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EDITORIAL BOARD

Chair John A. Dal Pino, S.E. Claremont Engineers Inc., Oakland, CA chair@STRUCTUREmag.org

Jeremy L. Achter, S.E., LEED AP ARW Engineers, Ogden, UT

Erin Conaway, P.E. AISC, Littleton, CO

Linda M. Kaplan, P.E. Pennoni, Pittsburgh, PA

Charles “Chuck” F. King, P.E. Urban Engineers of New York, New York, NY

Nicholas Lang, P.E. Vice President Engineering & Advocacy, Masonry Concrete Masonry and Hardscapes Association (CMHA)

Jessica Mandrick, P.E., S.E., LEED AP Gilsanz Murray Steficek, LLP, New York, NY

Jason McCool, P.E. Robbins Engineering Consultants, Little Rock, AR

Brian W. Miller Cast Connex Corporation, Davis, CA

Evans Mountzouris, P.E. Retired, Milford, CT

Kenneth Ogorzalek, P.E., S.E. KPFF Consulting Engineers, San Francisco, CA (WI)

John “Buddy” Showalter, P.E. International Code Council, Washington, DC

Eytan Solomon, P.E., LEED AP Silman, New York, NY

Jeannette M. Torrents, P.E., S.E., LEED AP JVA, Inc., Boulder, CO

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STRUCTURE ® magazine (ISSN 1536 4283) is published monthly by The National Council of Structural Engineers Associations (a nonprofit Association), 20 N. Wacker Drive, Suite 750, Chicago, IL 60606 312.649.4600. Periodical postage paid at Chicago, Il, and at additional mailing offices. STRUCTURE magazine, Volume 30, Number 6, © 2023 by The National Council of Structural Engineers Associations, all rights reserved. Subscription services, back issues and subscription information tel: 312-649-4600, or write to STRUCTURE magazine Circulation, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606.The publication is distributed to members of The National Council of Structural Engineers Associations through a resolution to its bylaws, and to members of CASE and SEI paid by each organization as nominal price subscription for its members as a benefit of their membership. Yearly Subscription in USA $75; $40 For Students; Canada $90; $60 for Canadian Students; Foreign $135, $90 for foreign students. Editorial Office: Send editorial mail to: STRUCTURE magazine, Attn: Editorial, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606. POSTMASTER: Send Address changes to STRUCTURE magazine, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606.

STRUCTURE is a registered trademark of the National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

Features

30 LISTENING TO THE STRUCTURE

By David J. Odeh, Carl Giegold, and Alan Joslin

The bucolic farmlands of eastern Massachusetts play host to a new music education and performance facility known as Groton Hill Music Center.

Cover Feature

12 EMORY WINSHIP HOSPITAL PEDESTRIAN BRIDGE

By Georgi I. Petrov P.E., A.I.A., Raymond Sweeney P.E., Darin Appleton P.E., and Charles Besjak P.E., S.E., F.A.I.A

Pedestrian bridges are a vital part of modern hospital complexes, providing weather-protected connections for patient transfer and staff connectivity.

Columns and Departments

46 OLD BONES - NEW PURPOSE

By Michael J. Richard, Ph.D., P.E., Paul E. Kassabian, P.E., P. Eng., CEng

Located in Boston’s Historic Newspaper Row District and at the location of Benjamin Franklin’s birth, One Milk Street is composed of three connected buildings: the Boston Transcript Building (circa 1873), the Boston Post Building (circa 1874), and a reinforced concrete connector building from the 1930s.

Performance with Elastomeric Adhesives

By Bilal Alhawamdeh, Ph.D., A.M.ASCE, and Xiaoyun Shao, Ph.D., P.E., M.ASCE

53 Structural Systems

Modular Composite Core Wall System

By Ahmad Rahimian, Ph.D., P.E., S.E., F. ASCE, Konstantin Udilovich, P.E., Ilya Shleykov, Ph.D., P.E., LEED AP BD+C, and Jeffrey Smilow, P.E., F. ASCE

64 Structural Design

The Devil in the Details

By John P. Phelan, P.E. and Michael F. Hughes, P.E., S.E.

In Every Issue

5 Advertiser Index

27 Wind/Seismic Guide

58 NCSEA News

60 SEI Update

62 CASE in Point

Lessons Along the Path to Licensure

By Kenneth O’Dell, P.E., S.E.

Job interviews are stressful. The interviewer and interviewee alike often play an unconscious game of chicken, each trying to second guess what the other really wants to hear. Engineering graduates and early career professionals often share their goal of attaining P.E. or S.E. licensure. Yet this is often just a rote answer because it sounds like what an employer wants to hear. Clearly, an engineering license is a critical step in a career; it leads to promotion, status, and increased responsibility. Yet, many new graduates and engineers-in-training (EITs) do not fully grasp what it will take or what it means. This is not surprising considering most firms eagerly await the chance to announce the new P.E. or S.E. within their ranks and often push passage before preparation. In lieu of pushing staff to pass an exam, the profession of structural engineering will be better served when firm leaders and mentors prepare staff members for success.

Success and confidence go hand in hand, yet hiring models do little to instill confidence. Employers seek qualified people to fill roles and firms seek individuals with experience for entry level positions. The perfect candidate is someone with a good internship or preferably several good internships. Firms want someone who can “run with it.” Unfortunately, this very notion contradicts the learning model of most university programs. For many, the transition from classroom to office is disconcerting because there are gaps between what was taught in class and what is needed in practice. Therefore, from day one, they feel behind the curve and their confidence is undermined. However, if expectations are aligned with the known gaps newly graduated engineers are understood to have, leaders and managers can empower success. In this alignment, confidence is built toward success when working together to bridge the gaps.

Strong bridges are built purposefully, not by random chance. General engineering curricula provides foundational experience, it does not provide everything a new engineering candidate needs to know. When this is ignored

during hiring, tension is created before the process even begins. Firms and supporting organizations such as NCSEA should recognize this and plan training programs that acknowledge the gaps and work to fill them. This needs to be purposeful, the gaps need to be defined and understood, and thoughtful (not random) steps taken to prepare training programs. Importantly, firms must invest time and energy to create internal training, or they need to invest in programs offered by professional organizations and empower their staff to participate. Equally important, new engineers need to understand that their learning is not done. Just as they invested in their college learning experience, they will

and project partners. Without a grand view of training opportunities and their outcomes, the distraction mentality takes hold and just like setting a coin aside allows it to lose its luster, untrained engineers will lose momentum and question their next steps.

need to invest in their career and professional learning. A university education is not a singular transaction but an investment in future learning. The greatest gap filled by training is the confidence built on new knowledge. When a coin is rubbed between fingers both sides are polished equally. Training is like a coin; a well exercised training program refines both sides--the trainee and the trainer. In developing internal training programs, firms are well served in understanding that engineers are not educators by nature. However, effective training benefits the entire firm by lifting each tier of the team. To be effective, wholistic programs focus on interpersonal skills as well as technical skills. Interpersonal skills should be prioritized, not seen as a distraction from billable work. Every step of training is an investment in the future as the entire team gains skills. Trainees learn new technical information and trainers learn better ways to interact with colleagues while enhancing their skills for working with clients

Licensure is an outcome of good experience; it should not be a step toward experience. Just as good training builds confidence and capabilities, these in return build opportunities for increased responsibility and better project experience. When on the job training and direct project experience do not relate to core knowledge necessary for passing an exam, an opportunity toward licensure may be provided but without the tools necessary for success. Expecting an exam should be taken before well-rounded project experience has been gained highlights that our industry requires new engineers to do the work on their own. This expectation is fraught with failure and has little opportunity to help build an engineer’s confidence. Unconsciously telling ourselves “We can do this faster than training someone else to do it” empowers no one. In today’s fast paced market, hanging on and doing it ourselves increases the stress level of experienced engineers and does little to help junior engineers gain critical skills. Instead, it reinforces the feeling that new engineers don’t “know anything.” We must break this unconscious cycle by training our engineers to share experience and knowledge regardless of the number of years we have in the profession. In doing so licensure becomes the natural outcome of experience gained rather than the artificial stressor.

The path to licensure should be walked together. Leaders must empower every team member to succeed by providing the tools behind the opportunity. With effective training and on-the-job knowledge sharing prospective examinees will gain what they need from colleagues working together.■

Kenneth O’Dell, P.E., S.E., is a principal with MHP Structural Engineers in Long Beach, CA, and a member of the NCSEA Board of Directors.

Licensure is an outcome of good experience;it should not be a step toward experience.

structural INFLUENCERS

Dr. S.K. Ghosh

Dr. S. K. Ghosh is known internationally for his work in earthquake engineering. He specializes in analyzing and designing reinforced and prestressed concrete structures, including wind- and earthquake-resistant designs. He has influenced seismic design provisions in the United States for many years by serving on or chairing numerous committees and advisory panels. Dr. Ghosh is active on many national technical committees; he is a Fellow of the American Society of Civil Engineers (ASCE), ASCE’s Structural Engineering Institute (SEI), the Precast/Prestressed Concrete Institute (PCI), and an Honorary Member of the American Concrete Institute (ACI). He is a member of ACI Committee 318 and an Emeritus member of the ASCE 7 Standard Committee.

