Environmental & Engineering Geoscience by Association of Environmental & Engineering Geologists (AEG)

Environmental & Engineering Geoscience FEBRUARY 2017

VOLUME XXIII, NUMBER 1

THE JOINT PUBLICATION OF THE ASSOCIATION OF ENVIRONMENTAL AND ENGINEERING GEOLOGISTS AND THE GEOLOGICAL SOCIETY OF AMERICA SERVING PROFESSIONALS IN ENGINEERING GEOLOGY, ENVIRONMENTAL GEOLOGY, AND HYDROGEOLOGY

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EDITORIAL BOARD ROBERT H. SYDNOR JEROME V. DEGRAFF USDA Forest Service Consulant THOMAS J. BURBEY CHESTER F. WATTS (SKIP) Virginia Polytechnic Institute Radford University SYED E. HASAN University of Missouri, Kansas City ASSOCIATE EDITORS PAUL M. SANTI JOHN W. BELL Nevada Bureau of Mines and Colorado School of Mines Geology ROBERT L. SCHUSTER U.S. Geological Survey RICHARD E. JACKSON Geofirma Engineering, Ltd. ROY J. SHLEMON R. J. Shlemon JEFFREY R. KEATON AMEC Americas & Associates, Inc. PAUL G. MARINOS GREG M. STOCK National Technical University National Park Service of Athens, Greece RESAT ULUSAY Hacettepe University, Turkey JUNE E. MIRECKI U.S. Army Corps of CHESTER F. “SKIP” WATTS Radford University Engineers TERRY R. WEST PETER PEHME Waterloo Geophysics, Inc Purdue University NICHOLAS PINTER Southern Illinois University SUBMISSION OF MANUSCRIPTS Environmental & Engineering Geoscience (E&EG), is a quarterly journal devoted to the publication of original papers that are of potential interest to hydrogeologists, environmental and engineering geologists, and geological engineers working in site selection, feasibility studies, investigations, design or construction of civil engineering projects or in waste management, groundwater, and related environmental fields. All papers are peer reviewed. The editors invite contributions concerning all aspects of environmental and engineering geology and related disciplines. Recent abstracts can be viewed under “Archive” at the web site, “http://eeg.geoscienceworld.org”. Articles that report on research, case histories and new methods, and book reviews are welcome. Discussion papers, which are critiques of printed articles and are technical in nature, may be published with replies from the original author(s). Discussion papers and replies should be concise. To submit a manuscript go to http://eeg.allentrack.net. If you have not used the system before, follow the link at the bottom of the page that says New users should register for an account. Choose your own login and password. Further instructions will be available upon logging into the system. Please carefully read the “Instructions for Authors”. Authors do not pay any charge for color figures that are essential to the manuscript. Manuscripts of fewer than 10 pages may be published as Technical Notes. For further information, you may contact Dr. Abdul Shakoor at the editorial office.

THIS PUBLICATION IS PRINTED ON ACID-FREE PAPER EDITORS ABDUL SHAKOOR Department of Geology Kent State University Kent, OH 44242 330-672-2968 ashakoor@kent.edu

BRIAN G. KATZ Florida Department of Environmental Protection 2600 Blair Stone Rd. Tallahassee, FL 32399 850-245-8233 eegeditorbkatz@gmail.com

Cover photo View of a cut slope instability affecting a hydroelectric power plant site, Turkey.. See article on page 23. Photo by E. Tuncay & R. Ulusay.

Environmental & Engineering Geoscience Volume 23, Number 1, February 2017 Table of Contents

Utilization of Polymers to Improve Soft Clayey Soils Using the Deep Mixing Method Seracettin Arasan, Majid Bagherinia, R. Kagan Akbulut, and Ahmet Sahin Zaimoglu

Three-Dimensional Stability Assessment of a Complex Landslide in the RjecË&#x2021;ina Vally, Croatia Chunxiang Wang

Assessment of the Mechanism of a Slope Failure in a Hydroelectric Power Plant Site and Considerations on Some Remedial Measures ErguÂ¨n Tuncay and Resat Ulusay

Assessing the Geological Sources of Manganese in the Roanoke River Watershed, Virginia Zachary A. Kiracofe, William S. Henika, and Madeline E. Schreiber

Utilization of Polymers to Improve Soft Clayey Soils Using the Deep Mixing Method SERACETTIN ARASAN Civil Engineering Department, Engineering Faculty, Ataturk University, Erzurum, Turkey, 25240, arasan@atauni.edu.tr

MAJID BAGHERINIA Civil Engineering Department, Engineering Faculty, Ataturk University, Erzurum, Turkey, 25240, altinkaradagli84@gmail.com

R. KAGAN AKBULUT1 Technical Vocational School of Higher Education, Ataturk University, Erzurum, Turkey, 25240, rkakbulut@atauni.edu.tr

AHMET SAHIN ZAIMOGLU Civil Engineering Department, Engineering Faculty, Ataturk University, Erzurum, Turkey, 25240, zaimoglu@atauni.edu.tr

Key Terms: Deep Mixing, Polyester, Guar Gum, Unconfined Compressive Strength, Freeze-Thaw ABSTRACT Deep mixing is an improvement method performed in the creation of mixed columns, which includes the in situ mixing of soil and lime or Portland cement using special equipment (using rigs with counter-rotating augers). In this study, series of unconfined compression tests were carried out on soft clayey soils using deep mixing with polymers. In the experiments, two polymers (i.e., polyester and guar gum) and lime were used as binder materials at different ratios. Samples cured for 14, 28, and 150 days were exposed to five and 10 cycles of freezethaw, and samples that were not exposed to freeze-thaw cycles were tested in order to investigate the freeze-thaw effect. The unconfined compressive strength increased continuously with the increase of polyester and curing period, while the changes in unconfined compressive strength with increase of freeze-thaw cycles were insignificant. The overall evaluation of the results has revealed that polyester and guar gum showed potential as candidates for deep mixing applications in soft clayey soils. INTRODUCTION The deep mixing method is mostly used in slope stability, embankment supports, hydraulic cut-off walls, 1

Corresponding author email: rkakbulut@atauni.edu.tr.

excavation support walls, liquefaction mitigation, environmental remediation, in situ reinforcement, and large-volume ground treatment applications (Bruce et al., 1998; Bruce and Bruce, 2003; and Terashi and Kitazume, 2009, 2011). The deep mixing method involves in situ mixing of soil and binder materials with special equipment, frequently by using rigs with counter-rotating augers (Taki and Yang, 1991). The method was first performed in the United States in 1954. The Intrusion Prepakt Co. (United States) developed the earlier deep mixing application (the mixed-in-place technique), which saw only sporadic use in the United States in 1954 (Bruce et al., 1999). Japan and Scandinavia have carried out large-scale researches and applications regarding deep mixing since 1967. Specifically, soft clays have been improved with lime columns in Sweden. Similarly, the Japan Ministry of Transport studied stabilization of soft sea soils with grained lime in 1967. Over the past few years, the deep mixing method has often been used to solve infrastructure and seismic problems and to enhance environmental remediation in the United States (Bruce et al., 1998; Bruce and Bruce, 2003). Lime and cement have been used as the primary binder materials in deep mixing applications since the 1970s, and they are now currently used extensively worldwide. Commonly, two different techniques are utilized: column installation and slurry pressure injection. These two techniques decrease soil moisture content, thereby reducing shrinkage and swelling while enhancing strength and compaction properties (Glendinning and Rogers, 1996; Threadgold, 1996; Bruce

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et al., 1998; Rogers et al., 2000; Ahnberg et al., 2003; Bruce and Bruce, 2003; and Arasan and Nasirpur, 2015). The binder type and its mixing ratio are two important factors influencing deep mixing performance. Numerous experimental studies dealing with the effects of binder types and ratios on deep mixing method are available in the literature. Some of these studies have focused on the soil/binder (cement and/or lime) ratios and the water/cement-lime (w/c) ratio (Okumura and Terashi, 1975; Terashi and Tanaka, 1981; Ahnberg, 1996; Porbaha et al., 1998, 2000; Bahner and Naguib, 2000; Jacobson, 2002; Lorenzo and Bergado, 2004, 2006; Rutherford, 2004; Pathivada, 2005; Lewsley, 2008; Shrestha, 2008; Şengör, 2011; Miura et al., 2002; Horpibulsuk et al., 2005; Maher et al., 2007; Tang et al., 2011; and Dias et al., 2012), while others have studied curing periods (Hartlen and Holm, 1995; Andromalos and Bahner, 2004). Most researchers have pointed out that the unconfined compressive strength (UCS) values increased when binder ratio and curing time were increased, and the w/c ratio was decreased. The UCS values obtained by these researchers were between 0.1 and 8.0 MPa with 28 days of curing time. Matsuo et al. (1996) reported that UCS values between 1.4 and 7.4 MPa were obtained when marine clays were mixed with 13–32 percent cement (cement to dry soil). Similarly, Bergado and Lorenzo (2005) reported that UCS values of 0.3–1.0 MPa were obtained when Bangkok clays were mixed with 10–15 percent cement (cement to dry soil). Taki (2002) also reported that UCS values between 1.7 and 4.9 MPa were obtained when marine clays were mixed with 200–300 kg/m3 cement (cement per unit volume of wet soil). Filz et al. (2005) and Liu et al. (2008) mentioned similar results from their studies. There are a limited number of studies in the literature on the utilization of materials other than cement or lime in the improvement of soils using the deep mixing method (Ahnberg and Holm, 1996; Ahnberg, 2006; and Ajorloo, 2010.). Silica fume, fly ash, and slag are generally used with or without cement in deep mixing applications. However, one group of researchers focused on the stabilization of soils using polymers (Ahmed, 1995; Bishop et al., 1998; Al-Khanbashi and Abdalla, 2006; Gallagher et al., 2007; Gupta et al., 2009; Naeini and Ghorbanalizadeh, 2010; and Cabalar and Canakci, 2011). Polymers are widely used in industry, but they are rarely used in soil stabilization applications. In this respect, some commercially available and non-traditional additives, such as emulsions, acids, lignin derivatives, enzymes, tree resins, and silicates, can be used as soil stabilization binders. These binders may exist in liquid or solid form and may be applicable for most soils (Newman and Tingle, 2004).

Table 1. Index properties of clay (Bagherinia, 2013). CL Clay Clay content, ,0.002 mm (%) Finer content, ,0.075 mm (%) Specific gravity, GS Liquid limit, wL (%) Plastic limit, wP (%) Plasticity index, IP (%) Optimum moisture content*, OMC (%) Maximum dry unit weight, γdmax (kN/m3) Hydraulic conductivity**, k (cm/s)

10 80 2.77 40 23 17 15 18.3 6.974 6 10−7

*From standard Proctor test. **From standard Proctor optimum moisture content.

A polyester is defined as a condensation or stepgrowth polymer containing in-chain ester units, because it is the long chains that give polymers their unique properties (McIntyre, 2004). With an 18 percent market share of all plastic materials produced, polyesters are widely used as thermoset plastic, fiber, and fabric for many industries, such as auto and marine body parts, building panels, clothing, carpet, furniture, etc. Significant research efforts have therefore been performed in polymer science (McIntyre, 2004; Scheirs and Long, 2005). However, very limited information exists on polymers in concrete technology (Ates, 2008; Lim et al., 2009; Jamshidi et al., 2013; and Martínez-Barrera et al., 2013) and soil stabilization (Akbulut et al., 2013; Bagherinia, 2013). Biopolymers are polymers that are produced by living organisms. In other words, they can be described as polymeric biomolecules. Commercially available biopolymers include gum arabic, guar gum, and locust bean gum from botanic sources; starches from corn or tapioca; xanthan gum from bacteria; and gelatin derived from animal skin or bones (Van de Velde and Kiekens, 2002; Chang and Cho, 2012). Biopolymers are used for soil improvement in the fields of geotechnical engineering and geo-environmental engineering (Cabalar and Canakci, 2011; Chang and Cho, 2012; Bagherinia, 2013; Khatami and O’Kelly, 2013; and Arasan and Nasirpur, 2015), reducing permeability of soils (Bouazza et al., 2009; Wiszniewski et al., 2013), and as soil drilling mud and temporary excavation supports (Mitchell and Santamarina, 2005). Similar to this trend, theoretical and experimental verifications of the interactions between various types of biopolymers and soil media are necessary in the geotechnical and geoenvironmental fields (Chang and Cho, 2012). Moreover, it can be said that biopolymers could reduce the degree of saturation and contaminant leaching (He et al., 2013; Arasan and Nasirpur, 2015). Soft clayey soils have great potential for stability problems, such as landslides, bearing capacity failure,

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Polymers to Improve Soft Clay

Figure 1. XRD analysis result for the soft clay.

and collapse. For that reason, studies on the stabilization of clay soils with polymers using the deep mixing method became more important for geotechnical knowledge and practice. The current study was performed to investigate the effects of polyester and guar gum on unconfined compressive strength and freeze-thaw properties of soft clayey soil, i.e., low-plasticity clay (CL) under three different curing periods (14, 28, and 150 days) and three different freeze-thaw cycles (0, 5, and 10 cycles).

to pass through a 1 mm sieve for easy processing and uniform water content. Then, the clay was mixed with the required water content. The properties of clay are shown in Table 1. Water content of 40 percent (mass) was chosen for experiments in this study on the basis of the liquid limit value of clay. According to the X-ray diffraction results for soft clay, the material sampled is composed of clay (56 percent), quartz (28 percent), cristobalite (11 percent), and tridymite (5 percent). The clay particles are also composed of only kaolinite minerals (100 percent), as shown in Figure 1.

MATERIALS AND METHODS Soft Clayey Soil

Polymers

The CL clay was brought from the field (Çankırı, Turkey) and then dried. The dry clay was powdered

The experiments were carried out with polyester used as the polymer, guar gum as the biopolymer,

Table 2. Some properties of unsaturated polyester and chemicals. Properties

Unsaturated Polyester

Accelerator

Catalyst

Name Formula Color Solid ratio Specific gravity

Polyethylene terephthalate (C10H8O4)n Colorless 65% 1.13

Cobalt(II) naphthenate CoC22H14O4 Bluish-red 6% (Co+2) 0.96

2-hydroperoxy-2-([2-hydroperoxybutan-2-yl]peroxy)butane C8H18O6 Water white — 1.17

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Arasan, Bagherinia, Akbulut, and Zaimoglu Table 3. Some properties of guar gum. Properties

Table 5. Preliminary test results of additives (Bagherinia, 2013).

No.

Percentages of Polyester (%)

Percentages of Guar Gum + Lime (%)

UCS for 14 Days (kPa)

UCS for 28 Days (kPa)

0.02 + 3 0.25 + 3 0 0.02 0.05 0.15 0.25 0.5 1 0.02 + 3 0.25 + 3 0 0.02 0.05 0.15 0.25 0.5 1 0.02 + 3 0.25 + 3 0 0.02 0.05 0.15 0.25 0.5 1 0.02 + 3 0.25 + 3 0 0.02 0.05 0.15 0.25 0.5 1

425.4 409.4 470 436 550 750 680 684 700 540.8 520 682 1,350 1,200 1,400 1,430 1,400 1,410 723.4 710.5 855 1,500 1,450 1,400 1,587 1,600 1,610 7,322.5 6,973 1,208 1,678 1,686 1,600 2,188 2,150 2,190

418 413 — — — — — — — 544 568 — — — — — — — 745 735 — — — — — — — 8,033.5 7,207 — — — — — — —

Guar Gum

Name Solubility Physical state pH Bulk density

E412 Insoluble in ethanol Free flowing white powder 5.0–6.5 600 g/L

and lime. An unsaturated polyester (casting type), an accelerator (cobalt naphthenate), and a catalyst (methyl ethyl ketone peroxide [MEKP]) produced by Dewilux Company (Turkey) were used to prepare the polyester. Some physical and chemical properties of those chemicals, obtained from manufacturers, are given in Table 2. Some previous studies regarding the cobalt naphthenate and MEKP suggested that they should be used at the ratio of 2 percent and 0.4 percent (by the weight of polyester), respectively (Akbulut et al., 2013; Bagherinia, 2013). Physical and chemical properties of guar gum provided by the manufacturer (A&D Chemicals Industry Company, Turkey) are given in Table 3. Specimen Preparation and Testing In order to determine which percentage of both accelerator and hardener can be used, a preliminary study was performed by Bagherinia (2013). Table 4 summarizes the preliminary study results. The polyester was prepared by adding 2 percent cobalt naphthenate as the accelerator and 0.4 percent MEKP as the Table 4. Preliminary test results of accelerator and hardener percentages (Bagherinia, 2013). Specimen No.

