shape curvature can be used for localization of damage in the structure [24] [25]. Damage is considered as a localized reduction in structural stiffness. Mode shape curvature can be used to identify damage instead of displacement mode 70

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

shapes. Mode shape curvature directly related with binding strain. Decreasing the flexural stiffness of the beam subsequently causes an increase in curvature. Curvature or bending strain (BS) is calculated by the following equations.

BS =

Eqn. 1

Where Y is the mode shape component at two adjacent elements and h is length of element. The curvature based method by central difference in approximation to mode shapes is given by Equation 2. Where,

Eqn. 2

is the mode shape component at two adjacent elements and h is length of element. A local change of stiffness arises from local change of mode shapes

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curvature and hence in the mode shapes [8]. For ith mode shape, the is Modal Curvature Difference calculated as given in Eqn. 3. ∆Φ″ = Φ″id − Φ″i0 Where,

Eqn. 3

d and 0 denotes the damaged and intact state and Φ″ is the Mode shape Curvature. Although the modal curvature difference between the experimental cases of the model bridge with single, double and

triple damage were obtained at all nodes, however the difference was more prominent in the location of the induced damage in the bridge. The single damage induced in the model bridge is clearly indicated in the 1st mode of the downstream truss (Fig. 6a) although at the higher mode (Fig. 6f) the higher MCD values are obtained. Thus emphasizing the fact that values obtained in the lower modes are to be relied upon. Similarly for the double damage nodal point.

(a)

(b)

(c)

(d)

(e)

(f)

(g)

(h) st

Fig. 6 : Modal Curvature Difference for the Downstream and Upstream Truss of (a) 1 D/S (b) 1st U/S (c) 2nd D/S (d) 2nd U/S (e) 3rd D/S (f) 3rd U/S (f) 4th D/S (g) 4th U/S

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Volume 46 │ Number 3 │ September 2016

change is the ratio of absolute difference of modal strain energy of damaged and undamaged element to the modal strain energy of undamaged element. [Eqn. 4].

4.4 Modal Strain Energy Change The mode shape and mode shape curvature reveal the damages introduced in the bridge Model. However both the mode shape and mode shape curvature involves the computation of relative variation of the respective nodal parameter along the bridge model. The elements of the bridge across the span are subjected to various levels of strain in different modes of vibration due to the corresponding applied strain energy. The strain energy changes with the change in the stiffness of the joint or the stiffness of the element as the stiffness varies with the damage in the bridge. The changes can be judged with the changes in the modal strain energy due to damage in the structure. The elemental modal strain energy is the product of the elemental stiffness matrix and second power of its mode shape component [19]. The modal strain energy

MSECR =

Eqn. 4

Where the undamaged and damaged modal strain energy is the product of the elemental stiffness matrix [Kj] and second power of its mode shape component [ÎŚi] of undamaged and damaged state respectively.

MSEu =

MSEd =

Eqn. 5a

Eqn. 5b

Hence Eq 4 can be rewritten as

MSECR =

Eqn. 6

(a)

(b)

Volume 46 â&#x201D;&#x201A; Number 3 â&#x201D;&#x201A; September 2016

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In the present study for evaluating MSECR, Modal strain energy for the damaged and intact state are compared for both the upstream and downstream damages truss as shown in Fig. 7. The benefit of using the Eq. 6 is that the variation in both the truss could be judged considering strain energy as the parameter. Also this equation did not involve any previous knowledge of elemental stiffness. Since the method involves modal shape coefficient hence this parameter will not be applicable to determine the variation in the node of the structure that does not vibrate. The Modal Strain Energy Change showed that modal strain energy change ratio is maximum at node 13 as the damage was introduced between node 13 and 14 as single damage. MSECR in the U/S truss at node 2 and 3 is maximum because the node points are just opposite to nodes 13 and 14 which was damaged. For the case of Double Damage the modal strain energy change is maximum at nodes 2, 3 and 13, 14 for U/S and D/S truss. For triple damage case when the damage was further introduced in the member 8 and 9 the MSECR values at the nodes are more than other nodes indicating the damages at these nodes.