Dr. Ghosh heads the consulting practice, S.K. Ghosh Associates LLC (SKGA), in Palatine, IL, and provides code-related, seismic, and specialized structural consulting services in the United States and abroad. SKGA is now a subsidiary of the International Code Council (ICC).

Why did you decide to enter the field of structural engineering?

I grew up in India, where a particular profession is favored during a particular period. In my father’s time, it was the law; in my time, it was engineering; shortly after that, it became medicine; now it is IT. My father was very influential in the decision process; he strongly suggested civil engineering. However, when it came time to determine which area to focus on, I decided on structural engineering.

What do you consider your most significant contribution to the engineering profession?

The aspect of my career that gives me the most satisfaction is the opportunity I’ve had to teach so many other engineers, especially young engineers. A primary focus of SKGA is education. I’ve had the opportunity to teach hundreds of seminars and, more recently, webinars all over the world. One example in the mid-1990s was an annual 2½-day seminar titled “Engineering and Economics of Reinforced Concrete Buildings” for structural engineering professors. I think being able to train the trainers is an excellent way to impart the knowledge and experience I’ve gained over the years.

During my brief academic career at the University of Illinois at Chicago, I supervised seven Ph. D. and five M.S. candidates, all of whom successfully completed their degree requirements. I also taught seven different courses at the University of Illinois at Chicago, one of which I developed.

What are some highlights of your building codes and standards work?

I served many years as Director of Engineering Services, Codes, and Standards for the Portland Cement Association (PCA). I supervised a sizable staff and administered a significant annual budget in that role. We generated publications and computer programs for use by structural engineers engaged in analyzing and designing reinforced concrete structures.

I was long an official observer on the Seismology Committee of the Structural Engineers Association of California, starting in the mid-1980s. In that capacity, I played a significant role in developing and passing the 1994 and 1997 Uniform Building Code (UBC)

design provisions for reinforced concrete shear walls in regions of high seismicity. Those eventually found their way into ACI 318. More recently, I played a leading role in developing and approving new diaphragm design provisions in ASCE 7-16 (more details about that appear in a March 2016 Structure Magazine article). I was also intimately involved in developments that have led to the inclusion of coupled shear wall systems as distinct seismic force-resisting systems in ASCE 7-22 (read more about that in the September 2022 issue of Structure Magazine).

I’ve been a member of the Provisions Update Committees of the Building Seismic Safety Council (BSSC) since 1989. I’ve been involved in ACI 318 committee and subcommittee work since 1990. I served on the Masonry Standards Joint Committee for many years and withdrew after completion of the 2013 standard. I’ve also been a long-standing member of ASCE 7.

What about all the books and technical publications you have authored or co-authored?

In 2007, PCI published the first edition of its manual, Seismic Design of Precast/Prestressed Concrete Structures, which I co-authored with Ned Cleland. The second edition was published in 2012. This manual is the only book regarding this subject and is widely used by designers in the United States.

The book, Seismic and Wind Design of Concrete Buildings: 2000 IBC, ASCE 7-98, ACI 318-99, which I co-authored, was the only book of its kind at the time and was widely used by those engaged in the seismic design of concrete structures. The book was updated to the 2003, 2006, and 2009 codes and referenced standards. A predecessor book, Design of Concrete Structures for Earthquake and Wind Forces, which I also co-authored, was published jointly by

PCA and the International Conference of Building Officials and was also widely used.

The 2000 IBC Handbook: Structural Provisions, which I coauthored, found wide acceptance. An update of part of this book, Application Guide for the 2003 IBC Concrete Provisions, was also published. The entire handbook was updated to the 2006 IBC and published by the International Code Council (ICC). The 2009 IBC Handbook: Structural Provisions is the latest available edition of the publication. An Application Guide for the 2012 IBC Concrete Provisions has also been published.

What leadership roles have you held, and what would you tell other engineers about stepping into leadership?

I was, until recently, a member of the Board of Governors of the Structural Engineering Institute (SEI) of ASCE. I served as a member of the Board of Direction of the Earthquake Engineering Research Institute (EERI), the American Concrete Institute (ACI), and the Building Seismic Safety Council (BSSC). I’ve also chaired various technical conferences and BSSC Issue Teams. I would tell other engineers that leadership roles provide you with opportunities to advance your profession, and there is much satisfaction in that.

You’ve received dozens of awards. Which one or two are most meaningful to you and why?

I think the two most meaningful were: Titan of the Precast/ Prestressed Concrete Industry Award, 2004, because the list of other recipients, such as T. Y. Lin, made me wonder how anyone could think I belonged in that company and 2) Walter P. Moore

Jr. Award, Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE), because of its association with Walter Moore whom I truly admired and because it is for my codes and standards work, which has been so important to me.

What advice would you give to younger engineers regarding a successful career path?

Try to understand what you are doing beyond the computer output. Make sure you comprehend things at the most basic level. For example, load path is the most fundamental concept a structural engineer must always keep in mind in the design process.

Who was your mentor or who influenced you most as a young engineer?

Mark Fintel brought me to PCA and groomed me to be his successor. Mark produced some great work in collaboration with Fazlur Khan. Two things he said have stayed with me all these years. The first was: “You and I have a nationwide constituency.” In other words, we had to be aware of questions and problems facing structural engineers no matter where in the country they were located and had to try to provide solutions. The second was: “No project is ever completely done until it is published.”

Closing thoughts?

After five decades in the profession, I still enjoy what I do. I have found structural engineers, in general, to be remarkably honest. Ours is a noble profession because we build, as opposed to destroying, and rebuild after destruction, which is God’s work.■

Building higher

Photo courtesy of Richard Jarvis, Owner’s Representative

Emory Winship Hospital Pedestrian Bridge

By Georgi I. Petrov P.E., A.I.A.,, Raymond Sweeney P.E., Darin Appleton P.E., and Charles Besjak P.E., S.E., F.A.I.A.

Pedestrian bridges are a vital part of modern hospital complexes, providing weather-protected connections for patient transfer and staff connectivity. They are also a great opportunity for merging engineering creativity with architectural expression. Emory University Hospital Midtown is a large hospital in the SoNo district of Atlanta, Georgia, with an over 100-year history.

In the spring of 2023, a new building for the Winship Cancer Institute, designed by Skidmore, Owings, & Merrill (SOM) and constructed by Batson-Cook Construction (BCC), was completed across Linden Avenue from the main hospital campus. The hospital planned a doublelevel bridge to link the new building to the rest of the complex. The lower level would connect public spaces on both sides, and the design team wanted it to be as open and transparent as possible. The upper level would provide patient access and was required to have fritted glass to offer the necessary patient privacy. The bridge would also carry mechanical, electrical, and plumbing services across the street to the new hospital.

Design

When designing an efficient truss, it is common to adhere to typical span-to-depth ratios for a single level truss. Thus the sides of the top level

of the bridge are full-depth trusses, and the lower level hangs from HSS hangers at each truss point. Besides these efficiency and constructability considerations, this choice also reflected the functional differences between the two levels. The top private level is more enclosed, while the lower public level is as transparent as possible.

One challenge for the designers was that the elevations of the corresponding levels between the old and new buildings do not align. The lower level has to reconcile an elevation change of 4 feet 1 inch and the upper level a change of 1 foot 4 inches. It was decided to connect the two springing points with straight and slightly inclined walking surfaces to meet the maximum allowable slope. The team then investigated several

options for forming the geometry of the structural elements to accommodate these elevation constraints. The solution was to keep the top and bottom chords of the truss horizontal to simplify fabrication, but more importantly, to preserve the clarity and purity of the truss geometry.

Additionally, this configuration maintained a consistent node geometry to facilitate fabrication and provided a regular datum for the attachment of the facade modules. The slope of the walking surface was achieved using a slab-on-grade over-shaped geofoam. The bottom level framing, which hangs from the truss, was sloped to follow the walking surface. The lower level facade units thus extend down from the bottom of the structure along the bridge in order to avoid trapezoidal glass elements.

Between the existing hospital and the new building lies the busy three-lane Linden Avenue, a patient drop-off area for the current hospital’s Peachtree Building, and an exterior courtyard. While locating supports for the new pedestrian bridge adjacent to the street would provide the shortest spans for the bridge, it would have resulted in the significant rework of the dropoff area and entrance to the Peachtree Building. Rework for both of these areas was not an option for the facility to remain open during the bridge construction. As a result, the team determined that the most appropriate approach was to provide one 134-foot-long span across both the street and drop-off zone, supported by the Winship building to the north and a pair of columns in the landscaped courtyard to the south. From there, a shorter adjacent span and second support location carry the cantilevered south end of the bridge structure. The cantilevered end of the bridge eliminates two potential problems. First, it allows the bridge to interface with the Peachtree Building without imposing new loads on the existing building’s structure while simultaneously locating any new bridge foundations away from the

existing footings. This approach permitted the continuous operation of the Peachtree Building with no structural enhancements required. All told the bridge covers 184 feet in length between the two buildings with minimal impact on Linden Avenue, the hospital drop-off, and the existing building.