Accelerator (%)

Accelerator/ Hardener

UCS (kPa)

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18

2 2 2 5 5 5 10 10 10 20 20 20 30 30 30 40 40 40

2 4 5 2 4 5 2 4 5 2 4 5 2 4 5 2 4 5

44,300 82,940 117,000 32,420 58,500 90,000 18,750 35,220 47,000 6,755 10,750 13,500 4,700 5,038 5,335 Boiling occurred Boiling occurred Boiling occurred

catalyst to the unsaturated polyester in the study. First, the unsaturated polyester and accelerator were mixed for 1 minute by hand, and then the catalyst was added and mixed for 1 minute by hand. After producing the polyester, series of preliminary tests were also conducted to determine the percentage of polyester and other additive types and percentages (guar gum and lime) to be used in this study (Bagherinia, 2013). The preliminary test results are shown in Table 5. These tests determined the poly‐ ester percentages to be 5 percent, 10 percent, and 20 percent (mass). On the other hand, 0.02 percent, and 0.25 percent of guar gum and 3 percent (mass) of lime were chosen for the experiments. The percent‐ ages of additives were determined by the total weight of soft clay at 40 percent water content (clay and

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Polymers to Improve Soft Clay Table 6. Percentages of used additives. Materials (%) Specimen Name

Clay at 40% Water Content

Polyester

Guar Gum

Lime

Total

S1 S2 S3 S4 S5 S6 S7 S8 S9

100 96.98 96.75 91.98 91.75 86.98 86.75 76.98 76.75

— — — 5 5 10 10 20 20

— 0.02 0.25 0.02 0.25 0.02 0.25 0.02 0.25

— 3 3 3 3 3 3 3 3

100 100 100 100 100 100 100 100 100

water). The quantities and percentages of soil sample additives are summarized in Table 6. Subsequently, the prepared polyester, guar gum, and lime were added to the clay (40 percent water content CL clay) and mixed for 5 minutes using a 150 rpm mechanical mixer. The test samples were prepared in accordance with the procedures described in JGS 0821 (2000) and Euro Soil Stab (2001). The prepared mixtures were then placed into metal cylinder molds measuring 38 mm in diameter and 76 mm in height in three layers. The inner surfaces of the molds were lubricated to make extrusion of the samples easier (Figure 2a). The molds were vibrated slightly to remove air bubbles. The prepared samples were removed from the molds after 7 days (Figure 2b) and cured in a moist room where the temperature was kept at 20 ¡ 3 uC and humidity was at 70 percent relative humidity (Figure 2c). To investigate the effects of curing periods on strength, 14 days, 28 days, and 150 days were determined to be the curing periods. At the end of each curing period, the samples were subjected to the UCS test. UCS tests were performed according to ASTM D 2166 (2013). The testing load was applied at a rate of 0.8 mm/min. Each sample was prepared at least in triplicate, and all tests were done in triplicate as well. The freeze-thaw tests were performed using a programmable cabinet. It should be mentioned that all cured samples were subjected to the freeze-thaw test. The samples were placed in to the cabinet and conditioned at −20 uC for 6 hours; then they were thawed at +25 uC for 6 hours (Zaimoglu, 2010). This process was designated as “one cycle.” All samples were subjected to five and 10 freeze-thaw cycles.

RESULTS AND DISCUSSION In the following, the effects of polyester ratio, curing periods, and freeze-thaw cycles on the UCS of the

Figure 2. (a) Samples in metal cylinder molds. (b) Two samples removed from molds after 7 days. (c) Curing of samples in the moisture room.

samples are presented. Additionally, the findings from the experimental tests are compared with those from other studies in the literature and are discussed. Effects of Polymer Ratio and Curing Periods The relationship between polyester ratio and UCS of the 14 day cured samples is given in Figure 3. The UCS values of all reinforced samples increase with increasing polyester ratio (Figure 3). As is seen in Figure 3, the increment of UCS values for polyester ratios between 5 percent and 10 percent is generally lower than the values for ratios between 10 percent and 20 percent. The sample with 0.02 percent guar gum gave maximum UCS values during the curing period of 14 days. The UCS values of this sample

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Arasan, Bagherinia, Akbulut, and Zaimoglu

Figure 3. UCS versus polyester ratio for the samples that were cured for 14 days.

at 0 percent, 5 percent, 10 percent, and 20 percent polyester ratios were obtained as 0.43, 0.54, 0.72, and 7.3 MPa, respectively. On the other hand, the UCS values of the 0.25 percent guar gum sample at 0 percent, 5 percent, 10 percent, and 20 percent polyester ratios were obtained as 0.41, 0.52, 0.71,

and 7.0 MPa, respectively (Figure 3). It should be also noted that the lowest UCS value obtained in this study is higher than that of the lower UCS limit of fine-grained soil (0.2 MPa) for the deep mixing method reported by Bruce and Bruce (2003) and Bruce et al. (1998).

Figure 4. 3-D column illustration of UCS/curing period/polyester ratio of samples with respect to the guar gum (GG) ratio.

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Figure 5. SEM micrographs of samples cured for 150 days: (a) raw clay, (b) 0.25 percent guar gum with 20 percent polyester, sample S9, (c) 0.02 percent guar gum with 20 percent polyester, sample S8, and (d) pure polyester.

On the other hand, Figure 3 shows that the UCS values of samples without adding polyester (only guar gum and lime added to clayey soils samples S2 and S3) are higher than 0.2 MPa, the lower UCS limit of fine-grained soil. It is also mentioned that the guar gum reduces the moisture degree of the samples as observed by Gupta et al. (2009), Bagherinia (2013), and Arasan and Nasirpur (2015). Due to the decreasing moisture degree and the chemical reactions occurring between guar gum, lime, and clayey soil, the samples became strengthened structures. On the basis of these results, it could therefore be concluded that

polyester and guar gum may be used for deep mixing applications in clayey soils. The column graphs of Figure 4 are drawn in three dimensions to display the combined effect of polyester ratio, guar gum ratio, and curing periods on UCS. It can clearly be seen in Figure 4 that UCS values increase drastically after a 20 percent polyester ratio is reached for each curing period. Nevertheless, the UCS values of all samples are higher than 0.2 MPa (the lower UCS limit of fine-grained soil). However, the increase during the curing period has no important effect on the UCS values of 0 percent, 5 percent, and

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Figure 6. SEM micrographs at different magniﬁcations of samples cured for 150 days: (a1–a2) raw clay sample, and (b1–b2) 0.25 percent guar gum with 20 percent polyester, sample S9.

10 percent polyester samples. The highest UCS value of this study was obtained from sample S9 with 0.25 percent guar gum and 20 percent polyester ratio and a curing period of 150 days. The UCS values of this sample for the curing periods of 14 days, 28 days, and 150 days were 7.0 MPa, 7.2 MPa, and 13.4 MPa, respectively (Figure 4). However, sample S8 (0.02 percent guar gum with 20 percent polyester) gave maximum UCS values for the curing periods of 14 and 28 days. The UCS values of sample S8 for the curing periods of 14 days, 28 days, and 150 days were recorded as 7.3 MPa, 8.0 MPa, and 12 MPa,

respectively. This different behavior suggests that the guar gum ratio is more effective on the UCS values after 28 days of curing time. On the other hand, the UCS values of the 20 percent polyester samples gave higher values than 7 MPa. This positive behavior indicates that the samples become a polyester-clay composite material when 20 percent polyester is added to the clayey soil. In other words, polyester dominates the behavior of the samples mixed with 20 percent polyester. This striking increase in UCS of the 20 percent polyester samples is attributed to the adhesion

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Polymers to Improve Soft Clay

Figure 7. Effects of freeze-thaw cycles on cured samples.

properties of polyester. Similarly, other researchers have found that the polymers used in their studies increased the UCS (Al-Khanbashi and Abdalla, 2006; Ates, 2013; and Arasan and Nasirpur, 2015). The reason for this strength increment was considered to be that clayey soil formed a new solid structure with polyester. Figures 5 and 6 show scanning electron micrographs of cured specimens supporting this conclusion. It is clearly seen in Figures 5 and 6 that the polyester covers the clay particles and causes them to adhere to each other. Similar findings were reported by researchers who conducted studies on both polymer (Al-Khanbashi and Abdalla, 2006; Ates, 2013; and Arasan and Nasirpur, 2015) and polyester (Akbulut et al., 2013). Additionally, Al-Khanbashi and Abdalla (2006) indicated that polymers alter the properties of soil by structural changes. Effect of Freeze-Thaw An experimental study was conducted on samples that were cured for 14, 28, and 150 days to determine the freeze-thaw effect on samples. The samples were exposed to five and 10 freeze-thaw cycles. The UCS values of samples subjected and not subjected (zero cycle) to freeze-thaw cycles are given for 14, 28, and 150 days of curing time in Figure 7. It was discovered (Figure 6) that the freeze-thaw cycles did not significantly affect the UCS values of the samples. The reason for that is considered to be the durability of polyesters to extraordinary environmental impacts. Similarly, Welling (2012) reported that the durability of stabilized soils increased with an increase in the polymer content. Polyester can be used as a binder material to improve the mechanical behavior of clayey soils that are exposed to freeze-thaw cycles.

CONCLUSIONS In this study, UCS tests were performed on samples to investigate the effect on properties of clays stabilized with polyester and guar gum. Based on the test results and the discussion presented in this study, the following conclusions were made: . The UCS values of all samples increased with the increase in the polyester ratio. . The maximum UCS values of the samples that were cured for 14 days were obtained from the sample with 0.02 percent guar gum. The UCS values of this sample for 5 percent, 10 percent, and 20 percent polyester ratios were determined to be 0.54, 0.72, and 7.3 MPa, respectively. . The highest UCS value of this study was obtained from the sample with 0.25 percent guar gum and 20 percent polyester ratio, and a 150 day curing period. The UCS values of this sample for the curing periods of 14 days, 28 days, and 150 days were obtained as 7.0 MPa, 7.2 MPa, and 13.4 MPa, respectively. . It was determined that UCS values of samples with no added polyester (only guar gum and lime added to clayey soils) are higher than 0.2 MPa, which is the lower UCS limit of fine-grained soil. It is also mentioned that guar gum reduced the moisture degree of samples and that the samples then became strengthened structures. . It was observed that the freeze-thaw cycles did not have an effect on the UCS values of stabilized samples. It could be said that soil stabilizations using polyester can also be used as an alternative to cement and lime in cold climate environments and in areas where freeze-thaw is a factor.

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On the basis of this study, it could therefore be concluded that polyester and guar gum may be used for deep mixing applications in clayey soils. It should be noted that model tests and field studies of polyester and guar gum applications are needed to confirm the results of this study. It is also recommended that a detailed cost analysis of the use of polymers be made, taking into consideration the relative costs of cement, lime, and the polymers considered for use. REFERENCES AHMED, N. B., 1995, Chemical stabilization of Baiji sand dunes in Iraq. 1. Effect of some soil stabilizers on the infiltration rate of sand: Qatar University Science Journal, Vol. 15, No. 1, pp. 109–113. AHNBERG, H., 1996, Stress dependent parameters of cement and lime stabilized soils. In Yonekura, R.; Terashi, M.; Shibazaki, M. (Editors), Grouting and Deep Mixing: Proceedings of the 2nd International Conference on Ground Improvement Geosystems, Tokyo, A.A. Balkema, Rotterdam, Netherlands, pp. 387–392. AHNBERG, H., 2006, Strength of Stabilized Soils—A Laboratory Study on Clays and Organic Soils Stabilized with Differ‐ ent Types of Binder: Doctoral Thesis, Lund University, Sweden. AHNBERG, H. AND HOLM, G., 1996, Stabilization of some Swedish organic soils with different types of binder. In Brendenberg, H.; Broms, B. B.; and Holm, G. (Editors), Dry Mix Methods for Deep Soil Stabilization: Balkema, Rotterdam, pp. 101–108. AHNBERG, H.; JOHANSSON, S. E.; PIHL, H.; AND CARLSSON, T., 2003, Stabilizing effects of different binders in some Swedish soils: Ground Improvement, Vol. 7, No. 1, pp. 9–23. AJORLOO, A.M., 2010, Characterization of the Mechanical Behavior of Improved Loose Sand for Application in Soil-Cement Deep Mixing: Doctoral Thesis, University of Lille, France. AKBULUT, R. K.; ZAIMOGLU, A. S.; AND ARASAN, S., 2013, Utilization of unsaturated polyester in improvement of sand with deep mixing method. In 5th Geotechnical Symposium: Cukurova University, Adana, Turkey (in Turkish with an English summary). AL-KHANBASHI, A. AND ABDALLA, S. W., 2006, Evaluation of three waterborne polymers as stabilizers for sandy soil: Geotechnical and Geological Engineering, Vol. 24, pp. 1603–1625. ANDROMALOS, K. B. AND BAHNER, E. W., 2004, The application of various deep mixing methods for excavation support systems. In Johnsen, L. F.; Bruce, D. A.; and Byle, M. J. (Editors), Grouting and Ground Treatment: Proceedings of the Third International Conference on Grouting and Ground Treatment: Geotechnical Special Publication 120, ASCE, New Orleans, Louisiana, United States, pp. 515–526. ARASAN, S. AND NASIRPUR, O., 2015, The effects of polymers and fly ash on unconfined compressive strength and freeze-thaw behavior of loose saturated sand: Geomechanics and Engineering, Vol. 8, No. 3, pp. 361–375. ASTM D 2166 (2013). ATES, A., 2013, The effect of polymer-cement stabilization on the unconfined compressive strength of liquefiable soils: International Journal of Polymer Science, 2013, Vol. 2013, Article number 356214. ATES¸, E., 2008, Optimization of compression strength by granulometry and change of binder rates in epoxy and polyester resin concrete: Journal of Reinforced Plastics and Composites. In press.

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Polymers to Improve Soft Clay GUPTA, S. C.; HOODA, K. S.; MATHUR, N. K.; AND GUPTA, S., 2009, Tailoring of guar gum for desert sand stabilization: Indian Journal of Chemical Technology, Vol. 16, pp. 507–512. HARTLEN, J. AND HOLM, G., 1995, Deep stabilization of soft soils with lime-cement columns. In Wong, K. S.; Zhao, J.; and Goh, A. T. C. (Editors), Proceedings of Bengt B. Broms Symposium on Geotechnical Engineering: Singapore, pp. 173–179. HE, J.; CHU, J.; AND IVANOV, V., 2013, Mitigation of liquefaction of saturated sand using biogas: Geotechnique, Vol. 63, No. 4, pp. 267–275. HORPIBULSUK, S.; MIURA, N.; AND NAGARAJ, T. S., 2005, Claywater/cement ratio identity for cement admixed soft clays: Journal of Geotechnical and Geoenvironmental Engineering, Vol. 131, No. 2, pp. 187–192. JACOBSON, J., 2002, Factors Affecting Strength Gain in Lime-Cement Columns and Development of a Laboratory Testing Procedure: Master Thesis, Virginia Polytechnic Institute and State University. JAMSHIDI, M.; ALIZADEH, M.; SALAR, M.; AND HASHEMI, A., 2013, Durability of polyester resin concrete in different chemical solutions: Advanced Materials Research, Vol. 687, pp. 150–154. JGS 0821-00 (2000), “Practice for making and curing stabilized soil specimens without compaction (Translated Version)”, Geotechnical Test Procedure and Commentary, Japanese Geotechnical Society. KHATAMI, H. R. AND O’KELLY, B. C., 2013, Improving mechanical properties of sand using biopolymers: Journal of Geotechnical and Geoenvironmental Engineering, Vol. 139, No. 8, pp. 1402–1406. LEWSLEY, G., 2008, On the Strength of Saturated Cement-Treated Soil Reconstituted by Wet-Mixing: Master Thesis, University of British Columbia. LIM, S. K.; HUSSIN, M. W.; ZAKARIA, F.; AND LING, T. C., 2009, GGBFS as potential filler in polyester grout: Flexural strength and toughness: Construction and Building Materials, Vol. 23, No. 5, pp. 2007–2015. LIU, S. Y.; ZHANG, D. W.; LIU, Z. B.; AND DENG, Y. F., 2008, Assessment of unconfined compressive strength of cement stabilized marine clay: Marine Georesources and Geotechnology, Vol. 26, No. 1, pp. 19–35. LORENZO, G. A. AND BERGADO D. T., 2004, Fundamental parameters of cement-admixed clay—New approach: Journal of Geotechnical and Geoenvironmental Engineering, Vol. 130, No. 10, pp. 1–9. LORENZO, G. A. AND BERGADO, D. T., 2006, Fundamental characteristics of cement-admixed clay in deep mixing: ASCE Journal of Materials in Civil Engineering, Vol. 18, No. 2, pp. 161–174. MAHER, A.; DOUGLAS, W. S.; YANG, D.; JAFARI, F.; AND SCHAEFER, V. R., 2007, Cement deep soil mixing (CDSM) for solidification of soft estuarine sediments: Marine Georesources and Geotechnology, Vol. 25, pp. 221–235. MARTÍNEZ-BARRERA, G.; MENCHACA-CAMPOS, C.; AND GENCEL, O., 2013, Polyester polymer concrete: Effect of the marble particle sizes and high gamma radiation doses: Construction and Building Materials, Vol. 41, pp. 204–208. MATSUO, T.; NISIBAYASHI, K.; AND HOSOYA, Y., 1996, Studies on soil improvement adjusted at low compressive strength in deep mixing method. In Yonekura, R.; Terashi, M.; and Shibazaki, M. (Editors), Grouting and Deep Mixing: Proceedings of the 2nd International Conference on Ground Improvement Geosystems, Tokyo, A.A. Balkema, Rotterdam, Netherlands, pp. 521–526. MCINTYRE, J. E. (Editor), 2004, Synthetic Fibres: Nylon, Polyester, Acrylic, Polyolefin: Woodhead Publishing Limited, Cambridge, England.

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Technical Note Three-Dimensional Stability Assessment of a Complex Landslide in the Rjecˇina Valley, Croatia CHUNXIANG WANG1 Research Institute for Natural Hazards and Disaster Recovery, Niigata University, Nishi-ku, Ikarashi Ni-no-cho 8050, 950-2181 Niigata, Japan

Key Terms: Landslide, Stability Analysis, Interpolation Method, Simulation

ABSTRACT The Grohovo landslide is located on the northeastern slope in the central Rjecˇina River Valley, and the river mouth is located in the center of Rijeka City, Croatia. It is the largest active landslide along the Croatian coast of the Adriatic Sea. This complex retrogressive landslide was induced by an earthquake that occurred in Rijeka City at the end of the 18th century and was reactivated in December 1996. The Grohovo landslide poses a natural hazard to Rijeka City, and so it was necessary to perform a three-dimensional (3D) slope stability assessment. Slope deposits in the Grohovo landslide are mainly a mixture of clayey silt that was formed by weathered Paleogene flysch rock and fragments of limestone originating from the cliffs on the top of the slope. The basal failure surface is located at the contact between the slope deposits and the flysch bedrock. Based on borehole data, geological mapping, and the reconstruction of geological cross sections through the slope, this study interpolated the 3D shape and position of the sliding surface using a modified inverse distance weighted interpolation method. The entire Grohovo landslide and 12 separate sliding blocks, which have the same 3D sliding surface, were analyzed using a modified 3D slope stability analysis based on a simplified force-equilibrium scheme.