Summary and Conclusion

●

It is observed that the frequency of the bridge model decreases consistently with increase in the damage introduced in the structure. The decrease in the frequency indicates the damage in the structure but does not indicate the location of the damage induced.

●

The mode shapes showed non-desired flexible joint of the bridge and uneven behavior of the upstream and downstream truss contrary to the expected mode shape of the bridge. The constructed joint of the steel bridges that has been designed for complete rigidity may not exhibit the required behavior thus signifying the importance of generating bridge signature after construction.

●

The mode shape derivatives give better indication of damage location in the structure. Higher Modal curvature difference at nodes indicates the damage members. For lower modes the model curvature difference clearly indicates the damaged points but for higher modes a damaged portion is being indicated. The members damaged in the downstream truss also affects the opposite members of upstream truss.

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●

The modal strain energy change clearly indicates that the defective portion showed erratic modal strain energy values whereas the structural members with good integrity showed uniform distribution of strain energy. The structure with undamaged elements distributes the load efficiently through the development of strain whereas the damaged elements develop higher modal strain energy values.

●

It is emphasized that newly constructed bridge should be accompanied by modal derivatives such as Mode Shape Curvature and Modal Strain Energy Change Ratio so that any defect developed during the construction could be rectified.

References 1.

WAHAB M.M.A. and Roeck G.D., “Damage Detection In Bridges Using Modal Curvatures : Application To A Real Damage Scenario”, Journal of Sound and Vibration, Vol. 226, No. 2, 1999, pp. 217–235.

AHLBORN T.M., SHUCHMAN R., SUTTER L.L., BROOKS C.N., HARRIS D.K.,BURNS J.W., ENDSLEY K.A., EVANS D.C., VAGHEFI K. and OATS R.C., “The State-of-the-Practice of Modern Structural Health Monitoring for Bridges: A Comprehensive Review”, Transport Research Board, Washington D.C., Task 2 Report, June 2010.

REN W.X., ZATAR W. and HARIK I.E., “Ambient vibration-based seismic evaluation of a continuous girder bridge”, Engineering Structures, vol. 26, no. 5, April 2004, pp. 631– 640.

SALAWU O.S. and WILLIAMS C., “Review of full-scale dynamic testing of bridge structures”, Engineering Structures, Vol. 17, No. 2, 1995, pp. 113–121.

PIRNER M. and FRYBA L., “Load tests and modal analysis of bridges”, Engineering Structures, Vol. 23, 2001, pp. 102–109.

WIBERG J., “Bridge Monitoring to Allow for Reliable Dynamic FE Modelling A Case Study of the New Årsta Railway Bridge”, Royal Institute of Technology (KTH), University of Stockholm, Sweden, KTH/BKN/B-81-SE, March, 2006.

ADAMS R. D. and COPPENDALE J., “Measurement of the Elastic Moduli of Volume 46 │ Number 3 │ September 2016

Structural Adhesives by a Resonant Bar Technique”, Journal of Mechanical Engineering Science, Vol. 18, 1976, pp. 149–158. 8.

GANDOMI A.H., SAHAB M.G., RAHAEI A. and GORJI M.S., “Development in Mode Shape-Based Structural Fault Identification Technique”, World Applied Sciences Journal, Vol. 5, No. 1, 2008, pp. 29–38.

DUTTA A., “Damage detection in bridges using accurate modal parameters”, Finite Elements in Analysis and Design, vol. 40, No. 3, January 2004, pp. 287–304.