Lateral support for the bridge structure is provided in three locations; at the connection to the new Winship Building and at each vertical support in the existing courtyard. At the Winship Building, lateral forces are transferred from the bridge into the building structure through a series of sandwiched bearing plates at each level and distributed into the reinforced concrete floor framing and diaphragms. At the two bridge supports to the south, moment-resisting frames provide the lateral strength and stiffness required for the bridge. The frames consist of 24-inch square built-up columns in combination with W24 beams. In-plane diaphragm bracing of the steel framing and composite metal deck slab distribute the lateral loads from the bridge spans into the lateral supports.

Novel Optimized Truss Geometry

The primary distinctive feature of the bridge is the upper-level truss that utilizes web members arranged in an innovative geometry that increases the structure’s

Design space and form finding diagrams.

View of the bridge being constructed. Photo courtesy of Batson-Cook Construction

efficiency. SOM used several academic and internally developed structural optimization tools that iteratively remove underutilized material to arrive at the most efficient structural form. These tools help to determine the optimized typology and geometry to minimize total material usage for the given constraints. For example, one program, Polytop, starts with a design space of a solid 2D continuum of material and iteratively removes material to arrive at a solid-like approximation of the most efficient structural form. The other, Ground Structure, utilizes a design space of a densely interconnected grid of linear elements and iteratively removes members to arrive at a sketch-like approximation of the most efficient structural form. The results from these programs are then combined with an applied rationalization considering the fabrication and construction of the bridge. The resulting geometry consists of truss bays with skewed X-bracing that is symmetrically oriented about the mid-span of the bridge. SOM has successfully designed a number of high rises with a similar geometry for vertical bracing, including the recently completed 100 Mount Street in Sydney, Australia, and 800 Fulton Market in Chicago, Illinois. However, the Emory Winship at Midtown Bridge is the first constructed example of a long-span structure using this truss geometry.

The truss diagonals were considered Architecturally Exposed Structural Steel. As such, the team made many design choices for aesthetic purposes and to facilitate construction. First, the decision was made to maintain a consistent outer dimension for all the truss diagonals within each truss bay to ensure visual continuity of elements and avoid connections with members of varying dimensions. Similarly, changes in member dimensions between bays were kept to a minimum to transition along the truss seamlessly. At the asymmetric connection of the truss diagonals, a milled steel node was used to provide a clean, easily repeatable piece for a consistent connection of the diagonals. The resulting truss consists of W12x96 top and bottom chord members, while the truss diagonals comprise HSS4x4 through HSS 8x8 elements.

Erection

The Emory Bridge spans Linden Avenue, a highly trafficked road that functions as an on-ramp to Interstate 85. As the team engineered the erection sequence, they wanted to minimize the time the road was closed to traffic. In addition, the team wanted to erect the largest sections possible to minimize the amount of work that would take place over the road. The trusses were fabricated and assembled in the shop and delivered to the site in 40-foot pieces. One lane of Linden Avenue was shut down and used as a laydown area to assemble the pieces into large truss sections on the ground and minimize the amount of overhead welding that would need to be done on the bridge while suspended. Next, the top and bottom chords of the trusses were welded together, and then the intermediate beams were installed to create two large box trusses that could be erected quickly and safely over the road.

Once the box truss sections were complete, a total road closure for Linden Avenue was instituted, and two mobile cranes were used to lift the two box trusses into place. These sections were bolted together while suspended in the air, and the cranes remained hooked to the trusses until all back welds were complete. Next, a team of four ironworkers worked through the night to complete 12 large completepenetration welds and the bridge tie-in. After 23 hours of continuous work, the welds were inspected, and the cranes released the trusses. Over the next several days, the bridge’s lower-level was hung from the trusses, allowing Linden Avenue to be reopened. The deflection of the bridge was measured and compared with predicted values at the completion of the trusses, after the erection of the lower level steel and after the placement of each of the three slabs.

Conclusions

The new pedestrian bridge connecting the Winship Cancer Institute to the rest of the existing Emory Midtown Campus utilizes a steel truss with a novel optimized geometry providing an efficient and elegant span over Linden Avenue.■

Georgi I. Petrov, P.E., A.I.A., is a Senior Associate Principal in the New York, NY office of Skidmore, Owings & Merrill. He can be reached at (georgi.petrov@som.com).

Raymond Sweeney, P.E., is an Associate Principal in Skidmore, Owings & Merrill.

Darin Appleton, P.E., is Senior Vice President/General Manager at BatsonCook Company.

Charles Besjak, P.E., S.E., F.A.I.A., is a Principal and leader of structural engineering in the New York, NY office of Skidmore, Owings & Merrill.

Interior view of the bridge. Photo courtesy of Dave Burk © SOM.

structural MODELING

BIM of Cold-Formed Steel Framing

The key is discovering value and avoiding waste.

By Daniel Stadig, P.E., Kirsten Zeydel, S.E., & Jesse Hasenfus

Building Information Modeling (BIM) is used extensively throughout the construction industry, with some trades fabricating directly from BIM as standard practice. Even steel rebar for cast-in-place concrete is being modeled and fabricated using BIM. So why isn’t all Cold-Formed Steel (CFS) framing included in a model, and why do some CFS contractors consider BIM a waste of time?

What Makes CFS BIM Unique?

Unlike most structural materials, the application of CFS framing on a project often varies significantly. For many projects, the CFS framing is a non-bearing component like interior walls/ceilings or exterior envelope framing, where many member locations are highly adjustable in the field. However, in other projects, the CFS framing represents part of the primary structural system where member locations are critical. Therefore, the strategic use of CFS BIM needs to acknowledge these differences.

Modeling

Additionally, the volume of parametric data used for CFS framing models can be significant. For example, with the spacing of CFS framing at 16 or 24 inches on-center, CFS models can require thousands of model objects. If connectors and fasteners are required to be modeled, the number of objects can quickly increase two-fold. While the latest software can automatically populate CFS framing into wall objects, the model volume and maintenance can still be costly. So, when is the reward worth the cost?

CFS Modeling and VDC

It is important to understand the BIM terms most commonly used by design teams and Virtual Design and Construction (VDC) teams. Most contracts requiring BIM deliverables include a BIM Execution Plan (BXP). The BXP typically makes reference to the BIM Forum Level of Development (LOD) Specification for the required amount of modeling and detail for every scope involved. Below are the LOD specifications for CFS framing:

•LOD200 – Generic geometry modeled to represent an overall CFS assembly or system, approximate in terms of overall size and shape. (Modeling of individual CFS framing members is not required.)

ˏ Allows for minimal trade coordination and only very basic takeoff.

•LOD300 – Geometric objects modeled to represent an overall

CFS assembly or system, accurate in terms of actual overall size and shape, including nominal openings. (Modeling of individual CFS framing members is not required.)

ˏAllows for minimal trade coordination and basic takeoff.

• LOD350 – Critical CFS framing modeled as individual elements for coordination purposes. Openings are modeled to actual rough-opening dimensions and include critical framing members such as jambs, headers, and sills. Diagonal kicker braces are also modeled for coordination purposes. Typical on-center infill framing may be omitted per a project’s BXP.

ˏ Allows for trade coordination with critical elements and standard takeoff.

•LOD400 – All CFS framing and connections are modeled as individual elements, accurate in terms of size, shape, and location.

ˏAllows for comprehensive coordination, exact takeoff, and is fabrication-ready.

VDC teams are accustomed to mechanical, electrical, plumbing, and fire protection (MEPF) trades producing BIM models varying from LOD350 to LOD400, which facilitates the prefabrication of many of these systems. CFS framing is a unique player in the VDC coordination space, as many CFS contractors choose to field frame the CFS system. Field framing allows the CFS contractor to easily make adjustments in the field. Therefore, determining the adequate LOD for CFS BIM should always be the first step. Adhering to a lower-value LOD may help save the overall project time and money while still offering significant coordination benefits.

Note: While LOD400 may be required for MEPF systems, the CFS framing may be best utilized at an LOD350 without the infill framing.

only critical members reduces file size and speeds up processing

General Value for Engineers and Contractors

CFS BIM can provide CFS engineers and contractors with useful information to help with critical decision analysis, add value through enhanced coordination, and help to ensure better quality control. Precise model-based material takeoffs, generation of shop drawings, and even running machinery at fabrication facilities are possible through CFS BIM implementation. The benefit of modeling complicated geometries in 3D can help determine how CFS framing can achieve a particular architecture and also serves as a very effective visual aid for collaboration with design teams and field crews. CFS BIM offers value to CFS engineers and contractors in various ways, but there is no rule book to follow when choosing which methods and techniques to adopt as standard practice. It is important to explore what works best case by case to achieve the best outcome. Note: Deciding to implement BIM internally often requires a longterm outlook. Added quality control and coordination may not realize immediate tangible benefits, but long-term return on the investment can outweigh the upfront cost.

Field-Framed Non-Bearing Exteriors

In most buildings, minimal MEPF items run through or penetrate the CFS envelope framing. Thus, CFS BIM may not significantly impact external VDC team coordination efforts, especially if the CFS framing is field-installed. However, building exterior walls with significant MEPF can benefit from CFS BIM. Examples include process structures, data centers, and lab facilities.

Note: Useful model coordination items can include base and top-of-wall conditions, slab edge clearance, tight tolerance conditions, and wall layout against other trades.