INTRODUCTION The Grohovo landslide is located on the northeastern slope in the central Rjecˇ ina River Valley, and the river mouth is located in the center of Rijeka City, Croatia (Figure 1). It is the largest active landslide along the Croatian coast of the Adriatic Sea. The Valic´i reservoir, which was created by a 35-m-high 1

Corresponding author email: chunxiangwang@hotmail.com

concrete gravity dam is located just upstream of this landslide. The most unstable part of the broad Rijeka region lies between the Valic´i reservoir and the Pašac bridge, approximately 1.6 km downstream of the Grohovo landslide. Here, mass movements occur mainly at the contact of fractured and karstified carbonate rocks with the flysch rock complex (Benac et al., 2005). The Grohovo landslide is not just a recent phenomenon in this area; numerous historical records and maps of landslides in the area surrounding the village of Grohovo in the Rjecˇina River Valley were found in the Croatian State Archives in Rijeka (Vivoda et al., 2012; Mihalic´ and Arbanas, 2013). These historical documents indicate that the Grohovo landslide was caused by an earthquake in 1750, with its epicenter in Rijeka. A recent large displacement was observed on December 5, 1996 (Benac et al., 2005). After the initial landslide displacement of the 1996 failure, there was a retrogressive development from head-to-toe as well as the formation of smaller landslides. During reactivation, the sliding mass completely buried the Rjecˇina River channel, and the landslide deposit formed a dam and a lake. After it stopped moving, the landslide mass was immediately removed from the river channel, which eliminated the risk of dam collapse and potential flooding that could have caused serious damage to Rijeka City. Cretaceous and Paleogene limestones are located on the top of the slope, whereas Paleogene flysch is found on the lower slope and in the bottom of the valley (Figure 2). Unlike the limestones, the flysch rock mass is more prone to weathering, particularly the calcitic silts, marl, and shales, which are the predominant rock types (Benac et al., 2005). As a consequence of weathering, a clayey weathering zone formed on the surface of the Eocene flysch bedrock (Eocene flysch bedrock of middle and late Eocene age is labeled as E2,3). Over time, the coarse-grained fragments originating from the cliff on the top of the slope were mixed with clay from the weathered flysch zones, forming slope deposits a few meters thick. Based on field investigations and monitoring, Benac et al. (2002, 2005) indicated that the complex landslide can be divided into 12 sliding blocks with different surface

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Figure 1. Location of the Grohovo landslide.

displacements. The thickness of the displaced material has been estimated from geologic mapping and geophysical surveys. The failure surface is considered to be located at the interface between the slope deposits and flysch bedrock. The Grohovo landslide poses a natural hazard to Rijeka City, which is the largest Croatian port situated on the northeastern Adriatic coast. Sliding is highly probable during unfavorable hydrogeological conditions (Benac et al., 2002, 2005). In order to determine the hazard and risk of potential of landslide occurrences in the future, it is necessary to perform three-dimensional (3D) slope stability analyses. Slope deposits in the Grohovo landslide are mainly a mixture of clayey silt that was formed by the weathered Paleogene flysch rock and fragments of limestone originating from the cliffs on the top of the slope. The basal failure surface is located at the contact between the slope deposits and the flysch bedrock. Based on borehole data, geological mapping, and the reconstruction of geological cross sections through

the slope, this study interpolates the 3D shape and position of the sliding surface using a modified inverse distance weighted interpolation method. The entire Grohovo landslide and 12 separate sliding blocks, which have the same 3D sliding surface, are analyzed using a modified 3D slope stability analysis based on a simplified force-equilibrium scheme. INTERPOLATED 3D SLIP SURFACE The shape and position of the 3D slip surface are important for determining slope stability. Various methods have been proposed in the past to simulate a slip surface. These methods can be classified into two categories. One approach assumes the slip surface is symmetrical, for example, cylinder plus conical surface (Hovland, 1977), ellipsoid end surface (Chen and Chameau, 1982; Ugai, 1985), other different end surfaces (hyperbola, straight line, exponential, parabola; Gens et al., 1988), or an ellipsoidal surface (Zhang, 1988).

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Figure 2. Grohovo landslide map with positions of vertical profile lines: (a) map of slope morphology, (b) vertical profile line 1-2, and (c) vertical profile line 7-8.

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Another approach assumes the shape of the slip surface is arbitrary. Thomaz and Lovell (1988) recommended a general routine for random generation of a 3D sliding sur‐ face. Yamagami and Jiang (1997) combined dynamic programming and random number generation to determine the critical 3D slip surface and the corresponding minimum factor of safety. Lam and Fredlund (1993) used the Kriging method to generate a 3D sliding surface based on random points in space. Depending on the type of landslide and geological features, the failure surface of a landslide exhibits a complex shaped-sliding surface involving a curved and planar surface. Where asymmetrical conditions prevailed, using a 3D method based on a symmetrical sliding surface, force, and/or moment equilibrium led to incorrect safety evaluations (Huang and Tsai, 2000). In this study, the irregular 3D shape and position of the sliding surface are interpolated using a 3D interpolation method (Zhang, 1995). Based on this technique, a smooth surface of designated points can be generated from discrete points in space through a series of cross sections. Most of the 3D geological data, such as geologic contact surfaces and groundwater table field, can be obtained only by discrete sampling. Suppose there is a point i (xi, yi, zi) where a physical datum Ai is recorded. The value of Ai then has an influence on its surroundings. Generally, the influence value W(ri) at ri can be written as follows. W ðri Þ ¼ Ai

r2i r2i r2i ; ln þ 1 R2 R2 R2

(1)

where r2i ¼ ðx xi Þ2 þðy yi Þ2 ; and R is the influence radius. The influence value W(ri), which decreases as ri increases, ranges from Ai to zero, and when ri 5 0 or ri 5 R, dW(ri)/dri 5 0. Given a set of n measurement points (xi, yi, zi) (i 5 1,2,…,n), which represent the sample data obtained from a surface, the fitting function can then be made by adding Eq. 1 N times with different independent ri and different coefficients Ai as follows: Z ðx; yÞ ¼

N X i¼1

r2i r2i r2i : ln þ 1 R2 R2 R2

(2)

In order to determine the coefficient Ai, it is necessary to establish a set of N linear equations in N unknowns. Zij ¼

N X j¼1

! r2ij r2ij r2ij ln þ1 2 ; R2 R2 R

(3)

2 2 where r2ij ¼ xi xj þ yi yj ; i,j 5 1,2,…,N; and rij is the distance from the sample data to the interpolation point. Thus, Eq. 2 describes a spatial curved surface that passes through all of the sample data and is continuous and smoothed everywhere. The interpretation method outlined above was developed for a system prototype, referred to as 3D-GMSSA (Geological Modeling and Slope Stability Analysis), by using the C++ programming language and OpenGL graphic library on a Windows 7 environment. During modeling of the Grohovo landslide, the basal failure surface is considered to be located at the contact between the slope deposits and the flysch bedrock. The locations of the cross sections and two vertical profiles are shown in Figure 2. Figure 3a is a sketch of the 3D landslide, which illustrates the sliding surface and the boundary line. Figure 3b shows the control points along the cross sections and the entire landslide boundary. Figure 3c shows the simulated slip surface using the interpolation method on the basis of the control points in the vertical cross sections located within the Grohovo landslide and the elevation points of the ground surface.

MODIFIED 3D EQUIVALENT OF THE JANBU SIMPLIFIED METHOD The 3D stability analysis becomes important when the slope geometry varies significantly in the lateral direction, the slip surface has irregular 3D configurations, the material properties are inhomogeneous and anisotropic, or the slope is locally surcharged (Chang, 2002). The 3D limit equilibrium approaches are widely used for 3D slope stability analyses. The sliding mass is divided into a number of columns with vertical interfaces. Differences between each method are based on the arbitrary assumptions made regarding intercolumn forces. In most of the 3D limit equilibrium methods, all of the sliding mass columns are assumed to move in a single direction (the sliding direction); namely, the acting direction of the shear force mobilized at the base of each column is parallel to the sliding direction. However, one of the main characteristics of 3D slope failures is that the direction of the shear force or the inclination of each column is different along the sliding surface. The sliding direction of the potential sliding mass is determined by the average dip direction of the slip surface and topographic boundary constraints. The direction of shear force S on the column base, as shown in Figure 4a, is defined as h1 (Figure 4b). Here, cos h1 ¼

sin ax ; sin b

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(4)

Stability Assessment of a Complex Landslide, RjecË&#x2021; ina Valley, Croatia

Figure 3. The interpolated sliding surface by control points: (a) sketch of the 3D landslide, (b) control points, and (c) the simulated 3D sliding surface.

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Figure 4. 3D sliding model and forces acting on a column: (a) forces acting on a column, and (b) geometric characteristics of sliding surface of a column.

where b is the dip angle, and cos b 5 (1/tan2ax + tan2ay + 1)1/2 has been derived by Hovland (1977). If h1 5 0 for each column, then ax 5 b, which implies that the direction of the shear force is parallel to the average sliding direction. This case is the method developed by Hungr et al. (1989). The forces acting on the various faces of each column are shown in Figure 4A. This paper also assumes that vertical shear forces acting on both the longitudinal and the lateral vertical faces of each column can be neglected in the equilibrium condition (Bishop, 1955; Hungr, 1987; and Hungr et al., 1989). In reference to Figure 4A, the total normal force N acting on the base of each column is derived from the vertical force equilibrium equation, W ¼ Nz þ Sz

ðN mÞ tan u cA ¼ N cos b þ þ sin b; Fs Fs

(5)

where W is the weight of the column, Sz is the z component of the shear force acting on the column base, m is the pore-water force acting in the center of the column base, A is the base area, c is the cohesion, φ is the frictional angle, and Fs is the 3D factor of safety.

From Eq. 5, we obtain W N¼

cA sin b uA tan u sin b þ Fs Fs ; ma

(6)

where ma 5 cos b[1 + (tan φ tan b/Fs)], and the area of the column base, A, is

A ¼ DxDy

ð1 sin2 ax sin2 ay Þ1=2 : cosax cosay

(7)

Hungr et al. (1989) used the horizontal force equilibrium in the direction of motion to drive the safety factor. Here, the force equilibrium (Eq. 8) for each column along the slip surface in the sliding direction is used to derive the 3D safety factor. Sz cos h1 ¼ ðPxa Pxb þ Txb Txa Þ cos ay þ W sin b cos h1 ;

(8)

where Pxa and Pxb are inter-column normal forces in the x direction, and Txa and Txb are lateral intercolumn shear forces in the x direction.

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Figure 5. Grohovo landslide: 12 separate landslide blocks.

P For all the columns, ðPxa Pxb þ Txb Txa Þ ¼ 0; and the 3D safety factor can be expressed as P Fs ¼

ððN mAÞ tan u þ cAÞ cos h1 sec ay P : W sin b cos h1 sec ay

(9)

The calculation of the 3D safety factor using the equations described in this section has been coded into the 3D-GMSSA software. 3D STABILITY ANALYSIS OF THE GROHOVO LANDSLIDE The complex Grohovo landslide consists of 12 individual landslide blocks, and the geometry of the entire landslide is shown in Figure 5. The landslide blocks labeled 1, 3, 4, 8, 9, 10, and 11 are located in the upper part of the slope. The landslide blocks labeled 2, 5, 6, 7, and 12 are located in the lower part of the slope. The landslide displacement started with block 7, followed by landslide blocks 6, 12, and 5 and then retrogressive landslide blocks 2, 11, 4, 1, 10, 9, 8, and 3. Borders between different landslide blocks were identified by monitoring surface displacements on the site (Benac et al., 2002, 2005, 2014). According to the boreholes and geological cross sections, the 12 separate sliding blocks are considered to have the same slip surface at the contact between the slope deposits and the flysch bedrock. During 3D analysis, the sliding

directions were determined as the angle between the movement direction and the direction of the x axis. According to Benac et al. (2005), two soil samples from the landslide were collected for laboratory testing to provide specimens for determination of their mineralogical, physical, and geotechnical properties. One of the specimens, S1, was taken from drill cores near the sliding surface at the upper part of the slope, and a second specimen, S2, was taken from drill cores near the sliding surface at the lower part of the slope. The residual shear strength parameters of the soil samples were obtained from ring shear testing and are listed in Table 1. The groundwater pressure is also considered in this analysis. The groundwater flow in the bedrock varies from very rapid to slow depending on the type of deposit. The quantity of water in the superficial deposit zone fluctuates depending on rainfall conditions, which may become surface flow after periods of intense precipitation. There is evidence of water flow at the contact between the bedrock and overlying disturbed material of the rupture zone (Benac et al., 2005). Groundwater levels in the lower part of the slope were derived from water levels measured from piezometers. There were no piezometers at the upper part of the slope, and the groundwater level was assumed to be 1–3 m above the contact between the bedrock and the overlying disturbed materials. Using the soil physical and geotechnical parameters that are listed in Table 1, the 3D slope safety factors

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Wang Table 1. Physical and geotechnical parameters.

Table 2. 3D safety factors (SF3D) of the entire landslide and 12 separate landslide blocks.

Soil Sample

Unit Weight γ (kN/m3)

Cohesion c (kN/m2)

Frictional Angle φ (u)

S1 S2 Flysch bedrock

20 20 21

7.5 16 25

25 16 32

of the complex landslide and 12 separate local landslide blocks were calculated and are listed in Table 2. The calculations indicate that landslide block 3, located near the top of the slope, has a minimum safety factor. This block is considered unsafe because new displacements are expected here in the future, e.g., triggered by heavy rainfall and unfavorable hydrogeological conditions, or a strong earthquake. The 3D safety factor of sliding block 10, located in the middle of the landslide, is greater than 1.2 due to the gentle incline of the base slip surface. The global safety factor of the complex landslide is 1.06, which indicates that the reactivated landslide is in a quasi-stable state.

DISCUSSION AND CONCLUSION The Grohovo landslide on the northeastern slope of the Rjecˇina River Valley is the largest active landslide along the Croatian coast of the Adriatic Sea. Mapping and landslide characterization revealed that there are 12 separate landslide blocks, which have the same base slip surface. The basal failure surface is at the interface between the slope deposits and the flysch bedrock. In order to assess the potential hazards and risks of potential landslides in the future, 3D stability analyses were performed using a modified 3D slope stability analysis based on a simplified force-equilibrium scheme. All slope failures are 3D in nature, and one of the main characteristics of 3D slope failures is that the sliding surface is asymmetric in the direction perpendicular to the movement. While arbitrarily shaped failure surfaces based on field data are seldom considered, the critical 3D slip surface shape is mainly spherical or ellipsoidal for 3D slope stability analyses based on limit equilibrium methods. This paper used an interpolated function to simulate the sliding surface. Based on an interpolation using borehole data, seismic refraction data, geological mapping, and adapted geological cross sections of the slope, the 3D shape and position of the slip surface were generated. Another major characteristic of 3D slope failures is that the direction of shear force or the inclination of each column is different along the sliding surface. In this study, the shear force direction at the column

Landslide Block

Sliding Direction (u)

SF3D

entire_landslide block_1 block_2 block_3 block_4 block_5 block_6 block_7 block_8 block_9 block_10 block_11 block_12

−57 −64 −77 −57 −61 −62 −58 −63 −60 −58 −58 −45 −45

1.06 1.16 1.14 1.02 1.16 1.12 1.07 1.06 1.11 1.16 1.21 1.13 1.08

base is defined as cos h1 5 (sin ax/sin b). However, in the 3D extended Janbu simplified method developed by Hungr et al. (1989), h1 5 0, which is equivalent to ax 5 b in the method presented in this study. Using the 3D method presented here, the global 3D safety factor of the complex landslide and the safety factors of 12 separate landslide blocks were analyzed. The stability analysis indicates that landslide block 3, which is located near the top of the slope, has a minimum safety factor, suggesting that new displacements can be expected here in the future, especially those triggered by heavy rainfall and unfavorable hydrogeological conditions, or by strong earthquakes. As a part of the joint research activities for the Japanese-Croatian scientific project on “Risk Identification and Land-Use Planning for Disaster Mitigation of Landslides and Floods in Croatia,” which was launched in 2010, a comprehensive integrated real-time monitoring system has been installed in the Grohovo landslide. The monitoring system consists of geodetic and geotechnical monitoring (Arbanas et al., 2012). More attention should be given to the movements of the upper part of the Grohovo landslide. Part of the geotechnical monitoring is ongoing measurement of pore pressures in the slope, which will provide better understanding of groundwater variations in the slope and enable more accurate slope stability analysis. ACKNOWLEDGMENTS The research presented in this paper was carried out within the Japanese-Croatian joint research project on “Risk Identification and Land-Use Planning for Disaster Mitigation of Landslides and Floods in Croatia,” which was funded by the Japan Science and Technology Agency–Japan International Cooperation Agency (JST-JICA) Science and Technology Research Partnership for Sustainable Development

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Stability Assessment of a Complex Landslide, Rjecˇ ina Valley, Croatia

Project (SATREPS). The author thanks the three reviewers for their thorough and thoughtful comments, which significantly improved the paper.