.10. DAWARI V.B. and VESMAWALA G.R., “Structural Damage Identification Using Modal Curvature Differences”, IOSR Journal of Mechanical and Civil Engineering, pp. 33–38. 11. VESTRONI F. and CAPECCHI D., “Damage detection in beam structures based on frequency measurements”, Journal of Engineering Mechanics, Vol. 126, No. 7, 2000, pp. 761–768. 12. ALAMPALLI B.S.(MEMBER, ASCE), FU G. (MEMBER, ASCE) and DILLON E. W., “Signal Versus Noise In Damage Detection By Experimental Modal Analysis”, Journal of Structural Engineering, Vol. 58, February 1997, pp. 237–245. 13. SHI Z.Y., LAW S.S. and ZHANG L.M., “Damage localization by directly using incomplete mode shapes”, Journal of Engineering Mechanics, June 2000, pp. 656–660. 14. YUEN K.V., “Updating large models for mechanical systems using incomplete modal measurement”, Mechanical Systems and Signal Processing, August 2011, pp. 1–12. 15. WANG S.Q. and LI H.J., “Assessment of structural damage using natural frequency changes”, Acta Mechanica Sinica, Vol. 28, No. 1, January 2012, pp. 118–127. 16. DOUGLAS B. M. and REID W. H., “Dynamic tests and system identification of bridges”, Journal of the Structural Division, Vol. 108, No. 10, 1982, pp. 2295–2312. 17. SHI Z. Y., LAW S. S. and ZHANG L. M., “Structural Damage Localization from Modal

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Strain Energy Change”, Journal of Sound and Vibration, Vol. 218, No. 5, 1998, pp. 825–844. 18. PRADEEPK. R., RAO B. N., SRINIVASAN S. M. and BALASUBRAMANIAM K., “Modal strain energy change ratio for damage identification in honeycomb sandwich structures”, Canadian Journal of Basic and Applied Sciences, Vol. 02, No. 01, 2014, pp. 10–24. 19. NDAMBI J.M., VANTOMME J. and HARRI K., “Damage assessment in reinforced concrete beams using eigenfrequencies and mode shape derivatives”, Engineering Structures, Vol. 24, 2002, pp. 501–515. 20. SAP2000 “Structural analysis program”, Computers and Structures Inc., Berkeley, CA., 2006. 21. HSIEH K., HALLING M. and BARR P., “Overview of vibrational structural health monitoring with representative case studies”, Journal of Bridge Engineering, Vol. 11, No. 6, 2006, pp. 707–715. 22. PICOZZI M., MILKEREIT C., ZULFIKAR C., FLEMING K., DITOMMASO R., ERDIK M., ZSCHAU J., FISCHER J., ŞAFAK E., ÖZEL O. and APAYDIN N., “Wireless technologies for the monitoring of strategic civil infrastructures: An ambient vibration test on the Fatih Sultan Mehmet Suspension Bridge in Istanbul, Turkey”, Bulletin of Earthquake Engineering, Vol. 8, No. 3, 2010, pp. 671–691. 23. CHOUBEY A., SEHGAL D. K. and TANDON N., “Finite element analysis of vessels to study changes in natural frequencies due to cracks”, International Journal of Pressure Vessels and Piping, Vol. 83, No. 3, 2006, pp. 181–187. 24. ROY K. and RAY-CHAUDHURI S., “Fundamental mode shape and its derivatives in structural damage localization”, Journal of Sound and Vibration, Vol. 332, No. 21, 2013, pp. 5584–5593. 25. PANDEY A. K., BISWAS M. and SAMMAN M. M., “Damage detection from changes in curvature mode shapes”, Journal of Sound and Vibration, Vol. 145, No. 2, 1991, pp. 321–332.

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Indian National Group of the IABSE Office Bearers and Managing Committee - 2016

Chairman 1.

Shri D.O. Tawade, Chairman, ING-IABSE & Member (Technical), National Highways Authority of India

Vice-Chairmen 2.

Shri B.N. Singh, Additional Director General, Ministry of Road Transport and Highways

Shri Alok Bhowmick, Managing Director, B&S Engineering Consultants Pvt. Ltd.,

Shri A.K.S. Chauhan, C.O.O., GR Infraprojects Ltd.,

Honorary Treasurer 5.