Comparing a CFS framing model to the architect’s and structural engineer’s models is an easy way to uncover potential design issues. For example, soffits and overhangs are common locations where deeper structural members may be required but aren’t clearly shown in architectural sections or details. CFS

BIM allows CFS engineers and contractors to develop solutions with the added assurance of fit and coordination.

Note: When working on a structural steel building, consider requesting the steel fabricator’s model and not the structural engineer’s model since the fabricator’s model represents the anticipated field condition. Similar considerations apply to concrete buildings; however, not all concrete gets modeled by the contractor.

Modeling diagonal kickers that extend into a building’s interior space can identify conflicts with MEPF systems and interior ceilings. Typically, kickers function as critical support members, and adjustment of their position or location can be limited. Identifying any issues before construction allows systems and designs to be altered without re-work in the field, saving everyone valuable time and money.

Field Framed Non-Bearing Interiors

BIM for non-bearing interior CFS framing primarily benefits the VDC team coordination process in buildings with dense MEPF systems. For coordination purposes, BIM models for interior non-bearing CFS framing are typically comprised of critical framing members only. These include jamb studs, headers, sills, end studs, and corner studs. This aligns with LOD350 and helps the MEPF trades identify where their systems need to be adjusted or re-routed away from critical CFS members.

Modeled objects for critical framing members do not have to be exact representations of the framing in the field. For example, rationalized geometry can represent the framing members to encapsulate the worst case. However, geometries must not be modeled overly conservatively, as this may lead to unnecessary coordination efforts from other involved parties.

Note: Jambs and headers can be modeled using a single worst-case size to simplify model insertion.

For top-of-wall conditions, especially fire-rated wall assemblies, MEPF must route their systems in a manner that does not impede the construction of the wall or prevent the assembly from functioning as intended. Modeling no-fly zones are a simple and effective way to coordinate and avoid framing issues without the extra effort of modeling all the individual framing members. Note: Top-of-wall no-fly zones can be modeled with solid extrusions, or top-of-wall tracks can use an elongated flange to create the no-fly zone.

Field Framed Load-Bearing Structures

For some buildings, the primary structural system utilizes load-bearing CFS framing. Minimal field adjustments are allowed with this system, as load-bearing CFS framing usually takes priority over other building systems. However, this may not hold in all cases, as some individual MEPF systems cannot adjust. For example, HVAC ductwork can be too large to fit between stud bays resulting in additional framed openings needing to be engineered, and consideration must be taken to avoid MEPF conflicts with shear walls and strap bracing.

Note: Toilet, shower, and tub drains, especially on floor joists, must be carefully coordinated as these drains can only be relocated by reconfiguring the space.

LOD400 vs. LOD350 without infill framing.

There are situations where CFS BIM, while helpful, may not be cost-effective. The MEPF trades may not have the expertise or resources to provide models for certain projects. Additionally, if the MEPF trades do not intend to fabricate according to the BIM model, there is not much guarantee that a model aids coordination. BIM for load-bearing CFS framing is typically beneficial, but when BIM lacks other trade involvement, it may not be pragmatic, resulting in a wasted effort.

Panelized Exterior or Interior Panels

(For this article, an unfinished panel is defined as a panel that is either just CFS framing members or CFS framing members with sheathing installed.)

Utilizing BIM for prefabricated unfinished CFS panels allows for preliminary coordination of stud layouts, panel joints, panel interactions with other components, and panel connection points to the structure. Minimizing field adjustments during installation is paramount to the success of prefabrication. A huge benefit of prefabrication is the speed of installation. Utilizing CFS BIM and MEPF coordination before panel fabrication reduces the need for field adjustments and ensures the most efficient installation process for both CFS contractors and MEPF trades.

Note: Coordination with every MEPF item is usually unnecessary for unfinished panels, as some adjustments in the field are acceptable. For example, sheathing can easily be cut in the field, and intermediate framing can be added to support MEPF systems that fit within a stud bay.

Panelized Exterior Finish Panels

(For this article, an exterior finish panel is defined as a panel that includes the air/weather barrier and exterior finish materials in addition to the framing and sheathing.)

BIM for CFS exterior finish panels is usually necessary

due to the precision required to properly design and locate all penetrations and attachments of anything that interfaces with the finish panel. As the air/weather barrier is usually covered by finish and no longer accessible, extreme care must be taken to ensure the air/weather barrier remains intact and functional to prevent performance issues. Items such as security cameras, light fixtures, louvers, vents or grilles, keypads, signage, and overflow scuppers are commonly occurring items that require coordination. BIM provides an excellent opportunity to ensure these items are properly located and installed.

Note: Not utilizing BIM for exterior finish panels is risky and can lead to costly field repairs that could compromise the appearance or performance of the exterior envelope.

Using simplified shapes can increase efficiency while achieving the same coordination goals.

Finish panels do not have the same opportunity for field modifications as bare framing.

Panelized Load-Bearing Structures

Load-bearing structural CFS panels can be installed at impressive rates, especially when multiple cranes are used. Some projects have 50% of the CFS panels fabricated before the first panel is set to keep up with this pace. Since issues in the field do not appear until most panels are fabricated, it is critical to utilize BIM to find and resolve conflicts before panel fabrication. Field modification of panelized CFS framing can have a detrimental effect on a project. Even moving one stud has the potential to cascade into a series of misalignments on levels above or below. Thus, CFS BIM for panelized load-bearing structures is almost always a necessity.

Recent Advancements

Cutting-edge integrations have allowed contractors to harness the ability to efficiently build directly from BIM with amazing results and accuracy. Examples of these advancements include CNC roll formers and robotic assembly lines. Information from BIM is exchanged with the equipment via output files. Output files are essentially digital instructions given to equipment so material can be formed, cut to length, and even printed with labels for assembly purposes. Output files can also control fully automated robotic assembly lines that assemble, screw, or weld members into panel assemblies.

Though CNC roll formers and robots are precision pieces of equipment designed to handle immense work volumes, downfalls exist with this technology. The equipment and software can be expensive to own and operate. Robots may struggle to accommodate fit-up tolerances of material causing problems during panel assembly. Machines may also need to run slower to achieve adequate accuracy and may jam while manufacturing complex or short pieces.

Final Considerations

As computers and software continue to improve, so will the benefits of BIM. However, heavy upfront modeling and coordination are required and generate costs that must be addressed by the parties involved. Therefore, it is important to remember the end-use of CFS BIM and how it relates to the CFS framing at the job site. When CFS framing is not panelized, or the CFS contractor does not intend to use BIM during installation, a full-scale CFS BIM effort may be a waste of time and money. However, the benefits can outweigh the costs when CFS BIM is utilized intelligently.■

Daniel Stadig, P.E. is Vice President at Salas O’Brien and specializes in ColdFormed Steel framing design and BIM. He is the immediate past Chair of the Cold-Formed Steel Engineers Institute (CFSEI) Executive Committee and the current Chair of the CFSEI BIM Committee. He also serves as a corresponding member of the AISI Committee on Specification. (daniel.stadig@salasobrien.com)

Kirsten Zeydel, S.E. is the Director of Design at Nevell Group, Inc. She is an active participant in the AISI Code Committees. Her passion is ensuring that CFS framing meets the building code and architectural requirements and is easy to manufacture and install. (kzeydel@nevellgroup.com)

Jesse Hasenfus is the BIM/VDC Director at Excel Engineering, Inc., in Wisconsin. He is a current member of the Cold-Formed Steel Engineers Institute (CFSEI). (jesse. hasenfus@excelengineer.com)

INSIGHTS

Steeling the Show

The benefits of steel framing in curtain wall systems.

By Jose Rodriguez

Whether used within the built environment or as an exterior façade, curtain walls made with transparent glazing can enhance the overall design of a building by improving access to daylight and helping to stabilize interior temperatures with U-values as low as 0.19. While discussion around these systems tends to focus on the glass, the framing systems are an equally important aspect to consider.

Choosing the right framing material can help optimize the benefits of a transparent curtain wall system. For example, the recent terminal expansion of Oklahoma City’s Will Rogers World Airport turned to cold-formed, structural steel framing systems to provide the largest possible spans and maximize the glazing area. However, the benefits of steel framing extend beyond strength to include improved thermal performance and system durability.

Increases Free Spans

Steel framing offers almost three times the strength of aluminum systems with a Young’s modulus of 29 million pounds per square inch (psi)—compared to aluminum’s 10 million psi. With greater strength comes larger spans and larger panels, or lites, of glass with similar shapes. Not only does this maximize the glazing area for better views and a greater percentage of visible light transmission, but it allows thicker glazing infills—over 3 inches (in) for some systems—which can help accommodate fire-rated glass when needed.

Further, steel frames can span heights much more significant than aluminum systems without the need for bulky and expensive support systems. Given a designed wind load of 30 pounds per square foot (psf), 5-foot (ft) vertical to vertical module spacing, and using captured rectangular verticals, steel curtain wall framing can span heights up to 21.41 ft without reinforcement (compared to aluminum’s 12.5 ft). The Will Roger World

Airport’s architectural team used this strength in conjunction with the building’s intermediate vertical structural members for anchorage to span heights of 33 ft without greatly increasing the frame’s dimensions.