REFERENCES ARBANAS, Ž.; SASSA, K.; MARUI, H.; AND MIHALIC´, S., 2012, Comprehensive monitoring system on the Grohovo landslide, Croatia. In EBERHARDT, E. B., (Editor), Landslides and Engineered Slopes: Protecting Society through Improved Understanding, Vol. 1, Taylor and Francis Group, London, pp. 1441–1447. BENAC, Cˇ.; ARBANAS, Ž.; JARDAS, B.; KASAPOVIC´, S.; AND JURAK, V., 2002, Complex landslide in the Rjecˇ ina River Valley (Croatia): Results and monitoring. In RIBAR, J.; STEMBERK, J.; and WAGNER, P. (Editors), Landslides, Proceeding of the 1st European Conference on Landslides: A. A. Balkema Publishers, Lisse, The Netherlands, pp. 487–492. BENAC, Cˇ.; ARBANAS, Ž.; JURAK, V.; OŠTRIC´, M.; AND OŽANIC´, N., 2005, Complex landslide in the Rjecˇ ina River Valley (Croatia): Origin and sliding mechanism: Bulletin of Engineering Geology and Environment, Vol. 64, No. 4, pp. 361–371. BENAC, Cˇ.; OŠTRIC´, M.; AND JOVANCˇ EVIC´, D. S., 2014, Geotechnical properties in relation to grain-size and mineral composition: The Grohovo landslide case study (Croatia): Geologia Croatica, Vol. 67, No. 2, pp. 127–136. BISHOP, A. W., 1955, The use of the slip circle in the stability analysis of slopes: Geotechnique, Vol. 5, pp. 7–17. CHANG, M., 2002, A 3D slope stability analysis method assuming parallel lines of intersection and differential straining of block contacts: Canadian Geotechnical Journal, Vol. 39, pp. 799–811. CHEN, R. H. AND CHAMEAU, J. L., 1982, Three-dimensional limit equilibrium analysis of slopes: Geotechnique, Vol. 32, No. 1, pp. 31–40. GENS, A.; HUTCHINSON, J. N.; AND CAVOUNIDIS, S., 1988, Threedimensional analysis of slides in cohesive soils: Geotechnique, Vol. 38, No. 1, pp. 1–23.

HOVLAND, H. J., 1977, Three-dimensional slope stability analysis method: Journal Geotechnical Engineering Division ASCE, Vol. 103, No. GT9, pp. 971–986. HUANG, C. C. AND TSAI, C. C., 2000, New method for 3D and asymmetrical slope stability analysis: Journal of Geotechnical and Geoenvironmental Engineering, Vol. 126, No. 10, pp. 917–927. HUNGR, O., 1987, An extension of Bishop’s simplified method of slope stability analysis to three dimensions: Geotechnique, Vol. 37, pp. 113–117. HUNGR, O.; SALGADO, F. M.; AND BYRNE, P. M., 1989, Evaluation of a three dimensional method of slope-stability analysis: Canadian Geotechnical Journal, Vol. 26, pp. 679–686. LAM, L. AND FREDLUND, D. G., 1993, A general limit equilibrium model for three-dimensional slope stability analysis: Canadian Geotechnical Journal, Vol. 30, pp. 905–919. MIHALIC´, S. AND ARBANAS, Ž., 2013, The Croatian–Japanese joint research project on landslides: Activities and public benefits. In SASSA, K., et al. (Editors), Landslides: Global Risk Preparedness: Springer-Verlag, Heidelberg, pp. 335–351. THOMAZ, J. E. AND LOVELL, C. W., 1988, Three-dimensional slope stability analysis with random generation of surfaces. In Bonnard, C., (Editor), Proceedings of the 5th International Symposium on Landslides, Vol. 1, A.A. Balkema, Rotterdam, The Netherlands, pp. 777–781. UGAI, K., 1985, Three-dimensional stability analysis of vertical cohesive slopes: Soils and Foundations, Vol. 25, No. 3, pp. 41–48. VIVODA, M.; BENAC, Cˇ.; ŽIC, E.; ÐOMLIJA, P.; AND DUGONJIC´, J. S., 2012, Geohazards in the Rjecˇ ina Valley in the past and present: Croatian Waters, Vol. 20, pp. 105–116. YAMAGAMI, T. AND JIANG, J. C., 1997, A search for the critical slip surface in three-dimensional slope stability analysis: Soils and Foundations, Vol. 37, No. 3, pp. 1–16. ZHANG, J. M., 1995, Design and display of three-dimensional geological model. In Liu, C., (Editor), Advancement of Chinese Mathematical Geology: Press of Geology, Beijing, pp. 158–167. ZHANG, X., 1988, Three-dimensional slope stability analysis of concave slopes in plan view: Journal Geotechnical Engineering Division ASCE, Vol. 114, No. 6, pp. 658–671.

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Assessment of the Mechanism of a Slope Failure in a Hydroelectric Power Plant Site and Considerations on Some Remedial Measures ERGÜN TUNCAY 1

RESAT ULUSAY

Hacettepe University, Geological Engineering Department, 06800 Beytepe, Ankara, Turkey

Key Terms: Slope Instability, Hydroelectric Power Plant, Planar Failure, Back-Analysis, Remedial Measure

ABSTRACT As a result of its mountainous nature, many hydro‐ electric power stations in the Black Sea region are constructed on rock slopes that require a high degree of engineering. Serious negative impacts related to construction safety, such as slope instabilities, that make the risk encountered excessive have also been experienced during the construction and operation stages of these plants. In this study, the mechanism of a slope instability, which occurred in the area of a stream-type hydroelectric power plant and threatened its loading pool and power house, and triggering factors were investigated and the applicability of some remedial measures were assessed. An assessment of the geological and engineering data from the study site indicated that the instability has developed in a certain part of the slope cut along the bedding plane in a sedimentary sequence, where the dip directions of the slope and bedding planes coincide. The back-analysis of the instability proved that the residual shear strength of the bedding planes was a critical factor. A series of analyses conducted along different slope profiles indicated that the most critical mode of failure in the study site would be planar failure, while circular failures through the rock mass and colluvial deposits are not anticipated. Some remedial measures, such as slope flattening, the use of rock bolts, and slope rotation, were also assessed based on preliminary analyses.

INTRODUCTION As a result of increasing demand for energy, efforts to find alternative energy sources continue. In addition to nuclear power plants and attempts in the field of 1

Corresponding author email: resat@hacettepe.edu.tr

wind energy, the construction of small-scale hydropower plants on rivers has been accelerated in recent years. Such plants have become very common particularly in the Black Sea region of Turkey. One of the primary energy sources in Turkey is hydropower. Hydroelectric power plants (HPPs) in Turkey had been constructed through tendering processes by the governmental organization called “General Directorate of State Hydraulic Works.” However, based on the laws known as “Built-operate-transfer, Privatization and Energy Market,” this sector has now also been opened to private companies. The installed power of the river- and channel-type HPPs planned for the entire region of Turkey, with an approximate number of 2,000. are expected to provide up to 25,000 Mw of power, and annual average production is going to be about 125,000 Gwh. The projects will meet only 5 percent of the country’s demand in 2023, the year during which the projects will presumably be complete (Özgür, 2012). The number of HPPs in Turkey, particularly those constructed on the major rivers and streams in the Black Sea region, is increasing day by day. As a result of its mountainous nature, many of the HPPs in the Black Sea region are constructed on rock foundations, forming high engineering rock slopes. Serious negative impacts related to construction safety, such as slope instabilities, which make the risk encountered excessive, have also been experienced during the construction and operation stages of these HPPs. One of the HPPs, which is located 1 km southeast of the town of Gökçebey in Zonguldak Province in the western part of the Black Sea region (Figure 1), is a stream-type hydroelectric power plant (without dam), in which the stream was directed into a tunnel to provide an elevation by which to gain a high flow rate, which results in potential energy. A certain fall to the tribune generates electric energy at the lower edge of the scheme, where the powerhouse is located. Its construction began in October 2007 and was completed in May 2011. With its installed capacity of 36.75 MWm, this HPP, operated by a private company, generates

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Figure 2. A view of the power house (PH), loading pool (LP), and cut slope (CS).

Figure 1. Location map of the investigated HPP site.

141,200,000 kWh of renewable energy annually (www. aksugroup.com) using the flow of Filyos Stream. The HPP consists of a regulator with a height of 12 m from the talveg, a spillway, a headwater channel with a height of 6.6 m, a headwater tunnel with a diameter of 6.2 m, a loading pool (LP), and a power house (PH). A view from the PH, LP, and the cut slope (CS), which is behind these two structures, is provided in Figure 2. The face of the CS has been covered with shotcrete. During the operation stage of the HPP an instability affecting the central part of the CS occurred in March 2013. As can be seen from Figure 3a, the failure initiated from the crest of the CS, involved the uppermost four benches, and moved toward the lower elevations, where the LP takes place. Therefore, the instability threatened the safety of the LP and the PH (Figure 3b) and a part of the protected area adjacent to the southern boundary of the HPP site. This study mainly aims to investigate the mechanism of the CS failure that occurred at the HPP site as well as the contributing factors and to assess possible remedial measures, which can be adopted to cope with the stability problem of the site. In order to reach that goal, in the first stage, field observations and studies, which include geotechnical boreholes, discontinuity and rock mass characterization, measurements

Figure 3. (a) A general view of the cut slope instability; (b) sliding material reached to the loading pool and the power house.

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Figure 4. Geological map of the investigated HPP site and its vicinity (rearranged from MTA, 2002).

along the failure surface, preparation of slope profiles from the failed area using topographical techniques and sampling from the slope forming materials, and failure surface in the unstable slope and its close vicinity, were conducted. Then laboratory tests were carried out on the collected samples to determine the geomechanical properties of the slope-forming materials and discontinuities, which are necessary for slope stability assessments. Finally, back-analysis of the failure and then a series of stability analysis were performed to assess the failure mechanism and contributing factors and possible remedial measures for the investigated slope, respectively.

GEOLOGY AND ENGINEERING GEOLOGY OF THE HPP SITE Geology As seen from the geological map in Figure 4, various formations of sedimentary, volcanic, and magmatic origin, with geological ages ranging from Precambrian to Quaternary, are observed in the vicinity of the HPP. The HPP takes place on/in the Akveren Formation, which is observed on the eastern valley slope of

the Filyos Stream (Figures 2 and 4). This Upper Campanianâ&#x20AC;&#x201C;Lower Eoceneâ&#x20AC;&#x201C;aged formation starts with sandy carbonates at the bottom of the sequence and consists of clayey limestone, reef limestone, marl, sandstone, conglomerate, and volcanics toward its top (MTA, 2002). However, in the PH site and through the failed CS, this formation is represented only by marl and clayey limestone. The slope investigated has been excavated in the alternating sequence of white-beige limestone and greenish-gray marl beds (Figure 5a). However, in the east of the failed CS, a limestone-marl alternation, which is partly weathered and has partly blocky nature and occasionally evident traces of bedding, is also observed above this alternation (Figure 5b). It is also clear from Figure 5a that the marl in the alternating sequence is the dominant lithology forming the CS when compared to the limestone. In the uppermost elevations in the southern part of the failed slope, these rock units are overlain by the Quaternary-aged and brown-colored colluvial deposits consisting of unconsolidated material with particle sizes of fine grains to angular rock fragments, which are present in sizes ranging from a few millimeters to about 3 cm. The fine material surrounding the rock fragments is

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Figure 5. (a) A view of the alternating marl-clayey limestone layering taken during the construction stage of the power house, (b) partly weathered zone at the top, (c) a view of the colluvial deposits affected by the instability at the crest of the slope, and (d) cores taken from the colluvial deposits and alternating sequence of the Akveren Formation (borehole S5).

the dominant component of the colluvial deposits and is represented by silt and clay-sized grains. In other words, these deposits can be classified as matrixsupported. The colluvial deposits have also been affected by the slope instability occurring in the HPP site (Figure 5c). The data from a total of eight geotechnical boreholes, which were drilled during this study, indicated that the depth of the colluvial deposits ranges between 2 and 4 m (Figure 5c and d), and it becomes thinner toward the south behind the crest. Engineering Geology In order to collect geotechnical data and to assess the engineering geology of the investigated site and the mechanism of the CS instability, field studies, including geotechnical boreholes and logging, engineering geological characterization of the discontinuities and

rock mass, construction of the slope instability profiles, and sampling for laboratory tests, were conducted. In the field program, a total of eight geotechnical boreholes with depths ranging between 12 and 25 m were drilled. Six of them were drilled behind the crest of the failed slope, while two boreholes were drilled on two different benches at the eastern part of the CS, as shown in the plan of the site in Figure 6. The boreholes reached to the elevations that were lower than that of the observed sliding surface, and their depths ranged between 12 and 25 m. Geotechnical logging of the boreholes indicated that, except at its occasionally observed uppermost weathered levels, the RQD values of the alternating marl-limestone sequence mainly forming the CS generally range between 60 percent and 90 percent, indicating that as a rock mass the rock sequence does not have a heavily jointed nature, and discontinuities can be considered as an important

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Figure 6. Plan view of instability features and locations of engineering geological site investigations.

governing factor for slope stability (see Figure 5d). Geotechnical borehole data also confirmed that the colluvial deposits directly over the marl-limestone alternation. However, it is also noted from the borehole data that at a few borehole locations, such as borehole S1 in the south behind the crest and boreholes S7 and S8 in the east (Figure 6), the marl-limestone alternation overlain by the colluvial deposits starts with a zone, which is composed of beige-oliveâ&#x20AC;&#x201C; and oil-brownâ&#x20AC;&#x201C;colored material partly transformed into a soil as a result of weathering and partly preserved by its bedded structure, and its thickness ranges between 2 and 10.6 m. The RQD values in this zone are between 3 percent and 33 percent. These properties suggest that this zone can be considered as partly heavily jointed rock mass and partly highly weathered rock. The

alternating fresh marl and clayey limestone appear below this zone. In addition to discontinuity orientation (dip/dip direction) measurements taken from the observation windows excavated in the shotcreted CS surfaces using a geologist compass, the dip of the bedding planes was also measured on the borehole cores using a goniometer. Since in this study no core orientation has been conducted through the boreholes, only the measurement of the dip of strata from the cores was possible. By considering the observations in the study site indicating that the dip of the bedding planes tends to decrease behind the crest of the failed CS, where six boreholes (S1 to S6) were drilled, the dip of the bedding planes in the marl-limestone alternation was evaluated in two separate groups: dips measured

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Figure 7. Histograms for dips of the bedding planes observed in boreholes (a) S1 to S6 and (b) S7 and S8 and (c) those of the joints observed in all boreholes; (d) nearly vertical joints (shown by arrows).

from boreholes S1 to S6 and those measured from boreholes S7 and S8. The histogram in Figure 7a suggests that the dip of bedding planes behind the slope crest ranges between 17u and 25u, with a mean of 21u, while the bedding dips measured from the cores of boreholes S7 and S8, which are located very close to the failed part of the slope (Figure 6), vary from 26u to 32u, with a mean of 28u (Figure 7b), indicating that the values of dip are greater than those measured in the boreholes drilled behind the slope crest. These higher dip amounts are consistent with those measured on the sliding surface. These assessments indicated that the bedding dip decreases behind the crest of the failed slope, while the failure occurred

in certain parts of the slope where the bedding dip increases. The dip amounts of the two main joint sets identified on the borehole cores were also measured. It is clear from Figure 7c that their dips range between 47u and 73u, with a mean of 63u, indicating the presence of steep joints. These joints observed on the cores likely represent the highly steep joints, which are nearly perpendicular to each other and are observed at the upper part of the sliding surface. They help separate the rock slabs bounded by them and the bedding planes into prismatic blocks (Figure 7d). It is also noted that no groundwater table has been encountered through all of the boreholes. In addition,

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Figure 8. (a) A view of an observation window excavated into shotcrete (the pen is aligned along a bedrock feature), (b) stereographic projection of the bedding planes in the alternating sequence of marl-clayey limestone layers (the great circle in the projection represents bedding planes), (c) the sliding surface along bedding plane, and (d) spacing between bedding planes.

no seepage was observed in the sliding area. Based on these observations, it was considered that the sliding surface was above the groundwater table and that it should be at deeper elevations, if any. Except for a few parts of the study site, where failure surface can be seen and small rock outcrops are available (since the failed CS has been covered by shotcrete), it is extremely difficult and/or almost impossible to observe the rock types and to geotechnically characterize their discontinuities in the study area. In order to overcome this difficulty, the shotcrete was excavated until the rock surface appeared, and observation windows with a dimension of about 30 6 30 6 10 cm were obtained to describe the rock type and to measure the orientation of the bedding planes from the rock surfaces reached (Figure 8a). However, at the locations of a few observation windows, it was not possible to reach the rock surface because of the presence of locally

thick shotcrete. In addition, surface features of the discontinuities, such as roughness, infill, and type of weathering, were also described according to the International Society for Rock Mechanics (ISRM, 2007) suggested methods. A total of 21 bedding orientations could be measured from the observation windows. The contour diagram for bedding planes depicted in Figure 8b suggests that the orientation (dip/dip direction) of the beddings is 29/321, and there is a small scattering in their orientation. The bedding dips, measured at different locations and marked in Figure 6, ranged between 28u and 32u, and except for very small deviations, their dip directions centered around the mean value of 321. These measured values show a good agreement with the orientation of the bedding plane (30uâ&#x20AC;&#x201C;31u/320) along which the instability occurred (Figures 7b and 8c). In addition, the strike of the CSs is generally in a northeast-southwest direction,