The Director General (Road Development) & Special Secretary to the Government of India, Ministry of Road Transport and Highways

Honorary Members 6.

Shri Ninan Koshi, Honorary Member, IABSE & Former Director General (Road Development) & Addl. Secretary Prof. S.S. Chakraborty, Honorary Member & Past Vice-President, IABSE

Persons represented ING on the Executive Committee and Technical Committee of the IABSE 8.

Dr. Harshavardhan Subbarao, Vice President & Member, Technical Committee of IABSE & Chairman and Managing Director, Construma Consultancy Pvt. Ltd., Past Member of the Executive Committee and Technical Committee of IABSE

Prof. S.S. Chakraborty, Honorary Member & Past Vice-President, IABSE

10. Dr. B.C. Roy, Past Vice President & Member, Technical Committee, IABSE

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Honorary Secretary 11. Shri R.K. Pandey, Member (Projects), National Highways Authority of India

Members of the Executive Committee 12. Shri A.D. Narain, Former Director General (Road Development) & Addl. Secretary 13. Shri A.K. Banerjee, Former Member (Technical), NHAI 14. Shri A.V. Sinha, Former Director General (Road Development) & Special Secretary 15. Shri G. Sharan, Former Director General (Road Development) & Special Secretary 16. Shri R.P. Indoria, Former Director General (Road Development) & Special Secretary 17. Dr. Lakshmi Parameswaran, Chief Scientist, Bridges & Structures Div., CSIR-Central Road Research Institute 18. Shri Ashwinikumar B. Thakur, Group Engineer, Atkins India 19. Shri Sarvagya Kumar Srivastava, Engineer-inChief (Projects), Govt of Delhi 20. Dr. Mahesh Kumar, Engineer Member, Delhi Development Authority

Secretariat 21. Shri R.K. Pandey, Member (Projects), National Highways Authority of India 22. Shri Ashish Asati, General Manager, National Highways Authority of India 23. Shri K.B. Sharma, Under Secretary, Indian National Group of the IABSE

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MEMBERS OF THE MANAGING COMMITTEE – 2016 Rule-9 (a): A representative of the Union Ministry of Road Transport and Highways 1.

Shri D.O. Tawade, Chief Engineer (CoordinatorII), Ministry of Road Transport & Highways

Rule-9 (b): A representative each of the Union Ministries/Central Government Departments making annual contribution towards the funds of the Indian National Group of IABSE as determined by the Executive Committee from time to time

19. Govt of Madhya Pradesh – nomination awaited 20. Dr. D.T. Thube, Chief Engineer, Govt of Maharashtra 21. Shri O. Nabakishore Singh, Additional Chief Secretary (Works), Govt of Manipur 22. Govt of Meghalaya – nomination awaited 23. Shri Lalmuankima Henry, Chief Engineer (Buildings), Govt of Mizoram 24. Govt of Nagaland – nomination awaited

CPWD - nomination awaited

25. Govt of Orissa – nomination awaited

NHAI - nomination awaited

26. Govt of Punjab – nomination awaited

Ministry of Railways - nomination awaited

27. Govt of Sikkim – nomination awaited

Rule-9 (c): A representative each of the State Public Works Departments/Union Territories making annual contribution towards the funds of the Indian National Group of IABSE as determined by the Executive Committee from time to time 5.