Improves Thermal Performance

Low-emissivity, double- or triple-glazed units, and other energy-efficient glass can significantly improve the thermal performance of a glass curtain wall system. In addition to providing the strength to hold these types of glazing, steel framing also contributes to stable interior temperatures. Steel’s thermal conductivity is approximately 74 percent less than aluminum’s, reducing the amount of heat transferred through the framing.

Comparable to both glass and concrete, steel’s thermal expansion

Cold-formed steel framing provides a close visual match to fire-rated and non-rated systems. Photo courtesy of Technical Glass Products.

Narrow-profile steel framing systems span larger heights and increase glazing areas. Photo courtesy of Simon Hurst Photography.

coefficient (0.00065 inch/inch/degree in Fahrenheit (in/in/deg F)) is almost half that of aluminum’s (0.00128 in/in/deg F). Reducing the amount of expansion the framing undergoes helps ensure a sound building envelope as temperatures change. Likewise, steel frames require fewer expansion joints, resulting in a more tightly knit curtain wall system. All these qualities improve the insulative capabilities of exterior curtain walls, decreasing energy consumption from HVAC systems. This works with daylight access, which limits the amount of electric lighting needed, to help buildings more readily achieve LEED credits.

Provides a Close Visual Match Between Fire-Rated and Non-Rated Systems

Cold-formed, structural steel framing can extend its benefits beyond a singular application to enhance a building’s design. Most notably, due to the strength and durability of these systems, they can provide a close visual match between non-rated and rated curtain walls.

Steel framing can be fire-rated for up to two hours if a fire rating is required. For fire-rated applications, the glass becomes heavier (and often thicker). In response, the frames must be stronger. In the past, this often meant enlarging a frame’s width and depth or adding either internal or external reinforcements to the system. Steel framing can provide fire-rated curtain walls with the strength to support large and heavy lites without dramatically increasing frame size or compromising the fire rating of the entire system.

This results in a narrow-profile frame that does not stand out compared to adjacent systems. For example, the Fulton Center in New York City used a fire-rated steel framing system to meet the requirements of its upper-level curtain walls while maintaining a seamless aesthetic with the non-rated curtain walls of the lower levels.

Exceed Historical Limitations

With all these benefits, some may wonder why steel is not used more prominently. Historically, the use of steel has been limited by concerns about corrosion in exposed conditions and an inability for manufacturers to form mullions precisely, resulting in bulkier frames. However, galvanizing these systems before assembly, along with both liquid and powder coat finishes, provides durable layers of protection against rust without compromising the design versatility of these systems. Additionally, many steel systems rely on rain-screen or pressure equalized principles for managing air and water resistance. For these systems, the primary seal at the back of the glazing pocket is extruded, with continuous gasketing across the face of all framing members. It provides increased air and water resistance. This bolsters the ability of a system to ensure a sealed building.

All Framed Up

Transparent curtain walls provide building professionals with many design possibilities—some of which were not possible even

20 years ago. While much of the conversation around these systems has understandably centered on glass, framing materials play an integral role in a curtain wall’s ability to meet and exceed design specifications. When steel framing is used, designers and engineers can increase the lite size and span height without vastly changing the frame shape, thereby maximizing the glazing area. Steel’s thermal performance capabilities also help curtain wall assemblies more significantly contribute to sustainable building initiatives like LEED and the Living Building Challenge.■

Steel-framed curtain walls improve access to daylight and visual connections between the inside and the outside. Photo courtesy of Technical Glass Products.

Jose Rodriguez is a Lead Mechanical Engineer at Technical Glass Products (TGP). He has over 20 years of experience engineering and managing curtain wall projects. Contact him at: (800) 426-0279.

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iconic STRUCTURES Statue of Liberty National Monument

Safety renovations for the Statue of Liberty.

By Denise L. Richards, P. E. and Thomas J. Normile, P. E.

When Keast & Hood, Structural Engineers were contacted one day in April 2009 and asked to be at the Statue of Liberty the following Monday, the authors had yet to learn what to expect. What followed was a four-year journey that would take them to nearly every interior corner of the iconic National Monument and UNESCO World Heritage Site. That journey, with a design team led by Mills + Schnoering Architects working under the auspices of the National Park Service, resulted in significant life-safety improvements to the Monument as described in the June 2014 Structure article Designing Life Safety Renovations for the Statue of Liberty

That article focused on the recently completed interior renovations to improve life safety and egress from the Monument’s pedestal and surrounding 200-year-old fort. In this article, we delve deeper into the history and structural details of the Monument.

Fort Wood

Before the Statue of Liberty on Liberty Island, there was Fort Wood on Bedloe’s (alt. Bedloo’s) Island, named for Isaac Bedloo, a former burgher of New Amsterdam whom the English granted the land after they captured the area in 1664. After changing hands several times, the island was conveyed first to the City of New York, then to the State of New York and ultimately, in 1800, to the United States Government. In 1806, as part of a larger effort to improve the defenses of New York Harbor, construction of a “star fort of masonry having a stone magazine, with barracks of brick, and a brick arsenal” was begun. The work was completed in 1811 and, in 1814, renamed Fort Wood for Lieutenant Colonel Eleazer Derby Wood, an 1806 graduate of the US Military Academy who died in 1814 in the siege of Fort Erie, one of the last battles of the War of 1812.

Fort Wood was manned throughout the War of 1812, although it never came under attack. The fort was fortified, improved, and repaired continuously in the period between 1814 and 1884. The result of those improvements is the fort as we see it today from the outside: an imposing, monumental structure with a commanding presence as seen from the surrounding New York Harbor. That presence is due, in large part, to its steep scarp (the front of the rampart or outer fort wall) faced with massive units of ashlar-cut grey/blue granite rising 30 feet from the surrounding grade. The 3ʹ-0 granite facing was added to the rubble masonry of the scarp of the Fort in 1844. The granite and rubble were bonded with concrete.

Through the mid-1800s, the Fort’s use evolved. It was used alternately as a Corps of Artillery garrison, ordinance depot, quartermaster depot, hospital, and recruiting station. That period was one of international peace, and, despite the continuous improvements, military activity

at the Fort gradually diminished.

In 1877, President Grant signed a joint Congressional resolution that authorized the President to make suitable provisions to accommodate the impending gift of a Statue from France on either Bedloe’s or the adjacent Governors Island. At that point, the remaining four officers and 32 men promptly moved out of Fort Wood, and the Fort’s military history ended.

Genesis Of The Monument

As most Americans know, the Statue of Liberty was a gift from France. The concept of the gift evolved in the late 1860s amid a group of French intellectuals enamored with the United States Constitution, its guaranteed rights and liberties, and the fact that its principles and ideals had recently survived the test of the American Civil War. The original idea was for the Statue to be presented to the United States in 1876, the centenary of the Declaration of Independence. It was another decade until the Statue was completed and opened on Bedloe’s Island.

In the early 1870s, the Franco-American Union was established to arrange for the Statue’s funding, creation, and installation. The French Committee was responsible for the Statue itself, while the American Committee was responsible for the pedestal and its siting. By the end of 1877, the US Government had officially selected Bedloe’s Island for the Monument’s location. The American Committee decided to locate the Monument within Fort Wood so that the Fort would appear to form the pedestal’s base.

View of the Statue of Liberty and Fort Wood taken from the harbor. Photo Courtesy of Brian Rose Photography.

The Pedestal

In 1881 the American Committee selected French-trained American Architect Richard Morris Hunt to design the pedestal of the Statue while US Army General Charles P. Stone was already at work designing the foundations.

True to the Committee’s intentions, Hunt and Stone produced a three-part design:

• Below Grade Foundation: Approximately 90 feet square and 15 feet thick foundation of poured concrete akin to a modern mat foundation. At the time of pedestal construction, the original parade ground inside of and below the terreplein was still there, and this portion of the foundation topped out at the parade ground level. The parade ground was later backfilled up to the terreplein level.

• Above Grade Foundation: Formed, cast-in-place concrete in the shape of a truncated pyramid, tapering from approximately 70 feet square to 40 feet square at its top - 60 feet above the parade ground. This pedestal section was originally a solid concrete mass except for a 10ʹ-0 × 10ʹ-0 shaft at its center and an access tunnel on one side. The shaft was later enlarged on one side by demolishing concrete to make more interior space for an internal stair starting at the base.

• Pedestal Proper: Approximately 38 feet square (outside dimension), granite-clad, cast-in-place concrete, rising approximately 85 feet from the top of the above-grade foundation to the base of the Statue. This pedestal section originally had an interior open shaft of approximately 25 feet square. Multiple sets of 50-inch-deep wrought iron girders traverse the pedestal proper. One set of girders at the top of the pedestal support the corner columns of the central pylon of the Statue. Another set, approximately 50 feet lower, is connected to the upper girders with wrought-iron eye bars and embedded in the pedestal walls. These lower girders resist uplift due to the overturning moment of the Statue under wind loading by engaging the weight of the pedestal. At various times, the pedestal girders have been referred to as “Eiffel Girders.” We expect that, at a minimum, Eiffel would have stipulated the forces to be resisted by the girders. Whether the girders were actually designed by Eiffel or an American engineer on the pedestal design team is unknown. It is known that the girders were fabricated by the Keystone Bridge Company in Pittsburgh, PA.