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and the slopes dip toward the northwest, with an overall angle ranging between 30u and 37u (see Figure 6). The bedding planes in the marl-clayey limestone alternation are planar and smooth. However, in some places they are covered by a thin clay coating. Neither weathering effects nor evident undulations on the bedding plane surfaces are observed, while joint surfaces are slightly rough and coated by yellowish-brownâ&#x20AC;&#x201C; colored iron oxide. The spacing between the bedding planes was measured in the sliding area, where bedding planes can be easily observed after the movement has occurred (Figure 8d). The values of the true spacing measured in this part of the slope were 47, 7, 37, 46, 5, 11, 45, and 47 cm, and a mean spacing value of 30 cm was estimated. In addition, spacing was also determined from the photographs, which were taken by the owner of the HPP at the lower elevations of the CS during its construction stage. These spacing measurements are very similar to those measured by the authors and range between 5 and 45 cm. Based on the limited observations by the authors in the study site, the spacing of the joints ranges between 1.5 and 2 m. No circular rock mass failure has been observed in the study site. However, since the height of the slope in the southern part of the site reached to 70 m, the possibility of a rock mass failure and its analysis are considered to be included in the investigation program. In the analysis of such failures, the use of rock mass strength is of prime importance. For this purpose, the Hoek and Brown failure criterion (Hoek et al., 2002) was utilized, and the Geological Strength Index (GSI), which is one of the main inputs of the criterion, was determined, using the quantitative GSI chart (SĂśnmez and Ulusay, 2002), as 51 based on the estimations from the available rock exposures and borehole cores. In order to use geomechanical laboratory tests, first, all core samples representing marl-limestone alternation were collected from the boreholes drilled in this study at the site. Some blocks from the partly weathered and partly jointed zone of the alternation and samples from the sliding surface were also taken. The samples from the sliding surface were obtained with an orientation parallel to the movement direction. In addition, undisturbed samples were extracted from the fine portion of the slope debris utilizing 6 6 6 6 2 cm metal specimen cutters. GEOMECHANICAL LABORATORY EXPERIMENTS The samples collected in the field were used to determine the physical, index, and geomechanical parameters of the bedding planes, which are of paramount

importance in terms of slope stability, alternating rock types, and slope debris and weathered zone. The laboratory experiments were conducted in accordance with the methods suggested by the ISRM (ISRM, 2007) and the procedures of the American Society for Testing and Materials (ASTM, 1994) for discontinuities and rock materials and for soil-like materials (slope debris, weathered zone), respectively. The unit weight of the marl and clayey limestone is the same (24.8 kN/m3), while the weights of the colluvial deposits and weathered zone are 17.2 and 15.9 kN/ m3, respectively. The uniaxial compressive strengths of the alternating marl and clayey limestone are 78.5 and 167.1 MPa, respectively. A mi value of 8 for the alternating rock sequence is found. Based on the Atterberg limits and grain size distribution analyses, the fines portion of the colluvial deposits and the weathered part of the alternating sequence are classified as MH/ CH (highly plastic silt-clay) and ML (silty-clayey fine sand with slight plasticity) class soils, respectively. The anticipated full range of in situ normal stresses acting on discontinuities in a bench or in an overall slope will vary by a few tens and/or hundreds of kilopascals, respectively. Therefore, it was considered essential that this condition should be duplicated in the laboratory in order to obtain reliable results from shear strength determinations. Since the portable shear box is rather sensitive and difficult to use at relatively low normal stresses associated with slope engineering investigations, in this study a motorized direct/residual shear test device with strain control and shear box assembly was employed for shear testing of bedding planes. For each test, specimens from the block samples of the natural bedding planes were cut to fit the shear box with dimensions 6 6 6 6 2 cm, and three test sets, each consisting of four specimens, were prepared. Similarly, shear strength parameters of the colluvial deposits and weathered zone were also determined on four specimens using the same device. A single-stage loading technique was adopted to prevent progressive damage to the specimens, with multiple reversals to achieve residual values. During the shear tests, no dilation was noted for the bedding planes. The majority of the bedding surfaces reached residual values at the end of the first forward motion of the shear box. Peak and residual friction angles of the bedding planes vary between 29.1u and 33.3u and between 24.3u and 27.9u, respectively, and they have no cohesion. To obtain a generalized failure envelope for the bedding planes, the results were transferred to a single plot. Statistical assessments showed that the test data better fit a linear envelope, and shear strength parameters of the bedding planes were derived from linear Mohr-Coulomb failure envelopes. It is clear from the generalized peak and residual failure envelopes and

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Figure 9. Generalized peak and residual failure envelopes for the bedding planes.

shear strength properties given in Figure 9 that the bedding planes have no cohesion. This behavior of the bedding planes in the study area shows good agreement with previous findings that the shear strength behavior of smooth and planar discontinuities show a purely frictional resistance proportional to normal stress (i.e., Udd and Betournay, 1983; Hencher and Richard, 1989; and Singh and Gahrooee, 1989). Shear strength properties of the colluvial deposits and weathered zone are given in Table 1.

TYPE AND MECHANISM OF THE SLOPE INSTABILITY The middle part of the CS failed in the HPP site in March 2013 and affected the central part of this area (Figures 3a and 6). The slope moved in a northwest direction, the failure involved the uppermost four benches of the CS (Figure 10a), and the sliding material reached to the road, passing from an elevation of +100 m, where the LP takes place, and closed the road next to the pool (Figure 3b). Fortunately, the sliding material did not cause any damage to the LP and did not fill it. During this event the shotcrete covering the slope face was also broken into the slabs and moved down (Figure 10b). The maximum height of Table 1. Shear strength parameters of the colluvial deposits and partly weathered material determined in laboratory. Material Colluvial deposits Weathered zone

cp (kPa)

ϕp (u)

cr (kPa)

ϕr (u)

9.2 0

23.0 38.1

6.5 0

17.6 31.7

cp: Peak cohesion; cr: Residual cohesion; ϕp: Peak friction angle; ϕr: Residual friction angle.

the failed part of the slope was about 50 m. Although the instability mainly occurred in the alternating marl and clayey limestone layering, it also involved some parts of the slope debris at its southern tip (Figures 5c and 6). The movement first occurred in the lower benches in the alternating marl-clayey limestone layering, and then the uppermost benches in the colluvial deposits also moved down, following the movement of the lower benches. The failure also resulted in the formation of some tension cracks in the colluvial deposits behind the slope crest. Site investigations and topographical measurements conducted by the authors suggested that it was a planar failure that occurred along the bedding plane with an inclination of about 30u in the alternating marl and clayey limestone sequence (Figure 8c). It is also clear from the photographs taken during the construction stage of the PH (Figure 10c) that the orientation of the bedding planes in the sliding area shows a very good agreement with that of the sliding surface measured in the site and that the strata dips toward the excavation. Using the measurements taken from the sliding area before and after the instability, geotechnical borehole data, and the position of the sliding surface, two cross sections (sections 1 and 2) showing the pre- and postfailure slope geometries and the sliding surface are depicted in Figure 11, and their directions are shown on the site plan in Figure 6. The HPP technical personnel inform the authors that heavy rains occurred in the region between January and March 2013. However, based on the information from the same personnel, no clear evidence and/or pore pressure measurements, indicating a possible contribution of these rains to the instability, were available. When the stable part of the slope located in the east of the failed area (Figure 10d) is considered, it can be

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Figure 10. (a) Sliding surface and the benches affected by the slope movement, (b) broken and deformed shotcrete, (c) bedding planes dipping toward the excavation (provided from the owner), and (d) stable (shown by arrow) and unstable parts of the cut slope.

seen that the difference between the dip direction or strike of the benches and those of the bedding planes are greater than 20u (Figure 6). Based on the rules of the slope kinematics (e.g., Hoek and Bray, 1977; Goodman, 1989), a difference between the strikes of the slope and bedding planes of greater than 20u explains why no planar failure has occurred in this part of the slope. BACK-ANALYSIS OF THE SLOPE FAILURE In the case of stability analysis for actual failed slopes, the shear strengths of material or discontinuity obtained from laboratory tests are sometimes not effective in determining their critical state. The most reliable way to obtain a statistical weighted mean value of shear strength parameters in an extended slope is back-calculation. In order to assess the failure mechanism of the instability that occurred in the south of the HPP site and to estimate the likely shear strength

parameters mobilized along the sliding surface at the time of failure, two-dimensional limit equilibrium back-analysis, assuming the linear Mohr-Coulomb shear strength criterion, was performed. The general approach employed in the back-analyses was based on the following assumptions: a) A condition of static equilibrium at the point of failure exists at the time of failure. That is, slope failure occurs when the factor of safety (FOS) is reduced to unity. b) Since roughness and coating of discontinuity surfaces are generally the same throughout the study site, there was no need for further simplifications in terms of homogeneity and isotropy. c) The shear strength obtained from the analyses is the weighted average shear strength of the sliding surface. Because of the spatial variations in the mechanical properties of the discontinuities (in this case, bedding planes), the back-calculation of cohesion (c) and

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Figure 11. Cross sections showing the pre- and post-failure slope geometries and the sliding surface.

internal friction angle (ϕ) from more than one, and preferably more than two, instability sections may give as many as n(n − 1)/2 points of intersection for n lines or curves of c (ϕ) (Sancio, 1981). For this purpose, cross sections 1 and 2, which are parallel to the movement direction (see Figure 6) and well represent the unstable part of the CS, were employed in these analyses. Post-failure topography along these sections was determined by topographical measurements, and pre-failure topography (slope profile) was provided by the owner of the HPP. Since no groundwater table and seepage were observed through the geotechnical boreholes and on the slope face, respectively, no

groundwater table was considered in the analyses. As can be seen from both cross sections (Figure 11), a very shallow seated separation surface at the tip of the main sliding surface passes through the colluvial deposits and weathered part of the alternating marlclayey limestone sequence. This shallow-seated and steep separation surface observed in both sections has no curvature and developed as a result of the sudden movement of the upper benches in the colluvial deposits down to fill the gap, which resulted from the movement of the lower benches in the alternating marl and clayey limestone sequence. Therefore, it can be concluded that there was not sufficient time for the shear

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Figure 12. Back-analysis results for the bedding planes along which the instability has occurred in the HPP site. (The ranges of peak and residual shear strength are those determined from laboratory testing.)

strengths of the colluvial deposits and weathered zone to drop their residual values, and they probably failed at shear strengths close to their peak values. Based on this conclusion, back-analysis of the failure was conducted by keeping the laboratory-derived peak shear strength of both materials constant and by trying to calculate c-ϕ pairs for the bedding planes that satisfy the limiting equilibrium condition. The analyses were conducted using the computer code SLIDE (Rocscience, 2009) according to the approach suggested by Hoek and Bray (1977) for planar failure. The results of the back-analyses obtained from the two failure sections are depicted in the form of c-ϕ envelopes, with the ranges of laboratory-determined peak and residual c and ϕ values of the bedding planes depicted in Figure 12. Since the bedding surfaces have no cohesion, the ranges of peak and residual ϕ fall on the y-axis. Although the c-ϕ lines in Figure 12 do not intersect each other, they tend to be very close when c approaches zero and better represent the residual shear strength range when compared to that of the peak value. In addition, a very good agreement between the back-calculated (c 5 0 and ϕ 5 26.5u) values at the c 5 0 condition obtained from section 2 (Figure 12) and the laboratory-derived residual shear strength (c 5 0, ϕ 5 26.3u; Figure 9) also suggests that the residual shear strength of the bedding planes has mobilized at the time of failure. Therefore, the use of the laboratory-derived residual shear strength of the bedding planes in further stability analyses seems

more realistic for the assessments of remedial measures in the HPP site.

SOME ATTEMPTS TO ASSESS POSSIBLE REMEDIAL MEASURES Since the presence of a large slope failure, the occurrence of new instabilities in the investigated HPP site seems also possible, and, therefore, the PH and LP are under their threat. Considering this, it became necessary to investigate possible remedial measures to be applied to the failed and stable parts of the CS behind the PH and LP. For this purpose, a series of stability analyses were performed, and possible effects of some remedial measures, such as slope flattening and application of rock bolts, were assessed using twodimensional limit equilibrium methods of analysis. In these analyses, in addition to planar failure, the possibility of circular failure surfaces passing through the rock mass was also considered. The first series of analyses were conducted for the failed central part of the CS using sections 1 and 2. Then second series of analyses were conducted for the western and eastern parts of the slope along sections 3 and 4, respectively (Figure 6), where no slope instability has occurred yet. Additional analyses based on kinematic approach (obtained by changing the strike of the slope) were also performed. These analyses and assessments of their results are briefly given in the following subsections.

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Figure 13. Stability analysis results along section 1: (a) after cleaning the failed material and flattening the slope to 30u, (b) flattening the slope to 25u in the failed part and designing the slope with an overall angle of 30u where the bedding dip decreases to 20u, (c) flattening the slope to 25 degrees both in the failed part and where the bedding dip decreases to 20 degrees, and (d) circular failure surfaces and the center of the surface with lowest FOS for rock mass failure.

Analyses for the Unstable Central Part of the Slope The analyses were first conducted for section 1. The initial analysis along this section was for a condition in which the failed material is cleaned and the failed uppermost part of the slope is benched with an overall angle of 30u (parallel to bedding planes), but no change is considered in the geometry of the benches located at lower elevations, and no rock bolt is applied (Figure 13a). The analysis resulted in a FOS of 0.828, indicating that this alternative cannot provide stability. In the second alternative trial, which considers cleaning the sliding material and flattening the overall angle to 25u in the failed part of the slope and at its south, where the bedding dip is 20u–21u, the overall slope angle was taken to be 30u. As seen from Figure 13b, the values of FOS smaller than unity and very close to 1

are obtained. This result also indicates that a sufficient safety condition cannot be achieved with the application of this alternative. It is also noted that although the overall angle in the upper part of the slope is reduced from 30u to 25u, no sufficient improvement (FOS 5 1.172) in the stability could be achieved (Figure 13c). In addition, the possibility of a circular failure through the rock mass was also investigated. For this purpose, the circular failure option of the program SLIDE (Rocscience, 2009), which considers the Hoek-Brown failure criterion for the estimation of shear strength of the rock mass, was utilized. In these analyses, a total of 4,500 failure surfaces were investigated, and the failure surfaces with the lowest FOS are shown in Figure 13d. The results indicate that the lowest FOS is 5.6, and, therefore, circular failure in

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Figure 14. Stability analysis results along section 2: (a) ďŹ&#x201A;attening the slope to 25u in the failed part and no change in the uppermost benches and (b) ďŹ&#x201A;attening the slope to 25u between the toe of the failed part and the crest of the slope.

this slope is not anticipated. The above-mentioned results suggest that even through the sliding material is removed and that the overall slope angle is reduced to 25u along this section, it is not possible to achieve a FOS of 1.3 to 1.4, which is generally recommended for such critical slopes (Huang, 1983). Flattening the slope to 25u will also not be an economic measure because of high amount of material that should be removed. Thus, it seems better to use reinforcement systems, such as rock bolts, rather than slope flattening. As the first-stage analysis along section 2, a scenario, which is based on cleaning the failed part of the slope and construction of new benches, based on an overall slope angle of 27u (the dip of the bedding planes) without any stripping behind the current crest of the slope, was considered. In the analyses, both the possibility of planar failure along the bedding planes and circular failure through the weathered zone were investigated. As

seen from Figure 14a, the FOS values that are very close to unity (1.039 and 1.1) were obtained for the upper part of the slope. Similarly, the analysis for the lower part of the slope, with an overall angle of 27u (Figure 14a), also yielded a FOS value that was lower than unity, indicating that this alternative cannot provide safety. In the second-stage analysis, starting from the uppermost elevations, where the dip of the bedding planes is 20u, to the toe of the failed section, the overall slope angle is flattened to 25u, and the possibility of a planar failure along this part was investigated. The calculated FOS of 1.6 suggests that stability will be achieved to a reasonable level only in the uppermost part of the slope (Figure 14b). But even if the failed portion of the slope is flattened to 25u along this section the FOS can approach only unity (0.99; Figure 14b), and this part of the slope will preserve its critical condition. The above-mentioned results suggest that starting from the elevation of +120

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m and designing the slope with an overall angle of 25u along this section will result in removal of a large amount of material and shifting of the crest of the slope toward a point that is beyond the legal boundary of the HPP site. In other words, the results from section 2 also call for the evaluation of reinforcement techniques to increase the stability. Analyses for the Parts of the Slope Adjacent to the Failure Given the presence of some local tension cracks observed on some benches between the elevations of 110 and 120 m and 120 and 130 m and the agreement between the dips of the bedding planes and the slope and its closeness to the failed part of the slope, as‐ sessment of stability along the southwest-northeast– directed section 3 (Figure 6) was considered to be useful. In the first series of analyses along this section, based on the current geometry of the benches forming the slope, the values of FOS against planar failure along the bedding planes at different depths were calculated. The results indicate that FOS values lower than unity even for the shallowest sliding surface only affect the single bench (FOS 5 0.836) (Figure 15a). It can be concluded that although an instability problem including some benches has not been experienced in this part of the slope, depending on time, stability may decrease and the slope may be threatened by instability. It is also noted that while no instability has occurred between 2011, when the construction of the HPP was completed, and 2013, the March 2013 failure, which occurred along the bedding planes daylighting on the slope face, can be considered as an indicator of the anticipated probable slide along this section. Then another slope configuration, in which the overall slope angle above an elevation of +130 m is flattened to 25u and the current geometry of the benches below this elevation is not changed, was analyzed. It is clear from the calculated FOS values smaller than unity (Figure 15b) that even when the slope above an elevation of +130 m is flattened to 25u, sufficient stability cannot be achieved, and reinforcement of the benches below this elevation seems necessary. Similar to what was concluded for sections 1 and 2, the analyses carried out for this section also suggested that the minimum FOS values for the circular sliding surfaces passing through the rock mass and colluvial deposits were 5.43 and 2.21, respectively (Figure 15c), and circular failure was not critical.

Figure 15. Stability analysis results along section 3: (a) analyses along the bedding planes passing from different depths for the current slope proﬁle, (b) ﬂattening the slope to 250 in the failed part of the slope and keeping the same geometry of the benches below an elevation of +130 m, and (c) analyses for the circular failures passing through the rock mass and the colluvial deposits.

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Figure 16. Stability analysis results along section 4: (a) Analysis of planar failure along the bedding in the partly weathered zone and between the colluvial deposits and alternating marl-clayey limestone layering and (b) application of rock bolts against planar failure.