Govt of Andhra Pradesh – nomination awaited

Govt of Arunachal Pradesh – nomination awaited

Shri Ajoy Chandra Bordoloi, Commissioner & Special Secretary to the Govt of Assam

Govt of Bihar – nomination awaited

Govt of Chattisgarh – nomination awaited

10. Shri Sarvagya Kumar Srivastava, Engineer-inChief (Projects), Govt of Delhi 11. Govt of Goa – nomination awaited 12. Govt of Gujarat – nomination awaited 13. Govt of Haryana – nomination awaited 14. Govt of Himachal Pradesh – nomination awaited 15. Govt of Jammu & Kashmir – nomination awaited 16. Govt of Jharkhand – nomination awaited

28. Govt of Tamil Nadu – nomination awaited 29. Govt of Tripura – nomination awaited 30. Govt of Uttar Pradesh – nomination awaited 31. Govt of Uttarakhand – nomination awaited 32. Govt of West Bengal – nomination awaited 33. Shri Mukesh Anand, Chief Engineer, Union Territory Chandigarh Rule-9 (d): A representative each of the Collective Members making annual contribution towards the funds of the Indian National Group of IABSE as determined by the Executive Committee from time to time 34. Major V.C. Verma, Director (Mktg), Oriental Structural Engineers Pvt. Rule-9 (e): Ten representatives of Individual and Collective Members 35. Shri G. Sharan, Former DG (RD) & Special Secretary 36. Shri A.K. Banerjee, Former Member (Technical), NHAI 37. Shri AV Sinha, Former DG (RD) & Special Secretary

17. Govt of Karnataka – nomination awaited

38. Shri R.P. Indoria, Former DG (RD) & Special Secretary

18. Shri K.P. Prabhakaran, Chief Engineer, Govt of Kerala

39. Shri. Atul D. Bhobe, Managing Director, S.N. Bhobe & Associates Pvt. Ltd.

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40. Shri N.K. Sinha, Former DG (RD) & Special Secretary

53. Shri Bageshwar Prasad, CEO (Delhi Region), Construma Consultancy Pvt. Ltd.

41. Dr. Lakshmi Parameswaran, Chief Scientist, Bridges & Structures Div, CSIR-Central Road Research Institute

54. Shri Alok Bhowmick, Managing Director, B&S Engineering Consultants Pvt. Ltd.

42. Shri Rakesh Kapoor, General Manager, Holtech Consulting Pvt. Ltd.,

Rule-9 (i): Honorary Treasurer of the Indian National Group of IABSE

43. Shri Inderjit Ghai, CEO, Consulting Engineers Associates

55. The Director General (Road Development) & Special Secretary to the Govt of India

44. Shri Ashwinikumar B. Thakur, Group Engineer, Atkins India Rule-9 (f): Four representatives of Bridge and Structural Engineering Firms

Rule-9 (j): Past-Chairman of the Society, for a period of three years, after they vacate their Chairmanship

45. Shri M.V. Jatkar, Executive Director (Technical), Gammon India Ltd.

Rule-9 (k): Secretary of the Indian National Group of IABSE.

46. The Managing Director, UP State Bridge Corporation Ltd.

56. Shri R.K. Pandey

47. Shri T. Srinivasan, Vice President & Head Ports, Tunnels & Special Bridges, Larsen & Toubro Ltd.

Rule-9 (l): Persons who have been awarded Honorary Membership of the Parent Body

48. Vacant Rule-9 (g): Two representatives of the Engineering Colleges / Technical Institutes / Universities / Research Institutes 49. Prof. A.K. Goel, Director, Indian Railways, Pune 50. Shri V.L. Patankar, Director, Indian Academy of Highway Engineers Rule-9 (h): Four representatives Engineering Firms

Consulting

51. Shri A.D. Narain, President, ICT Pvt. Ltd. 52. Shri Krishna Sandepudi, Vice President, Aarvee Associates Architects Engineers & Consultants Pvt. Ltd.

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57. Shri Ninan Koshi 58. Prof. S.S. Chakraborty Rule-9 (m): Persons represented ING on the Executive Committee and Technical Committee of the IABSE 59. Dr. Harshavardhan Subbarao Rule-9 (n): Past Members of the Executive Committee and Technical Committee of the IABSE 60. Prof. S.S. Chakraborty 61. Dr. B.C. Roy

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With Best Compliments From :

Reliance Infrastructure Ltd Mumbai

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