The Statue

French sculptor Frederic Auguste Bartholdi was well-known within the circle of intellectuals who first discussed the idea of the Statue of Liberty in the late 1860s. By 1870, Bartholdi was at work designing the statue of “Liberty Enlightening the World,” and by 1876, a team of French artisans and craftsmen began constructing the Statue in Bartholdi’s workshop in Paris.

For the design of the Statue’s structure, Bartholdi, in 1875, first consulted with the well-known French architect and engineer EugèneEmmanuel Viollet-le-Duc. In that era, a metal-skinned statue would commonly have had an interior masonry structure. However, Violletle-Duc proposed a stability system of compartments filled with sand from the base to the hip and an iron armature above the hips.

When Viollet-le-Duc died in 1879, the engineering work was passed to the company of Gustav Eiffel, already known for his mastery of aerodynamics and light structural latticework in the design of a series of elegantly proportioned railway bridges in France. (It wasn’t until the World’s Fair of 1889 in Paris that he did the Eiffel Tower).

Eiffel veered away from Viollet-le-Duc’s sand-filled compartments and designed an all-iron structure. Four central angle-iron columns, each with unequal legs measuring 28 × 24 inches, extend from the Statue’s base to the shoulders’ height, leaning inward as they ascend. The columns are laced together with smaller iron angles and form a vertically-cantilevered box truss or pylon – similar to a modern high-voltage electrical transmission tower. The central pylon measures approximately 7 × 6 feet at its top and 16 × 13 feet at its base. From this central pylon, a series of orthogonal and diagonal struts and ties are cantilevered to support the flat, wrought-iron armature to which the exterior, 3/32” thick copper skin of the Statue was connected.

This stair descending into the terreplein exposed the interior of the outer Fort wall (rampart) and the remnant brick arch of a bombproof shelter inserted below the terreplein in the mid-1800s. Photo courtesy of Brian Rose Photography.

The completed foundation and pedestal in April 1886 showed original access by wooden stairs. Photo courtesy of the Library of Congress.

The four main columns of the pylon land on the girders spanning the top of the pedestal.

The original design called for the bare concrete section of the foundation from the terreplein up to the third-level to be hidden beneath a planted berm traversed by monumental stairs to the third-level doorways on each side of the pedestal. For the first 20 years of the Monument’s

life, the concrete foundation was exposed, and public access to the pedestal was via exterior wood stairs to doorways on the third level. The berm and one set of permanent stairs were not completed until about 1907.

American Museum of Immigration (1960s)

In 1955, after decades of discussion about the lack of interpretive content at the Statue of Liberty and, separately, discussions about locating an immigration museum in Manhattan, the American Museum of Immigration, Inc. (AMI) was chartered by New York State at the base of the Statue of Liberty.

The design called for the museum to be located beneath the terreplein in the space between the tiered upper foundation of the pedestal and the exterior walls of Fort Wood. The parade ground of the original Fort had been filled in up to the level of the terreplein in 1907. This fill had to be removed to the level of the fort’s original parade ground in order to make way for the museum. Sadly, all remnants of the original fort structures within its walls were demolished and removed as well. The one surviving structure from the interior of the fort was the south sally port.

The AMI structure consists of cast-in-place concrete walls and pan joist floor systems with spread footing foundations. Expansion joints at the four corners of the pedestal divide the AMI into four structurallyindependent sections, and some slab sections bear on the stepped pedestal foundation. Construction of the AMI began in 1961 and was completed in 1972.

Centennial Restoration (1980s)

In the early 1980s, plans formed within the US Government to restore the Statue of Liberty National Monument to celebrate the 1986 centennial of the Statue’s opening. The New York architectural firm of Swanke Hayden Connell (now defunct), in association with the firm of French-born, New York-based architect Thierry W. Despont, was engaged to lead a multi-discipline consultant

Photograph by Albert Fernique of Bartholdi’s warehouse/workshop in Paris, circa 1882. Photo courtesy of Library of Congress.

Gustave Eiffel’s drawing for the Statue framework was sent to Richard Morris Hunt circa 1881. Image courtesy of Library of Congress.

Looking up within the core of the central pylon rising within the interior of the Statue. The pylon’s central core is occupied by the double helix stair that takes visitors from the top of the pedestal into the crown of the Statue and back down via the nested helix. Photo by Jet Lowe for HAER, courtesy of Library of Congress.

team to design and document the planned repairs and improvements to both the Statue and its pedestal. Ammann & Whitney (subsequently part of the Louis Berger Group and now WSP) served as the structural engineer for the project and worked with the French engineering and industrial research firm Cetim and Lehigh University for materials testing, fatigue modeling, and other advanced material science analyses. Lehrer/McGovern (subsequently Lehrer McGovern Bovis and now Lend Lease) served as the construction manager. Generally speaking, the work included the following:

• Replacing the wrought-iron flat-bar armature to which the copper skin is connected. After extensive analysis and material testing, Type 316L stainless steel was selected to replace the wrought iron bars. Extensive galvanic corrosion had resulted in significant section loss in the bars, and the iron bolts at the splices in the bars had corroded and failed. Stainless steel bars were made to match the dimensions of the original iron bars, and then custom bent to match the shape of the bar it was replacing. In all, approximately 1,800 bars were replaced.

•Evaluation of and localized improvements to the main structure. This work included blast cleaning and zinc-rich and epoxy coating of both the central pylon and the secondary angle strut and tie system that connects the skin’s armature to the central pylon. Also included in this work was strengthening the connection of the structure for the right arm to the central pylon. When the Statue was first assembled in Paris and presented to the United States ambassador to France on July 4, 1884, the arm’s position was altered, resulting in a deviation from Eiffel’s design. Some reinforcing of this joint was completed in the 1930s when the Statue was still under the control of the U.S. War Department. However, in the 1980s, the engineering team determined that additional improvements to the arm-to-pylon linkage were needed.

• In addition to reconnecting the copper skin to the new, stainless steel armature bars, cleaning and numerous repairs to both the interior and exterior of the copper skin, including the replacement of the spikes of the crown, were completed.

•By the 1980s, the original torch was in poor condition. The windows of the flame enabled the electric lights within to make the flame glow at night and were not watertight. As a

• Replacement of the stairs and elevator within the pedestal; repairs and modifications to the double-helix stair within the Statue.

The hold-down girders that embed and transfer tension forces into the pedestal walls approximately 60 feet below the top of the pedestal. Photo courtesy of Keast & Hood.

Terreplein excavation for the construction of the American Museum of Immigration. Photo courtesy of National Park Service (Statue of Liberty National Monument, STLI 44239).

The upper tiers of the American Museum of Immigration projected above the terreplein shortly after completion. Photo by Jack E. Boucher for HAER, courtesy of Library of Congress.

result, both the copper skin and the iron armature of the entire torch assembly were severely damaged. The restoration team decided they had no choice but to replace the torch. In lieu of the cladding perforated with windows enabling lights to shine from within, which was not part of the original design but, rather, a modification made in the U.S. in the days before the Statue’s 1886 opening, the team decided to replace the torch entirely and follow the original design. The original torch was removed, and a new torch was fabricated on-site at the base of the Monument. The work was completed by a team of French artisans using the same process of hammering copper sheets into wooden molds, known as repousse, that was used by Bartholdi’s team in the initial fabrication of the Statue. The flame was then covered in gold leaf and lit externally to enable it to shine at night.

Life Safety Improvements 2009–2013

In the aftermath of the attacks on the World Trade Center and the Pentagon on September 11, 2001, access to the Statue of Liberty National Monument was significantly curtailed. After several studies on visitor safety, a series of short-term improvements allowed the pedestal to reopen to the 6th level promenade in August of 2004, but the Statue remained closed. In the spring of 2009, the Park Service engaged Mills and Schnoering Architects, Keast & Hood Structural Engineers, and Kalimex Construction to design and install code-compliant modifications to the handrails on the 118 feet tall double-helix stairs within the Statue in 67 days, enabling its reopening in July of 2009.

Six months later, Park Service representatives and a much larger team of design professionals, still led by Mills + Schnoering Architects, embarked on a larger project to improve accessibility and safety at the Monument. At that time, the majority of the pedestal’s interior was essentially one large, open volume 25 feet × 25 feet × 85 feet high with stairs attached to the walls as they

wound their way up. In the pedestal foundation was a narrowed opening 17 feet × 17 feet. Studies had concluded that, in the event of an incendiary incident near the base of the pedestal, that volume could act like a chimney, drawing smoke up through it and making egress challenging for visitors near the top. The primary challenge to the design team: find a way to insert two separate, fire rated, pressurized stairs and an elevator into this volume while avoiding the large, Statue hold-down girders and the narrowed shaft opening in the foundation, fondly described as the “pinch point” during the project.

The complex, three-dimensional puzzle was solved by using Building Information Modeling (BIM) at a time when the technology was still in its early adoption period. By working in three dimensions, the design team achieved a high level of coordination and clash prevention in what was an unusually complex three-dimensional space. Relying on that modern-day advantage just increased our awe for the team that envisioned, designed, and built the Monument 125 years earlier – long before such modern “necessities”, like the computer, were even imagined.