Some cracks observed on the shotcrete, the presence of partly weathered and heavily jointed marl-limestone alternation at shallow depths and its closeness to the unstable part of the slope, the slope profile along section 4, close to the northwest-southeast direction (Figure 6), were also analyzed. In the first stage, the current slope geometry along the section was analyzed separately according to its two parts: slope debris and the weak alternation. The analysis results obtained for planar failure (Figure 16a) suggest that no stability problem is anticipated for the part of the slope resting in the slope debris (FOS 5 1.987), while it is approximately equal to unity (FOS 5 0.977) for the slope profile resting in the partly weathered and heavily jointed

alternating sequence. In addition, the stability of the slope in this zone was also investigated against circular failure, and a high FOS of 1.665 was found. This result indicated that planar failure in this part of the slope, represented by section 4, is the only critical mode of failure. In order to preliminarily assess the stabilizing effect of rock bolts against planar failure in this part of the slope, additional analysis was performed. In this analysis, rock bolts (having a reinforcing force of T 5 250 kN), which are 12 m long and installed with 3-m spacing and an inclination of 45u to the bedding planes, were considered (Figure 16b). The comparison of Figures 16a and 16b suggests that in the case of rock bolt application, the FOS

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Figure 17. (a) Kinematic conditions for planar failure (Norrish and Wyllie, 1966) and (b) simpliﬁed plan of the investigated slope showing the current benches and the recommended slope rotation.

considerably increases from 0.997 to 1.321, provid‐ ing an important contribution to stability. However, more detailed analyses based on rock bolts in conjunction with the application of shotcrete and wire mesh will be useful. Improvement of the Stability by Rotating the Slope The main reasons for the instability occurring in the HPP site are the following (which also kinematically satisfy planar failure conditions; Figure 17a):

a) Bedding planes dip into the excavation with inclinations (28u–32u) less than that of the slope (daylighting), and their dip is greater than the friction angle mobilizing along their surfaces. b) The difference between the strike or dip direction of the slope and bedding planes is less than 20u. Since the above-mentioned conditions are not satisfied in other parts of the CS, no evidence of instability has been observed in these parts. However, since the orientations of the bedding planes are very similar

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and their friction angles do not show important variation throughout the site, the main factor that negatively affects the stability is the local changes in the dip directions of the slope with respect to those of the bedding planes. In order to prevent such local movements, the slope can be rotated until a difference of greater than 20u between the strikes of the slope and the bedding planes is obtained. In such a case, even if condition 1 in Figure 17a is valid, condition 2 will not be satisfied and planar failure will not occur. By considering the range of dip direction of the bedding planes (310-335) measured in the study site, the current benches between the western boundary of the site and section 1 can be rotated until their strikes become parallel to the east-west direction (dipping toward the north) and then the strike of the benches between section 1 and the eastern boundary of the slope can be rotated to N10E (dipping toward N80W or 280u), as illustrated by red lines and also shown in the stereographic projection in Figure 17b. However, a final decision on the application of this alternative measure needs further assessments related to the amount of the material that will be removed and its cost. CONCLUSIONS In this study, the mechanism of the CS failure occurred in a HPP site located in the western Black Sea region of Turkey, and contributing factors were investigated and possible remedial measures, which can be adopted to cope with the stability problem of the site, were assessed. An assessment of the geological and engineering data from the study site indicated that the instability has developed along the bedding planes in a sedimentary sequence in a certain part of the slope cut, where the dip directions (or strikes) of the slope and bedding planes coincide. The back-analysis of the instability proved that there is a good agreement between the laboratoryderived residual shear strength of the bedding planes and the back-analyzed shear strength of the sliding surface. A series of analyses conducted along different slope profiles indicated that the most critical mode of slope failure in the study site would be planar failure, while deep- and shallow-seated circular failures through the rock mass and colluvial deposits are not anticipated. These analyses also suggest that slope flattening to improve the stability alone would not be sufficient. In addition, if the failed part of the slope, which is located above an elevation of +130 m, is cleaned and then flattened to 25u and if the other benches located below this elevation are supported by rock bolts, it seems that the stability can be considerably increased. However, it should be remembered that this remedial measure may result in a considerable

amount of overburden removal and/or occupation of a protected area adjacent to the southern boundary of the HPP, where there are some historical ruins. Rotation of the strike of the slope is also another alternative remedial measure. If the strike of the benches located between the western boundary of the site and section 1 is rotated until they become parallel to east-west and the other part of the slope between section 1 and eastern boundary of the site is rotated to have a strike of N10E, the difference between the strikes of the slope and bedding planes will be greater than 20u, and planar failure will not be kinematically possible. However, the amount of the material that will need to be removed can be a restriction for this application. ACKNOWLEDGMENTS The authors thank the company, who is responsible for the operation of the HPP, for its kind permission for the publication of this study. The authors are also grateful to Fatih Adil (Hendese Geotechnique Co.) for his kind permission to use the computer code SLIDE. The authors also acknowledge Richard M. Wooten, one of the three reviewers, and the other two anonymous reviewers for their constructive comments, which led to improvements in the article. REFERENCES AMERICAN SOCIETY FOR TESTING AND MATERIALS (ASTM), 1994, Annual Book of ASTM Standards—Soil and Rock, Building Stones, Section 4, Construction. V.04.08: ASTM Publications, Philadelphia, PA. 978 p. GOODMAN, R. E., 1989, Introduction to Rock Mechanics: John Wiley and Sons, New York. HENCHER, S. R. AND RICHARDS, L. R., 1989, Laboratory direct shear testing of rock discontinuities: Ground Engineering, Vol. 22, pp. 24–31. HOEK, E. AND BRAY, J. W., 1977, Rock Slope Engineering: Institution of Mining and Metallurgy, London, U.K. HOEK, E.; CARRANZA-TORRES, C. T.; AND CORKUM, B., 2002, Hoek-Brown failure criterion—2002 Edition: Proceedings 5th North American Rock Mechanics Symposium, Toronto, Canada, Vol. 1, pp. 267–273. HUANG, Y. H., 1983, Stability Analysis of Earth Slopes: Van Nostrand Reinhold Comp., New York. INTERNATIONAL SOCIETY FOR ROCK MECHANICS (ISRM), 2007, The Complete ISRM Suggested Methods for Rock Characterization, Testing and Monitoring: 1974–2006: R. Ulusay and J. A. Hudson (editors), Suggested Methods Prepared by the Commission on Testing Methods, International Society for Rock Mechanics, Compilation Arranged by the ISRM Turkish National Group, Ankara, Turkey, Kozan Ofset. MTA, 2002, Geological Maps of Turkey With a scale of 1:100,000— No. 29, Zonguldak F28 Sheet: General Directorate of Mineral Research and Exploration, Department of Geological Investigations, Ankara, Turkey (in Turkish). NORRISH, N. I. AND WYLLIE, D. C., 1996, Rock slope stability analysis. In Turner, A. K. and Schuster, R. L. (Editors), Special

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Mechanism of a Slope Failure Report 247: Landslides: Investigation and Mitigation: TRB, National Research Council, Washington, DC, pp. 391–425. ÖZGÜR, M. N., 2012, Hydropower plant safety and landscape in the period of ‘Energy Market Regulation Board’ (EPDK) in Turkey: Proceedings of IWAWCE2012, Paper No. 0143. ROCSCIENCE, 2009, Slide V5.043—2D Limit Equilibrium Analysis: Rocscience, Toronto, Canada. SANCIO, R. T., 1981, The use of back-calculations to obtain the shear and tensile strength of weathered rocks: Proceed‐ ings International Symposium on Weak Rocks, Tokyo, pp. 647–658.

SINGH, R. N. AND GAHROOEE, D. R., 1989, Application of rock mass weakening coefficient for stability assessment for slopes in heavily jointed rock masses: Journal International Surface Mining Reclamations, Vol. 3, pp. 207–219. SÖNMEZ, H. AND ULUSAY, R., 2002, A discussion on the HoekBrown failure criterion and suggested modifications to the criterion verified by slope stability case studies: Yerbilimleri (Earthsciences; www.yerbilimleri.hacettepe.edu.tr), Vol. 26, pp. 77–99. UDD, J. E. AND BETOURNAY, M. C., 1983, Slope stability in Trenton limestone of a Montreal area quarry: CIM Bulletin, Vol. 76, No. 809, pp. 72–78.

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Assessing the Geological Sources of Manganese in the Roanoke River Watershed, Virginia ZACHARY A. KIRACOFE WILLIAM S. HENIKA MADELINE E. SCHREIBER1 Department of Geosciences, Virginia Tech, Blacksburg, VA 24061

Key Terms: Manganese, Ore Deposits, Geochemistry, Geomorphology, Hydrogeology

ABSTRACT The release of geogenic elements to water supplies is an issue of worldwide concern. Because these elements occur naturally and are often present in a variety of geologic materials, delineating sources and fates of these elements can be challenging. In this study, we examine connections between manganese (Mn) in modern groundwater, bedrock geology, and ores in the Roanoke River watershed of the Piedmont Province, Virginia. In the watershed, Mn concentrations in groundwater are often elevated above secondary drinking water standards. Evaluation of chemical characteristics of groundwater and geologic materials within the region suggests that carbonate-bearing lithologies are likely sources of Mn to groundwater. The inverse correlation of Mn with dissolved oxygen concentrations in groundwater suggests that once released from chemical weathering, Mn persists in groundwater under reducing conditions that develop along flowpaths. Analysis of Mn ores of the James Riverâ&#x20AC;&#x201C;Roanoke River Manganese District provides support that the ore deposits are supergene in origin, consistent with previous models. However, in contrast to previous models suggesting ore formation from downward flow of groundwater, our analysis of ore formation in the context of groundwater geochemistry supports an upwelling model of ore formation in which Mn oxides were precipitated near discharge zones. Overall, our results suggest that Mn cycling in the region has been active over geologic time, as Mn-rich groundwater discharges to riverine systems, both past and present. Thus, the processes that formed the Mn ores in the past are still occurring in the modern day.

Corresponding author email: mschreib@vt.edu.

INTRODUCTION Manganese (Mn) is a naturally occurring metal, with average concentrations of Mn in soil and rocks of 850 parts per million (ppm) and 650 ppm, respectively (Gilkes and McKenzie, 1988). The U.S. Environmental Protection Agency (USEPA) has established 50 parts per billion (ppb) as the secondary drinking water standard (SDS) for total Mn for aesthetic reasons (taste, color, clogging potential) (USEPA, 2004). The USEPA has also set 300 ppb for total Mn as the human health benchmark (HHB), as studies have linked chronic exposure of children to total Mn concentrations of .300 ppb in drinking water with learning impairments (Bouchard et al., 2007, 2011; Khan et al., 2012). In addition to geogenic sources, Mn can also be released into aquatic environments through human activities, including steel manufacturing, municipal wastewater discharge, landfills, and mineral processing from mining activities (Nadaska et al., 2010). Manganese, like iron (Fe), is soluble in its reduced state (Mn2+) and, as a result, can reach elevated concentrations under reducing conditions. Under oxidizing conditions, which prevail in rivers and well-mixed surface waters, Mn4+ is stable, which allows it to form insoluble oxyhydroxides (Hem, 1972). However, in groundwater, as a result of the often prevailing suboxic conditions of groundwater and the longer water-rock interaction, soluble Mn can be released to groundwater. A recent study in Scotland (Homoncik et al., 2010) found that Mn concentrations in 30 percent of groundwater samples (n 5 475) exceeded 50 ppb. In the United States, Mn concentrations in groundwater routinely exceed the SDS; a nationwide survey conducted by the U.S. Geological Survey (USGS) NAWQA program from 1992 to 2003 found that 31 percent of groundwater samples (n 5 4,976) exceeded the SDS and 12 percent exceeded the HHB (Ayotte et al., 2011). Because Mn is naturally occurring and is also used in industry, identifying the source of Mn can be challenging, as is the case with other naturally occurring elements, such as arsenic (As), chromium (Cr), nickel (Ni),

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cadmium (Cd), zinc (Zn), and uranium (U), among others. One approach to evaluating sources of naturally occurring elements to groundwater is to examine the geological, geochemical, and hydrologic relationships that may link the element to specific sources or mobilization processes. This approach has been beneficial for evaluating geologic sources and release mechanisms for As in groundwater (see Smedley and Kinniburgh, 2002 and references therein) but has not been specifically applied to Mn, likely as a result of its abundance in geologic materials. The main objectives of this study were to (1) examine relationships between Mn and relevant geochemical parameters in groundwater in the Roanoke River watershed of Virginia and (2) evaluate possible connections between Mn concentrations in groundwater and geologic units of the Eastern Blue Ridge cover sequence and Smith River Allochthon, including the James River–Roanoke River Manganese District (JRRRMD), a regional mineralized zone located within the watershed. Manganese is of interest in the Roanoke River watershed of Virginia, as Mn concentrations in both the Roanoke River and groundwater in the watershed routinely exceed the SDS (Smith, 1997; Nelms and Harlow, 2003; USEPA, 2004; and Chapman et al., 2013). Study Area Encompassing 14 counties in southern Virginia, the Roanoke River watershed (Figure 1) covers approximately 16 percent of Virginia. Filtered Mn concentrations in groundwater in the watershed (data from NURE and other USGS data sets (Smith, 1997; Nelms and Harlow, 2003; and Chapman et al., 2013) are shown in Figure 1. In the watershed, the median filtered Mn concentration in groundwater samples is approximately 27 ppb. Most filtered Mn concentrations in groundwater are below 50 ppb; however, ,30 percent of wells have elevated Mn concentrations (.50 ppb). Locations of historic mines of the JRRRMD, a belt of Mn mineralization with a NE-SW strike, are also shown in Figure 1. In addition to Mn, this district was also mined for Fe and barite from the mid–19th century until the end of the Second World War, with peak production occurring between 1911 and 1919 (Edmundson, 1938; Gooch, 1954, 1955). Site Geology The general geology of the Roanoke River watershed is displayed in Figure 2. Digital geologic data derived from the 1993 geologic map of Virginia were obtained from the USGS (Dicken et al., 2008); descriptions of rock units were obtained from the Virginia Division of Geology and Mineral Resources (Berquist, 2003).

Geologic formations were grouped into geologic regions by factors such as location, structural relationships, and geologic terranes described by the USGS (2003). The majority of the known Mn mine locations in the JRRRMD are found within the Proterozoic Alligator Back Formation of the Eastern Blue Ridge cover sequence. This formation consists primarily of interlayered metamorphosed sedimentary rocks (e.g., quartzite, pelitic schists, metagraywacke, and marble) and metamorphosed igneous rocks (e.g., amphibolite schist) (Henika, 1992; Berquist, 2003). Some Mn deposits are also located in the Cambrian Candler Formation, which is composed of phyllite, metasiltstone, and quartz mica schist with occurrences of marble (Berquist, 2003). Extensive folding and faulting of geologic formations in the study area have constrained the JRRRMD to a narrow (,10-km) belt bounded by two thrust faults: the Bowens Creek Fault (labeled BCF on Figure 2) on the west side and the Ridgeway Fault (labeled RF on Figure 2) on the east side. Between the two faults, the lithologic units of the Alligator Back Formation, and to a lesser extent the Candler Formation, are highly folded with several anticlines and synclines that form a corrugated assemblage of geologic units along the terrane boundary between the Eastern Blue Ridge and the Smith River Allochthon. Historically, the Mn oxide minerals of the JRRRMD have been separated into two main groups: “psilomelane type” and “wad,” in which psilomelane refers to Mn oxides with high specific gravity and wad refers to soft, powdery weathering products of Mn-rich geologic materials (Espenshade, 1954). Specific Mn oxide minerals identified in the JRRRMD by Espenshade (1954) are cryptomelane, psilomelane, and pyrolusite. Manganese oxides of the JRRRMD are predominantly concentrated in saprolite and fractured bedrock and occur as (1) massive replacement, in which original rock material has been replaced by Mn oxides; (2) void spaces and fractures filled in with Mn oxides; (3) cementing material (matrix) of vein quartz and quartzite breccias; (4) nodular masses of Mn oxides; and (5) wad. It is important to note that Fe oxides are also abundant throughout the JRRRMD, often occurring as mixed mineral assemblages with the Mn oxides. METHODS Rock Sample Collection To conduct chemical and mineralogical analyses of the JRRRMD ore deposits and other Mn-bearing rocks, we designed a sampling program using geologic maps and known localities of the ore deposits. Grab samples were collected in the Leesville Lake (LVL)

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Figure 1. (A) Map of the Roanoke River watershed (blue shaded area). The Roanoke River (blue line) lies in the center of the watershed. Circles represent well locations, and colors correspond to ﬁltered Mn concentrations (ppb). Red diamonds represent the locations of known historic Mn mines. Filtered Mn data were obtained from the NURE data set (USGS) and the USGS (Nelms and Harlow, 2003; Chapman et al., 2013). Historic Mn mine location information was retrieved from the MRV database (courtesy of the Virginia DMME). These ore deposits are referred to as the James River–Roanoke River Manganese District (JRRRMD). (B) Locations of ore and marble samples within the JRRRMD collected in the ﬁeld and from museum collections. Three ores of interest are shown (Piedmont Mine [RBG]; Leesville Lake [LVL]; Hutter Mine).