A Monument for all Generations

In the words of the United Nations Educational, Scientific and Cultural Organization (UNESCO), whose mission includes the promotion of international cooperation in education, sciences, culture, and communication, the Statue of Liberty is one of the greatest technical achievements of the 19th century. Further, it is recognized as a bridge between art and engineering. It endures as a highly potent symbol – inspiring contemplation, debate, and protest – of ideals such as liberty, peace, human rights, abolition of slavery, democracy, and opportunity.

The value of this international treasure can and will be its enduring ability to motivate humankind to subscribe to its core principles of freedom, mutual respect, peace, and justice. On an entirely different plane, it can and should serve as both example of and motivation for overcoming obstacles in the pursuit of engineering achievement.■

Full references and acknowledgements are included in the online version of the article at STRUCTUREmag.org.

Denise L. Richards, PE, is a structural engineer, Partner and President of Keast & Hood Co. She worked on the 2009 – 2013 Life Safety Improvements to the Statue of Liberty. She can be reached at (drichards@keasthood.com).

Thomas J. Normile, PE, is a structural engineer and Senior Principal of Keast & Hood Co. He also worked on the 2009 – 2013 Life Safety Improvements to the Statue of Liberty. He can be reached at (tnormile@keasthood.com).

One vignette of the completed 2009-2013 Life Safety Improvements project. Each stair run is independently pressure-positive, fire rated and, separated from the other, as well as from the surrounding occupied spacesArchitectural Design by Mills & Schnoering Architects, LLC. Photo courtesy of Brian Rose Photography.

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Emerging Risks and Claim Trends in the Design Profession

Survey findings offer insight to potentially mitigate claim risk. Part 3

By Dan Buelow and Mark Blankenship

In our previous articles, we discussed several emerging claims trends identified in the WTW A&E Professional Liability Carrier Survey Report: social inflation, cyber liability, climate change, Covid-19, and risk-shifting in contracts. In this issue, we will conclude with an analysis of the enhanced risks associated with certain project types, the most important thing you can do to preserve your professional liability coverage, and the state of the professional liability marketplace.

Project Types

Another question in our WTW A&E survey was to list the top project types that experience the most professional liability severity claims. The responses listed many project types, including residential, apartments, condos, roads, highways, infrastructure, hospitals, schools, arenas, offices, and “other,” including oil and gas.

Residential traditionally has led the field in terms of both frequency and severity. The expectations in residential construction are very high, and residential owners have certain legal protections that are not available to commercial owners. For instance, there’s the implied warranty of habitability. Residential owners tend to need to be more sophisticated about maintenance requirements.

Condominiums have shown the poorest performance of any project type. There are several reasons for this: the developer will never occupy the building, so he has incentives to do things cheaply, especially regarding mechanical systems. Condo unit owners want to spend their money on their units rather than on the maintenance of the common areas. The condo board members can be unsophisticated about maintenance requirements, and the board can be easily intimidated into making claims based on their fiduciary duties to the other unit owners. It is interesting to note that these are primarily human factors.

Right now, apartments are trending badly. This experience is driven partly by inexperienced designers dabbling in the market and partly by shoddy work done by inexperienced contractors, especially “paper” general contractors. We have seen some design firm clients successfully manage these risks by insisting on a high level of service and implementing specific risk mitigation efforts. These include organizing their firm into studios where groups have a long history of working together. They strongly recommend qualified contractors and do much planning with pre-construction teams. They bring in independent consultants to do peer reviews of accessibility and waterproofing issues. They work with experienced attorneys to review all of their contracts. They ask that a “no condo conversion” clause be written into the deed. They have people who are dedicated to the construction administration phase and who are expert at it. They have a Chief Learning Officer and hold Project Manager meetings to share lessons learned. They require subconsultants to submit documents a week in advance of the due date

to allow time for review. They require owners to hire a construction inspector to verify that the work is being done in conformance with the plans. They offload specific consultants, including the geotechnical consultant. And in addition, they are willing to say no. They have a documented history of saying no to bad deals. Together, these measures have allowed them to thrive in a challenging market.

Schools have been a leading source of claims frequency due to several factors. One is inexperienced boards with champagne tastes and beer budgets. Second is low-bid contractors. And third is the unlevel playing field, where our potential jurors know that they, as taxpayers, will have to pay if the defendant does not.

There is frequently a powerful correlation between project size and claim severity. Larger projects are understandably more complex. While $5 million change orders on a billion-dollar project might be less than 1% of construction values, it’s still a lot of money. And so we see more significant lawsuits filed on larger projects generally - which firms can address by offering a higher level of service.

One carrier that responded to our survey noted, “The loss ratio for our large firm segment is running more than two times higher than the loss ratio on the rest of our book. We know we are not the only carrier seeing these claims. I believe the large firm segment will be a tough market for the professional liability markets to manage shortly.” This begs the question: what is a large firm? It’s a subjective question, but the authors consider any firm with over $20 million in annual billings to be a large firm.

Design Build projects have also become a concern for A&E PL carriers. Contractors have been giving GMPs based on 15% to 30% design and have increasingly leaned on the designer’s E&O coverage to cover contingencies. This phenomenon has been acutely apparent

in infrastructure design-build versus vertical design-build. As a result, we have seen some A&E PL carriers refuse to accept infrastructure design-build risks.

In a related trend, a couple of major players left the project-specific insurance marketplace. As a result, some of those risks that may have been covered in the past by a project-specific policy are now being picked up under the practice policies of the designers. Many large firms have ceased offering GMP pricing on schematic design documents and are leading a drive for a progressive approach to design-build, which requires at least 60% design before a GMP can be contemplated.

Staffing

Staffing is a real issue in the design community. There’s currently more work than can be completed by available engineers at the senior, mid-level, and new graduate levels. As a result, we are seeing a rise in technical claims. This is a real challenge for design professionals, the contractor community, and their clients. Successful firms are increasing their pipeline of applicants by reaching out to students with internships and engaging them before graduation, and adopting strategies that make them the preferred place to work for existing staff.

Coverage Issues

Another question we asked our carriers is, “What are the leading contractual risk design professionals assume that give rise to a reservation of rights or denial of a claim?” The coverage offered by the major providers of professional liability insurance is so broad that a design firm should be able to transfer virtually all of its professional liability risk to insurance. However, firms can jeopardize that coverage by failing to report claims timely or by assuming liability above what is required by the professional standard of care.

There was the one question all carriers in our survey agreed on: the number one reason coverage might be declined is failure to report a claim timely per the policy. We rarely see coverage declined, but when it is, it is almost always due to failure to comply with the reporting requirements of claims-made coverage. Claims need to be reported promptly during the policy year when the claim is first made or when the firm becomes aware of circumstances reasonably likely to result in a claim.

The one thing that keeps insurance brokers up at night is the possibility that our clients might fail to timely report a claim. If you don’t report the concern in accordance with policy requirements, you could void your coverage. In the case of professional liability, our advice is to report anything that qualifies as a claim (defined as a demand for money or services) - even if you think it will resolve under your deductible or if you think you can manage it and it will go away.

You also don’t want to settle a matter without the carrier’s consent. Your insurance policy is a contract, and there are two sides to that contract. Both the insured and the insurance company have specific responsibilities and obligations under the provisions of that policy. One obligation of the insured is not to admit or assume any liability without the prior consent of the carrier.

The Insurance Marketplace Response

Over the last few years, the increased cost of claims resulting from the trends we discussed has caused property-casualty carriers to increase

rates and adjust their risk appetite. Due to an abundance of capacity, insurance rates have generally fallen since 2008. And while the architects’ and engineers’ marketplace has been relatively stable, we have seen some recent changes that we need to monitor.

Recently, we’ve seen an increase in some rates. We’ve seen some carriers exit the marketplace, and we’ve seen some carriers reduce their appetite, not wanting to write larger firms, for example, or show decreased interest in firms with specific exposures such as bodily injury, design-build, and residential risks. And we are seeing more professional liability carriers reduce their available limits. Where carriers might have offered $10 million policy limits in the recent past, some of these carriers are now only offering up to a $5 million limit. As a result, larger firms that need higher limits to meet contractual requirements or protect the financial core of their business will have to go to excess markets, which will come with an additional expense.

In addition, we are seeing real pressures to increase self-insured retentions and deductibles. A&E PL carriers would like to see firms retain 1% of their annual revenue as their deductible. In reality, most firms retain 0.5% or less. But we see some pressure on this marketplace for firms to raise their retentions.

There are many companies out there that might want to sell you insurance, but you need to be careful when it comes to picking your long-term insurance partners. There are a finite number of tier-one carriers in the A&E marketplace. Tier-one carriers are those carriers with a history of backing their promises, have a strong balance sheet with an AM Best Rating of A- or better, broad policy forms, as well as good people that are backing those promises with underwriters and claim representatives exclusively dedicated to the A&E marketplace. WTW A&E has direct relationships with the major providers of professional liability insurance who provided the data summarized in this survey.■

Dan Buelow is Managing Director of Willis A&E. He leads a team of insurance and risk management experts that are exclusively dedicated to providing insurance and risk management solutions to Architects and Engineers. Dan can be reached at (Dan.Buelow@wtwco.com) or 312-288-7189.