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Figure 2. Generalized geologic map of the Roanoke River watershed in Virginia and the locations of historic Mn mines of the JRRRMD. Circles represent well locations, and colors correspond to ﬁltered Mn concentrations (ppb). Geologic data obtained from the USGS (Dicken et al., 2008); descriptions of rock units obtained from the Virginia Division of Geology and Mineral Resources (Berquist, 2003). BCF 5 Bowen’s Creek Fault; RF 5 Ridgeway Fault.

and the Rustburg (RBG) quadrangles from surficial exposures of Mn ore deposits within the ore belt. Many former Mn mining sites exist within the JRRRMD, but physical terrain and uncertainties regarding property access limited the possibility of sampling locations. The two sites selected—LVL and RBG—were chosen because their locations made access to Mn ore deposits manageable. The locality for LVL samples was an unnamed mine located approximately 0.8 km northwest of the Myers Hematite Mine in northern Pittsylvania County; RBG samples were collected from the Piedmont Mine in northern Campbell County, located ,10 km southeast of Lynchburg, Virginia. Ore samples were targeted visually in the field based on the appearance of black-colored minerals in hand sample. Additional ore samples were obtained from museum collections of the Virginia Department of Mines, Minerals, and Energy (DMME) and the Virginia Museum of Natural History. DMME samples were collected throughout

the JRRRMD by W. Henika from surficial outcrops of Mn ores. Samples from the Virginia Museum of Natural History were originally collected in 1996 from a single dump of mine tailings from the Hutter Mine, a former Fe mine located ,2.5 km southwest from the LVL sampling location. The dump of tailings was an estimated 100 m2 in area, and the depth from which samples were extracted is not reported (Beard et al., 2002). In addition to ores, seven grab samples of marbles from the Alligator Back and Candler formations were collected. One marble sample was collected from the same site as RBG ore samples. The remaining six samples were collected by W. Henika throughout the JRRRMD. Locations of collected samples are shown in Figure 1. Powder X-Ray Diffraction Powder x-ray diffraction (XRD) was used to determine minerals in bulk samples. Samples were powdered

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Figure 3. Concentrations (mg/kg) of K versus Mn, Ca versus Mn, Ba versus Mn, and Co versus Mn from ore samples from RBG (n 5 3; ďŹ lled squares) and LVL (n 5 4; open circles). Note log scale.

using either a percussion mortar or a porcelain mortar and pestle. XRD analyses were conducted using a Rigaku MiniFlex II Desktop X-ray Diffractometer. Copper (Cu) Ka radiation was used with a fixed tube output voltage of 30 kV and fixed tube output current of 15 mA. Samples were analyzed with an x-ray scanning range of 5u to 85u 2h, a step size of 0.020u 2h, and a scanning speed of 1u 2h per minute. XRD patterns were analyzed using the PDXL software package to match the peaks from the diffraction patterns with known mineral species from the software database. Plausible mineral species were narrowed by including Mn as part of the mineral-matching criteria. Scanning Electron Microscopy Ore textures and qualitative elemental abundances were observed using scanning electron microscopy (SEM; Hitachi TM3000). Relative elemental abundances were determined through the use of energy

dispersive spectroscopy (EDS) in the form of elemental maps. Samples were analyzed using back-scattered electrons to collect Z-contrast images with an accelerating voltage of 15 kV. A working distance of 12 mm was used. Elemental maps were generated using EDS; data were collected for approximately 400 seconds per sample image. The lower limits of EDS detection for individual point analysis are approximately 0.1 weight percent for any given element. Bulk Chemistry Seven ore and seven marble samples collected from the study area were selected for bulk chemical analysis. Samples were powdered and sent to Activation Laboratories Ltd. (Ontario, Canada) for analysis. Ore samples were digested with hydrochloric, nitric, perchloric, and hydrofluoric acids and analyzed by inductively coupled plasma mass spectroscopy (ICP-MS). Marble samples were fused using lithium metaborate/tetraborate and analyzed by ICP-MS.

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Kiracofe, Henika, and Schreiber Table 1. Chemical analysis of ore samples by near-total digestion and ICP-MS analysis, conducted by Activation Labs, Inc. Sodium (Na), magnesium (Mg), Al, K, Ca, and Fe are measured in weight percent. All other elements are measured in ppm. Values preceded by “, ” were below detection limits. Values preceded by “.” were above detection limits.

B Li Na Mg Al K Ca Cd V Cr Mn Fe Hf Ni Er Be Ho Hg Ag Cs Co Eu Bi Se Zn Ga As Rb Y Zr Nb Mo In Sn Sb Te Ba La Ce Pr Nd Sm Gd Tb Dy Cu Ge Tm Yb Lu Ta Sr W Re Tl Pb Th U

RBG-111

RBG-110

,1 82.2 0.22 0.03 0.92 0.9 0.58 139 14 15.2 .10,000 0.85 ,0.1 471 13.8 3.6 5.2 230 3.09 1.08 .500 6.41 0.03 4.8 4,030 147 5.2 26.8 119 9 1.2 14.2 ,0.1 ,1 0.5 0.2 .5,000 39 172 16.4 79 25.8 30.2 4.7 27.5 839 0.3 2 9.5 1.2 ,0.1 694 0.2 0.024 2.07 128 1 0.9

,1 8.9 0.04 0.05 1.41 0.13 0.06 2.3 17 19.4 4,490 38 0.1 104 7.5 3.8 2.5 ,10 0.3 0.24 337 2.95 0.15 2.5 1,100 2 38.7 5.8 56.4 10 3.6 1.52 ,0.1 ,1 0.4 0.3 635 38.7 94.5 13.8 57.9 13.4 12.8 2 12.7 117 0.4 1.1 6.3 0.9 ,0.1 6 0.4 0.018 0.38 24.8 4.7 6.5

RBG01-BH ,1 4.7 0.04 0.18 0.88 0.06 0.05 19 24 18.1 1,780 43.2 0.2 599 7.2 9.6 2.4 ,10 0.36 0.21 297 2.15 0.07 4.6 .10,000 4 64.2 4.3 84.3 8 1.2 9.07 ,0.1 ,1 14.6 0.3 48 94.3 150 19.8 59.3 8.7 10.6 1.6 10.4 1,540 0.6 1 5.4 0.8 ,0.1 6.5 0.6 0.019 0.14 777 3.1 7.6

LVL-100

LVL-101

LVL-101A

LVL-102

,1 1.6 0.21 0.01 0.91 0.88 0.05 0.7 1 12.1 2,870 7.75 ,0.1 30.2 0.4 1.6 0.1 ,10 0.11 0.2 20.5 0.2 0.02 0.9 256 0.5 3 16.8 4.6 4 0.5 1.22 ,0.1 ,1 0.8 0.3 453 4.7 7 0.9 3.2 0.6 0.6 ,0.1 0.6 28.1 ,0.1 ,0.1 0.3 ,0.1 ,0.1 10 0.2 0.023 0.36 12.3 0.8 0.4

3 30 0.05 0.15 2.91 1 0.08 6.9 58 37 .10,000 21.4 ,0.1 285 7.8 5.2 3 80 0.79 4.96 447 4.55 0.11 2.7 784 ,0.1 9.2 230 56.6 19 7.2 11.8 ,0.1 1 2.3 0.4 2,640 51.9 124 19.9 84.9 19.3 18.8 2.9 16.7 382 0.4 1.1 6 0.9 0.3 104 1.1 0.021 3.02 10.9 5.2 3.3

,1 1 0.03 0.01 1.16 0.07 0.01 0.5 300 136 354 38.2 0.3 61.1 2.4 5.9 0.8 ,10 0.17 0.08 6.9 0.96 0.03 2.6 164 4.3 94.3 2.8 11 8 1.1 40 0.2 ,1 4 0.5 35 4.8 16.6 2.4 11.4 3.8 3.7 0.7 4.1 344 0.7 0.4 2.3 0.3 ,0.1 3.6 0.5 0.015 0.16 32.9 2.6 22.1

5 21.6 0.04 0.15 2.81 1.3 0.02 0.4 40 48.8 7,940 31.7 0.4 116 2 4.9 0.7 ,10 0.24 8.62 52.9 1.22 0.03 0.6 671 9.5 2.1 121 22 25 7.1 1.67 ,0.1 ,1 0.4 0.2 469 23.7 32.5 5.3 22 4.6 4.7 0.6 3.7 57.8 0.3 0.3 1.4 0.2 0.2 17.2 0.8 0.015 0.92 6.8 6.4 1.3

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Geologic Sources of Mn to Groundwater Table 2. Chemical analysis of marble samples by fusion with metaborate/tetraborate and ICP analysis, conducted by Activation Labs, Inc. More details on sample locations provided in Supplementary Information. Sample RBG-115 ALT-262 ALT-263 PTV-191 AL-3 KLY-822 LST-791

SiO2 (%)

Al2O3 (%)

Fe2O (%)

MnO (%)

MgO (%)

CaO (%)

Na2O (%)

K2O (%)

TiO2 (%)

P2O5 (%)

1.43 1.36 1.4 0.91 1.54 10.61 5.29

0.46 0.38 0.38 0.37 0.68 0.95 1.06

2.08 1.2 1.03 1.14 1.46 0.45 1.23

0.669 0.676 0.689 0.703 0.406 0.024 0.626

2.67 1.6 1.51 2.43 17.73 0.99 1.05

49.68 51.52 51.18 50.11 31.04 47.67 48.18

0.02 0.01 0.03 0.02 0.03 0.41 0.04

0.005 0.07 0.07 0.02 0.21 0.09 0.37

0.02 0.023 0.016 0.001 0.049 0.019 0.05

0.05 0.09 0.07 0.12 0.03 0.02 0.05

RBG 5 Piedmont Mine; ALT 5 Altavista site; PTV 5 Hutter Mine; AL 5 quarry in Appomattox County; KLY 5 near James River, northern Campbell County; LST 5 near Roanoke River, southern Campbell County.

Soil Mapping Soil types, specifically soils derived from alluvial processes, were examined to evaluate models of ore formation. First, alluvial soils in Campbell County were examined using descriptions from Bullard (1977). Alluvial soils were then identified in the county using the Gridded Soil Survey Geographic Database (SSURGO) (USDA, 2014) obtained online from the U.S. Department of Agriculture (USDA) Natural Resources Conservation Service (NRCS) Geospatial Data Gateway at https://gdg.sc.egov.usda.gov/. Alluvial soils were separated into two categories: Piedmont Upland terrace soils (greater elevations) and modern stream valley soils (lower elevations), which were overlayed with the known locations of Mn ore deposits using ArcGIS 10.2 (ESRI, 2013). Spatial relationships in the xy-plane between alluvial soils and Mn mine locations were examined visually. In the z-direction (vertical), spatial relationships were evaluated by extracting the elevations of Mn mining locations from digital elevation maps for Campbell County also obtained online from the NRCS Geospatial Data Gateway. Elevations of soil horizons and Mn mines were modeled using ArcGIS. To do this, the original shapefile (the polygon of the mapped soil) was converted into a raster file using a 30 m 6 30 m grid. The raster file was then converted to points, which were put in the center of each 30 m 6 30 m grid for extraction of elevation data. Finally, a digital elevation model was overlain using a 30 m 6 30 m grid, and the GIS tool Spatial Analyst was used to extract the elevation values from each grid.

Groundwater Data Compilation Groundwater geochemistry data in the Roanoke River watershed were gathered from three sources: (1) the National Uranium Resources Evaluation (NURE) data set (Smith, 1997) containing filtered samples (0.8-μm membrane); (2) the Storage and Retrieval

(STORET) data warehouse (USEPA, accessed 2014), containing unfiltered samples; and (3) USGS studies (Nelms and Harlow, 2003; Chapman et al., 2013), containing filtered samples (0.45-μm membrane). Groundwater Levels Hydraulic head values for wells in a portion of the study area near Smith Mountain Lake were calculated using static water level data from the U.S. Environmental Protection Agency (USEPA, accessed 2014). RESULTS XRD and SEM The LVL ore samples contain Fe and Mn minerals within different matrices (see Supplementary Information for examples), including quartzite with fractures filled by Mn and Fe oxides (LVL-100); Mn and Fe oxides that appear to have replaced the mica-schist country rock (LVL-101); and brecciated quartzite cemented by Mn-oxide (LVL-102). For the LVL samples, XRD analysis could not confirm specific Mn minerals, but SEM showed associations of Mn with aluminum (Al) and potassium (K), suggesting mixtures of lithiophorite and cryptomelane. Two of the RBG samples (RBG-111, 112A) contained brecciated quartz cemented by Mn oxides, similar to LVL-102; XRD confirmed cryptomelane in RBG-111 (see Supplementary Information). In RBG112A, although Mn oxides were not detected by XRD, the crystalline habit of the hand sample was similar to that of manganite. The Mn oxide replacement of weathered schist observed in RBG-114 was similar to that of LVL-101; however, in this sample, XRD identified lithiophorite. RBG-110 was a massive Fe/Mn ore deposit, with country rock barely distinguishable. As a result of the abundance of Fe-oxides in this sample, Mn minerals were not identified by

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Figure 4. Representative soil maps in southwest Campbell County, Virginia, showing alluvial soils (terrace deposits) of the Piedmont Uplands and modern stream valleys as well as the locations of the historic Mn mines. Soil data from SSURGO database (USDA, 2014).

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Geologic Sources of Mn to Groundwater Table 3. Percentage of known mines (n 5 60) in and near Piedmont Uplands and modern stream valley soils. Search Distance (m) Piedmont Uplands 0 50 75 100 Modern stream valleys 0 50 75 100

Number of Mines

% of Mines

13 37 48 60

21.7 61.7 80.0 100.0

0 2 6 6

0.0 3.3 10.0 10.0

percents ranging from 0.406 to 0.703 (3,144–5,444 ppm Mn).

Soil Mapping

XRD, but the association of Mn with Al in the SEM/ EDS analysis suggests the presence of lithiophorite. Bulk Chemical Analysis Figure 3 displays plots of Mn versus selected major and trace elements using the bulk chemical data for the seven ore samples (data set presented in Table 1). Because two of the ore samples contained Mn concentrations greater than the detection limit, we did not conduct a statistical evaluation of relationships but instead looked at patterns. Manganese has a positive relationship with K, calcium (Ca), barium (Ba) and cobalt (Co), as shown in Figure 3. Results of marble sample analyses are shown in Table 2. All of the marble samples analyzed contained detectable Mn and, with exception of one sample, had MnO weight

Soil maps were compiled to examine the spatial relationships between the known locations of Mn mines in Campbell County (ore deposits) and alluvial soils, including the Piedmont Upland terrace deposits, found at high elevations, and modern stream valley soils, occurring in or near modern streams (Figure 4). Three types of analyses were conducted: visual inspection, calculation of the percent of known Mn mines within these soils units, and comparison of elevations of the known Mn mines and the soil units. Although not a perfect association, visual inspection of locations of known Mn mining localities suggest that they cluster either within or near alluvial (both ancient and modern) soil units (Figure 4). There are two important things to note when examining Figure 4 and evaluating the spatial association of Mn mines and alluvial soils. First, the SSURGO soil data are based on mapping of only complete soil profiles. Previous mapping in this region of the Piedmont documents that terraces are highly eroded, often leaving isolated deposits (straths) that contain rounded cobbles and gravels (Pazzaglia and Carter, 2015). Thus, the presence of terrace deposits is often unidentified in the field and is thus underreported in the database. Second, only the Mn mines archived in the Mineral Resources of Virginia (MRV) database are shown in Figure 4;

Figure 5. Boxplots of the elevations (modeled using ArcGIS) of modern stream valley deposits, Piedmont Upland terrace deposits, and known historic Mn mines in Campbell County, Virginia. Circles reﬂect outliers of the elevation data; asterisks represent extreme outliers.

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Kiracofe, Henika, and Schreiber Table 4. Summary statistics for Mn and other relevant parameter concentrations used in Figure 7. Data from STORET (USEPA, 2014) and from Chapman et al. (2013; using Virginia data only).

STORET (USEPA, 2014) Mn (ppb, unfiltered) Fe (ppb, unfiltered) Ca (ppm, unfiltered) Chapman et al. (2013) Mn (ppb, filtered) Fe (ppb, filtered) Dissolved oxygen (mg/L)

Mean

Minimum

* * *

343 340 342

72.93 493.62 29.17

168.41 1,130.28 34.14

10 10 0.4

0.2 or 1 3 to 10 —

87 87 85

162 948 4.3

334 2,045 3

0.17 1.5 0.05

Maximum 2,000 7,000 280 2,013 9,965 9.6

DL 5 detection limit. *The detection limit is not explicitly stated in the STORET database; we assume a DL of 10 ppb for Fe and Mn based on the presence of many

values of 10, and no values under 10, in the database. The DL for Ca is also not stated.

there are many other localities in which Mn deposits have been located that are not included here. A second way to evaluate relationships between the known Mn mine locations and the alluvial soil units is to calculate the percentage of Mn mines within the Piedmont Upland and modern stream valley soils. This calculation was conducted in GIS for 60 known Mn mines in the study area using several search distances (0, 50, 75, and 100 m) to account for likely uncertainties in locations for both the known Mn mine locations and in the soil units. Results (Table 3) show that 13 out of 60 (21.7 percent) of the mines are located within (0 m) Piedmont Upland soils; 100 percent of the mines are located with 100 m of this soil group. None of the known mines are located within mapped modern stream valley soils, but 10 percent of the known mines are located within 100 m of this soil group. Last, we compared the elevation data for the alluvial soil deposits and Mn mining locations modeled using ArcGIS (v. 10.2; ESRI, 2013) (Figure 5). Results show that the distribution of elevations is similar between Piedmont Upland terrace deposits (mean, 217 m; standard deviation [SD], 40 m) and Mn mining locations (mean, 225 m; SD, 33 m), suggesting that the ore deposits were at the same relative locations of discharge zones (streams) during their time of formation. Modern alluvial soils have a lower mean elevation (176 m; SD, 34 m) than the terrace and ore deposits but are still within the standard deviation of the terrace and ore deposit mean.

Groundwater Mn, Fe, Dissolved Oxygen, and pH Summary statistics of Mn and other relevant parameter concentrations in groundwater from STORET (USEPA, accessed 2014) and the Virginia portion of the Chapman et al. (2013) data set are shown in Table 4; data are shown in boxplots in Figure 6. Results show

that the mean Mn concentrations in groundwater samples for both data sets are higher than the SDS of 50 ppb and that maximum Mn concentrations reach 2,000 ppb. Statistical evaluation of the relationship of Mn with dissolved oxygen (DO), Fe, and Ca in groundwater was conducted to examine possible geochemical processes influencing Mn. Chapman et al. (2013) had previously documented a negative correlation of Mn and Fe with DO in groundwater in the regional Piedmont data set. To further examine these relationships, we compared the distribution and means of concentrations of DO, Fe, and Ca between three groups of Mn concentrations (group 1: Mn , 50 ppb; group 2: 50 , Mn , 300 ppb; group 3: Mn . 300 ppb). Because DO data are sparse in the study area, we used the Chapman et al. (2013) data set, which includes wells located within the Piedmont and Blue Ridge Crystalline aquifers in Virginia. The boxplots show the distribution of DO concentrations for the three groups (Figure 6). Results of a Student’s t-test (alpha 5 0.05) show that the mean for group 1 is significantly different from those of groups 2 and 3 (p , 0.0001), although it should be noted that group 3 has lower n values than do the two other groups. We used the same approach to assess relationships between Fe and Ca concentrations and the three Mn groups in groundwater but using the larger STORET data set for wells within the Roanoke River watershed (Figure 6). Both Fe and Ca concentrations were logtransformed prior to plotting as inspection of the quantile plots revealed log-normal distribution. However, for the statistical comparison, the orig‐ inal (untransformed) data set was analyzed using non-parametric methods. For the Fe data, using the Kruskal-Wallis test yields a p-value of ,0.0001, indicating a difference in the median values between at least two groups. The Wilcox signed-rank test shows that Fe concentrations for all groups are statistically different from one another (all p-values ,0.05).

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Geologic Sources of Mn to Groundwater

however, there is no statistically significant difference in the Ca means between groups 2 and 3. Similar to the analysis above, it should be noted that group 3 has lower n-values than the other two groups. Combined, these results show that increases in Mn concentrations in groundwater are associated with increases in both Fe and Ca concentrations, as well as decreases in DO. The distribution of total Mn concentrations (from the STORET data set) in geologic formations is shown in Figure 7, with summary statistics shown in Table 5. Results show that the Candler, Bassett, and Alligator Back formations are the three formations that have the greatest 50th percentile for total Mn concentrations in groundwater; wells in these formations also have mean Mn concentrations greater than the SDS. Groundwater Levels and Mn Concentrations Figure 8 shows the relationships between Mn concentrations and groundwater levels (hydraulic head) in 11 wells located within 2 km of Smith Mountain Lake, which is assumed to be a regional discharge area for groundwater. The concentrations of Mn in groundwater are lower in wells with higher hydraulic head and increase as hydraulic head decreases closer to Smith Mountain Lake, suggesting that Mn concentrations may increase with distance along a generalized groundwater flowpath. However, it is important to note that groundwater flowpaths are difficult to delineate in wells installed in fractured crystalline rocks because the wells are likely open to multiple fracture zones and thus reflect mixed conditions of hydraulic head and rock-water interactions. It is also the case that the data set is limited to only 11 wells. Further data collection would be needed to evaluate these patterns. DISCUSSION Origin of the JRRRMD Ore Deposits Figure 6. Boxplots of dissolved oxygen (DO), log Fe, and Ca for grouped Mn (unﬁltered) concentrations in the study area. Group 1: Mn , 50 ppb. Group 2: 50 , Mn , 300 ppb. Group 3: Mn . 300 ppb. Sample numbers (n) shown in boxplot. (Top) DO and Mn data from Chapman et al. (2013). (Middle) Fe and Mn data from STORET (accessed 2014). (Bottom) Ca and Mn data from STORET (USEPA, 2014).

For the Ca dataset, the Kruskal-Wallis test yields a p-value of ,0.0001, indicating a difference in the median values between at least two groups. The Wilcoxon signed rank test shows that the mean of Ca concentrations for group 1 is different than the means for group 2 (p , 0.0001) and group 3 (p 5 0.0004);

Results from XRD, SEM, and ore chemical analysis suggest that the ore deposits of the JRRRMD are supergene (formed at or near the Earth’s surface at low temperatures) in origin. XRD results show secondary Mn and Fe oxide minerals (e.g., cryptomelane, lithiophorite, goethite) that are commonly found in supergene ore deposits (Nicholson, 1992). Goethite (FeOOH) is not thermodynamically stable at temperatures that exceed 130uC (Schmaltz, 1959), implying the fluids that formed the ore deposits of the JRRRMD were lower than 130uC. Further evidence of a supergene origin is provided by ore morphologies, as observed using SEM. These morphologies, including fracture fillings by Mn and Fe oxides, botryoidal

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Figure 7. Boxplots of unfiltered Mn concentrations (ppb) in groundwater using data from STORET (USEPA, 2014) in wells open to different lithologies within the Roanoke River watershed. Note the log scale. The top line reflects the human health benchmark (300 ppb) and the bottom line reflects the secondary drinking water standard (50 ppb). Number of samples shown in the figure. ABF 5 Alligator Back Formation.

Fe oxides, botryoidal Mn oxides, and massive replacement of pre-existing material by Mn oxides, are typical of supergene ore deposits (Kim, 1984). Lastly, bulk chemical analysis showed chemical associations between Mn and trace metals (e.g., Ba and Co) that are also consistent with supergene ore deposits (Nicholson, 1992). By comparison, the Hutter Mine (see Figure 1) is a known hydrothermal deposit and contains an abundant array of Mn-bearing minerals formed at high temperatures. A few examples of these include manganosite (MnO), galaxite (MnAl2O4), and jacobsite (MnFe2O4) (Beard et al., 2002). Espenshade (1954) previously proposed a supergene origin of the Mn oxides of the JRRRMD. However, Espenshade’s model of ore formation invokes dissolution of Mn from soluble mineral phases in the country rock by downward-circulating groundwater, implying that the Mn oxides were precipitated at depth as Table 5. Summary statistics for Mn (unfiltered, ppb) concentrations in wells located in different geologic formations. Data from STORET (USEPA, 2014). Formation ABF AF BF CF FMF

Mean

Minimum

Maximum

142 24 94 16 66

105.4 24.3 62.4 224.5 199

268.3 31 101 193.9 42

5.3 0.3 0.05 18 3.2

1,406 136 690 500 199

ABF 5 Alligator Back Formation; AF 5 Ashe Formation; BF 5 Basset Formation; CF 5 Candler Formation; FMF 5 Fork Mountain Formation.

groundwater flowed downward through fractures and along planes of weakness (e.g., relict bedding planes) in weathered rock formations. Espenshade’s model of downward circulation and Mn ore precipitation at depth is problematic because it does not take into account how Mn oxides form from a redox perspective. At shallow aquifer depths, DO in groundwater is replenished by recharge, resulting in oxidizing conditions under which Mn4+ is stable as insoluble Mn ox‐ ides. With increasing depth in aquifers, DO becomes more depleted as it is consumed by microbes and is not replenished by recharge (Chapelle, 2001). As a result, aquifers generally become more reducing with depth, favoring the soluble Mn2+ as the stable Mn species. Thus, in typical aquifers, downward groundwater flow would not promote oxidation of Mn2+ to Mn4+ and subsequent precipitation of Mn oxides, as Espenshade’s model assumes. We propose an alternate “upwelling” model (Figure 9) that, similar to the Espenshade model, first invokes downward groundwater circulation in recharge areas. As groundwater flows downward through pore spaces and fractures, it can dissolve Mn from country rock through chemical weathering. During the diagenesis and subsequent metamorphism of sedimentary and igneous rocks, Mn2+ is the dominant oxidation state of Mn incorporated into mineral structures (Muller, 1971; McKenzie, 1980; and Dromgoole and Walter, 1990) and is the most probable source of Mn released into solution by weathering reactions. At depth, as the aquifer becomes increasingly reducing, Mn2+ is able to

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Figure 8. Manganese (unﬁltered, ppb) concentrations versus hydraulic head (m AMSL) for wells within 2 km of Smith Mountain Lake, Virginia. Data from STORET (USEPA, 2014).

persist in solution. This reduced, Mn-bearing groundwater then upwells toward regional discharge zones, such as Smith Mountain Lake, where reduced groundwater containing Mn2+ can mix with shallow, oxic groundwater and promote the precipitation of Mn oxides. Examples of other regions where upwelling models of ore formation have been invoked include the nearby Shenandoah Valley of Virginia, where Hack (1965) associated the occurrence of secondary Mn-oxides derived from a dolomitic protore with overlying alluvial soil deposits. Chan et al. (2001) also describe secondary Fe and Mn oxides precipitating from upwelling reduced groundwater encountering oxic groundwater in a Jurassic sandstone. In both Australia and Africa, upwelling models have been proposed for the deposition of other redox-sensitive metal ore deposits, such as uranium and vanadium, precipitating within hardened Ca-rich layers in soil profiles (Carlisle, 1983). Analysis of soil data in the study area suggests spatial associations between the location of known Mn mining localities, representing outcrop locations of Mn ore deposits, and ancient alluvial soils, which serve as relicts of previous discharge zones. Comparison of the elevations of alluvial soil deposits in the Piedmont Upland terrace deposits and locations of Mn mines show that the majority of mines occur within the elevation ranges of the Piedmont Upland terrace deposits. Field observations show that most mining locations are located on hilltops well above the elevations of nearby streams. Why are the Piedmont Upland terrace deposits currently found at high elevations on ridges? The results gathered in this study, combined with previous work, suggest that the topography of the study area, and of the Piedmont in general, has been inverted

throughout geologic time. Throughout the Piedmont, alluvium is found at greater elevations than—and often considerable distances from—active streams (Henika, 1971, 1977, 1992, 1997; Conley and Henika, 1973; Weems et al., 2009; and Lang et al., 2010). Over time, terrace dissection and stream migration have resulted in a lowering of topography relative to relict alluvial deposits, resulting in stream valleys that later become ridges. The explanation for why alluvial sediments remain in place while other sediments are eroded is related to mineralogy. The alluvial deposits are characterized by quartz gravel and cobble beds that are resistant to erosion; however, feldspar and lithic clasts in these deposits have weathered to clay, which, when exposed to heat, forms a low-permeability hard layer perched on the alluvial straths. In contrast, the crystalline and marble units of the Piedmont underlying the alluvial deposits are more easily erodible, allowing for recent drainage channels to cut down into these mica gneiss, marble, amphibolite, and schist units while the alluvial layer remains. Others have suggested that topographic inversion occurs in other regions of the Piedmont (Markewich et al., 1990), where differential weathering rates of geologic materials in this erosional terrain can promote such landscape evolution. The East Coast of the United States is, furthermore, considered a dynamic uplift zone, which could promote such weathering processes (Rowley et al., 2013). Work by Prince et al. (2010) suggests that stream migration and capture also play a vital role in isolating relict terrace deposits at relatively higher elevations in both the Blue Ridge and Piedmont physiographic provinces. Although the majority of Mn mines are located within or near alluvium of the

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Figure 9. (A) Conceptual upwelling model for the formation of the ore deposits of the JRRRMD. The dashed line represents the redox boundary between alluvial sediments, where oxic groundwater dominates, and fractured saprolite and bedrock, where reducing groundwater occurs. This redox interface is where shallow, oxic groundwater and deep, reducing groundwater can mix together, altering the redox chemistry to favor the precipitation of Mn oxides in close proximity to streams. (B and C) Conceptual diagram of topographic inversion in relation to the formation and occurrences of Mn oxides in the Piedmont Uplands within the JRRRMD. (B) Original formation model of ore deposits in stream valleys (discharge zones) as proposed in this study; possibly how ores are being deposited in the present day. (C) Inverted topography caused by extensive weathering, causing ancient stream valleys to become ridges.

Piedmont Uplands, several mines do occur at lower elevations either within or proximally close to alluvium in modern stream valleys. The occurrence of Mn oxides in modern alluvium and field observations of Mn-coated sediment in modern streams near historic Mn mining locations suggest that Mn ores are still forming in the present day.

Figure 9 shows our conceptual model of the geomorphological evolution involved with topographic inversion as it relates to the formation of the JRRRMD through upwelling. Mn oxides were originally deposited by upward-circulating groundwater that carried soluble Mn2+ from a reducing groundwater zone into an oxic groundwater zone, which

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promoted the oxidation and subsequent precipitation of Mn oxides (panel A). At some point after deposition, extensive weathering processes (e.g., stream migration, terrace dissection, preferential weathering) have caused the topography around the ore deposits in the Piedmont Uplands to become lowered (panel B). These weathering processes effectively cause topographic inversion, isolating the Mn oxides of the JRRRMD in the Piedmont Uplands on hilltops and ridges, where stream valleys were once present. It is important to note that development of this conceptual model is based on our limited data set; additional data need to be collected to allow for further testing. Geologic Sources and Controls on Mn Concentrations in Groundwater Analysis of the available chemical data suggests that there are likely several geologic sources of elevated Mn concentrations in groundwater, including Mn-rich marbles in the metamorphic rocks of the Eastern Blue Ridge cover sequence and the Mn ores of the JRRRMD. There are several pieces of evidence to support diffuse chemical weathering of Mn-bearing metamorphic rocks as a primary source of Mn to groundwater. First, although wells open to the Alligator Back Formation, which is the dominant formation containing Mn ores, contain elevated Mn concentrations, wells in other formations (Candler, Bassett, Ashe, Fork Mountain) also contain elevated Mn. Second, groundwater samples with elevated Mn also contain higher Ca concentrations, suggesting that carbonate-rich lithologies (e.g., marble, calcareous quartzite, calcareous schist) are also sources of Mn. Analysis of marble samples collected in the study area shows that Mn concentrations reach up to 5,500 ppm, thereby documenting that Mn is present in rocks. It is important to note that high concentrations of Mn in rocks are not necessary to cause elevated Mn concentrations in groundwater, as only a few parts per million of Mn, if fully released to groundwater, are needed to result in an exceedance of the SDS. In addition to carbonate rocks, Ca may also be introduced into solution from minerals within metavolcanic rocks in the study area (e.g., amphibole schists and amphibolites) as well as other Ca-bearing minerals. For example, the sediments and volcanic materials in the study area—the Alligator Back Formation in particular—were deposited in a Neoproterozoic rift basin. These tectonic settings are associated with both Mn and Fe formations resulting from metal-rich fluids being exhaled from oceanic crustal venting (Bühn and Stanistreet, 1997), which could become entrained within sediments during the diagenesis of the protoliths for the metasedimentary rocks.

The inverse relationship between Mn and DO in groundwater of the regional Piedmont and Blue Ridge crystalline aquifers in Virginia (Chapman et al., 2013) documents that higher Mn concentrations occur under reducing conditions. The increase of Mn concentrations along a generalized flowpath toward Smith Mountain Lake suggests that once soluble Mn is released to groundwater it can be transported along the flowpath. Because DO becomes depleted in aquifers as a result of microbial consumption and lack of replenishment from recharge (Chapelle, 2001), as groundwater flows for extended periods of time (years, decades, centuries), it is reasonable to assume that reducing conditions will develop with depth, distance, and time in an aquifer, promoting elevated Mn concentrations in groundwater. Deciphering if the Mn ores of the JRRRMD are sources of Mn to modern groundwater is difficult to interpret with the current data set. As described above, the majority of the JRRRMD ore deposits are located at high elevations in the Piedmont Uplands, which are likely local recharge areas for groundwater. As recharge waters are air-saturated, they are unlikely to promote dissolution of Mn oxides. However, if the upland soils contain sufficient organic matter, and if DO is depleted in recharge waters, it is possible that the Mn ores could be reductively dissolved, releasing Mn to groundwater. Further work would be needed to address this question. CONCLUSIONS In this study, we examine connections between Mn in modern groundwater, bedrock geology, and Mn ores in the Roanoke River watershed of the Piedmont Province of Virginia. Analysis of groundwater chemistry data shows patterns that reflect weathering of Mnbearing rocks, release of Mn into groundwater, and subsequent transport under reducing conditions toward the regional discharge area. Analysis of ore mineralogy from the JRRRMD, which occurs in the watershed, supports an earlier model (Espenshade, 1954) of supergene formation. However, in contrast with this earlier model, our results suggests that the ore deposits were formed from upwelling of reduced groundwater, carrying soluble Mn2+ dissolved from the country rock and mixing with oxic groundwater in discharge zones. Our interpretation is supported by the spatial associations between the locations of known Mn ore deposits and both relict and modern alluvial soil deposits. The presence of relict alluvial deposits on hilltops and ridges suggests that the topography in the JRRRMD has been inverted since the deposition of Mn ores in the Piedmont Upland

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terraces. This weathering process may explain why the majority of the ore deposits in the JRRRMD are located at elevations above modern streams and rivers. Combined, our results suggest that the mechanisms that formed the JRRRMD ores in the geologic past are still active in the modern day. Field observations of Mn staining on rocks and alluvial sediments near and within the Roanoke River suggest that Mn oxides are currently precipitating as groundwater discharges to the river; these oxides may in the future become concentrated enough to be considered ores. Overall, this study highlights the usefulness of integrating geochemical data with other geologic information to better delineate and evaluate geogenic element cycling in both modern and ancient environments. ACKNOWLEDGMENTS The authors gratefully acknowledge funding from Dominion Power and the Geological Society of America. We also thank Jim Beard (Virginia Museum of Natural History) for lending us samples from the Hutter Mine deposit and for mineralogy advice and Don Rimstidt (Virginia Tech) for helpful discussions on geochemistry. Thanks also goes to Joshua Rubinstein and William Lassetter (Virginia Division of Mines, Minerals and Energy), Bradley White (Virginia Department of Environmental Quality) and Neil Johnson, Luca Fedele, Andrew Muscente, and Lowell Moore (Virginia Tech). We thank three anonymous reviewers and the editor for their insightful comments that greatly improved the manuscript. REFERENCES AYOTTE, J. D.; GRONBERG, J. M.; AND APODACA, L. E., 2011, Trace Elements and Radon in Groundwater across the United States, 1992–2003: U.S. Geological Survey Scientific Investigations Report 2011–5059, 115 p. BEARD, J. S.; TRACY, R. J.; AND HENIKA, W. S., 2002, Minerals of the Hutter Mine: A new manganese-barium mineral locality in northern Pittsylvania County, Virginia: Rocks Minerals, Vol. 77, No. 5, pp. 320–325. BERQUIST, C. R. J., 2003, Digital Representation of the 1993 Geologic Map of Virginia—Expanded Explanation: Virginia Department of Mines, Minerals, and Energy (DMME) Publication 174. BOUCHARD, M.; LAFOREST, F.; VANDELAC, L.; BELLINGER, D.; AND MERGLER, D., 2007, Hair manganese and hyperactive behaviors: Pilot study of school-age children exposed through tap water: Environmental Health Perspectives, Vol. 115, No. 1, pp. 122–127. BOUCHARD, M. F.; SAUVÉ, S.; BARBEAU, B.; LEGRAND, M.; BRODEUR, M.-È.; BOUFFARD, T.; LIMOGES, E.; BELLINGER, D. C.; AND MERGLER, D., 2011, Intellectual impairment in schoolage children exposed to manganese from drinking water: Environmental Health Perspectives, Vol. 119, No. 1, pp. 138–143. BÜHN, B. AND STANISTREET, I. G., 1997, Insight into the enigma of Neoproterozoic manganese and iron formations from the

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APPENDIX 1

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