Mark Blankenship is the Director of Risk Management for Willis A&E. He draws on his 25 years of experience in professional liability claims, underwriting, and brokerage to provide risk management guidance, claims advocacy, and contract review support. Mark can be reached at (Mark.Blankenship@wtwco.com) or 312-525-2281.

Listening to the Structure

A new mass timber concert hall takes shape in massachusetts

By David J. Odeh, Carl Giegold, and Alan Joslin

The bucolic farmlands of eastern Massachusetts play host to a new music education and performance facility known as Groton Hill Music Center. Its owners, a not-for-profit organization founded in 1985, describe their new home as a “126,000-square-foot love letter to sound,” with studio classrooms for students of all ages, an orchestral rehearsal space, a 300-seat recital hall for soloists and small ensembles, and a grand

1000-seat concert hall that opens to view for a 500-seat lawn audience that is set in the surrounding outdoor fields in during supportive weather. From the early design stages, the owner set a goal to achieve worldclass natural and amplified acoustics in the building. Normally, engineers accomplish this by completely covering the structure to isolate it from the acoustic volume of the performance spaces. However,

Figure 1 Recital hall, showing the interplay of convex and concave wall shapes with timber decking attached to “tuning fork” columns.

Photo courtesy of Robert Benson Photography.

what made Groton Hill Music Center so unique was the architect’s unifying vision, expressed in early concept design sketches, to utilize exposed mass timber to integrate acoustics, structure, and aesthetic finishes with the building’s rural setting. Indeed, concertgoers who experience performances in the building, designed using a hybrid of steel, concrete, and timber, are literally “listening to the structure.”

An Integrated Structural System

Much of the landscape and vernacular architecture of Groton are characterized by open fields, cultivated orchards, barns, and other farm buildings. Seeking to harmoniously blend the new music center into its surroundings, Epstein Joslin Architects envisioned an abstraction of “barns and orchards” in the form of timber structural frames. The major performing spaces, or “barns,” are interconnected by tree-like “orchards” of columns, with a series of curvilinear forms giving shape to the roof structures, as shown in Figure 2. The architect chose mass timber for many reasons, including material warmth and ambiance; resonance with regional historic building methods; a desire to use renewable and sustainable products; and its ability to provide structure, spatial definition, and finishes as one system.

Through an integrated collaboration between architects, structural engineers, acoustic consultants, timber fabricators, and builders, these abstract forms were carefully engineered to achieve the project’s acoustic, aesthetic, and functional objectives.

The design team conceived of a superstructure comprising curved and straight-line generated shapes that could be supported by hybrid steel and glued-laminated timber

framing depending on the spans and structural demands of the members. Despite the seeming complexity of the architectural volumes, the structure consists of largely repeatable shapes:

• “Tree” columns consist of two or four straight-sloped gluedlaminated members connected at a common base plate on concrete piers – the “trunks.” (see Figure 3)

• “Tuning fork” columns consist of two curved glued-laminated members placed back to back and used in two different orientations (flared at the top or flared at the bottom). (see Figures 1, 4, 5, and 6)

• Utilizing tongue and groove decking, vertically curved walls were created along the height of the tuning fork columns. The reverse curvature of the column forms allowed convex surfaces on either the upper or lower walls, as needed acoustically. (see Figures 1, 3, 4, and 6)

Figure 3 Tree columns in the “orchard,” showing sinusoidal curved roof plane. Photo courtesy of Robert Benson Photography.

Figure 2 3D Model of Building Framing.

• Roofs in the “orchards” consist of curved glued-laminated beams supporting tongue and groove decking. Curvature repeats to form a sinusoidal wave along the length of the “orchard.” (see Figure 3)

• Roofs over the concert hall and recital hall utilize unique hybrid timber/steel trusses with exposed curved glued-laminated bottom chords and gable-shaped steel top chords. Ceilings in these spaces were constructed using tongue-and-groove timber decking, connecting the truss bottom chords to create convex curved volumes.

(see Figures 1, 4 and 5)

The team selected southern yellow pine for the timber, for both its structural properties and aesthetic qualities and worked with fabricator

Unalam to develop constructible forms to guide the design. Odeh Engineers designed the entire superstructure and also designed all of the unique connections for these elements in collaboration with Unalam and Epstein Joslin.

To achieve lateral stability and stiffness, a system of cast-in-place concrete shear walls and reinforced shotcrete shear walls that could be spray-applied to the curved shapes of the walls were designed. The decking served as an economical stay in place form, eliminating waste and reducing the construction schedule. As it turned out, these concrete elements, while critical to the buildings’ lateral force-resisting system, would also play an important role in achieving the acoustic goals for the performing spaces.

Acoustic Design: Making the Structure into an Instrument

For several centuries, Western music spaces have been structured massively, originally with masonry-bearing walls that are necessarily quite thick because of their height. Concert halls so constructed often sustain sound for two seconds or longer after the source has stopped (the final chord of a symphony, for example), and generations of composers have worked with this phenomenon in their minds’ ears – the music just does not sound right without all that mass around it.

But the structure in these halls was not working alone. Particularly in the 18th and 19th centuries, concert halls were the province of royalty or at least the wealthy, and they were finished lavishly with ornament inspired by ancient Greece and Rome. The acoustic diffusion offered by the statuary and filigree also lived in the ears of the

Figure 4 Concert hall space showing “tuning fork” columns and bottom chords of hybrid trusses and timber decking convex ceiling. Photo courtesy of Robert Benson Photography.

Figure 5 Concert Hall cross section showing key structural elements, including shotcrete placed on ceiling decking and walls for lateral stability and acoustic mass.

composers, taming the harshness of high woodwinds and brass and creating immersion and envelopment by scattering (not absorbing) reflections from the farthest reaches of a room that might otherwise be heard as echoes.

But concert halls of the present age are public realms; ornament for ornament’s sake seems vanquished for the long haul; rooms must be equally capable of presenting amplified and orchestral sound; and

construction culture is newly conscious of embodied carbon, all of which influence how we think about the design of spaces for music.

To address this present day paradigm, Mass timber offered an interesting approach to explore.

The acoustical analysis is compartmented into large; and small-scale analyses – there is not yet an analysis platform that allows for both room-scale geometry and surface-scale shaping (statuary and filigree, for example) in a single model. ODEON acoustic modeling software is used for room shaping studies, while Finite Difference Time Domain analyses written in MATLAB are used to evaluate diffusion, such as that applied to the operable wall at the rear of the Concert Hall. The architect favored a form more embracing and intimate than the traditional rectangular “shoebox” concert hall shape, and this was enabled by curvature in plan and curved glued-laminated timber members.

The “embrace” idea is treacherous because it relies on acoustically focusing concave surfaces, so careful study of overall room geometry was critical in both halls. They avoid the pitfalls of acoustic focusing by splaying the side walls enough to push the focus well behind the audience while counteracting the concavity in the plan with convex shaping in section facilitated by the curved columns. The plan and section were studied iteratively in the acoustic model and verified with the design team and client through auralization in Threshold’s simulation studio with its 22-channel audio system. The studio session led to a real-time change to the design — a 3-foot stretch in the height of the model to increase reverberation, made while the architect and client stepped away for lunch — a demonstration of both the design and the nimbleness of the design tool.

The columns and bottom chords of the trusses are the only curved elements of the structure, and all columns are identical in each of the two halls. The columns are arranged on a curve in plan, another move made possible by the unusual 8-foot structural bay and the 2¼-inch tongue-and-groove planks spanning between them to create the voluptuous structural shells that require no applied acoustical treatments.

These forms, which are simple convex shapes in each structural bay and step between the inside and outside legs of the double columns in the recital hall, are large enough to be diffusive even at the lowest audible frequencies. The freestanding columns and pilasters “see”

Figure 7 Musikvereinssaal, Vienna – ornament, ornament, ornament, plus lots of mass – an example of traditional concert hall design. Photo courtesy of Andreas Praefcke, CC BY 3.0.

Figure 6 Shotcrete detail at Tuning Fork Column (typical)

and diffuse mid-frequency sound, and architectural elements such as balcony fronts, rusticated stone in the Concert Hall, and chamfered edges of the tongue-and-groove planks themselves add some critical high-frequency diffusion. Where the twin legs of the columns join at the top and meet the clerestory windows, the deep recesses they create serve the same acoustic purpose as the ornaments of centuries past. Even the bolted connections add their voices to the chorus.

Given the role of mass in the history of orchestral acoustics, the use of “mass” timber represents a significant and acoustically risky reduction in mass and the volumes’ ability to provide an acoustically warm environment. The structural design mitigates that risk and perhaps turns it into an advantage in two ways. First, the stout laminated curved columns are inherently stiff, and the 8-foot structural bay limits the span of the planks to less than 7 feet, so the superstructure on its own is exceptionally stiff for a timber structure.

Secondly, eight inches of shotcrete on the walls and ceiling (which serves as the attic floor in each hall) resists in-plane lateral shear forces while providing just enough mass to carry the day acoustically — at far less than the 12 to 16 inches of concrete that are common to present-day concert hall construction. The result is a pair of rooms that seem to find a sweet spot on the acoustic spectrum spanning from solos and small ensembles through full symphony with chorus and organ and onward through jazz, folk, and Americana, and even heavily amplified rock, blues, and world music – a bass response that is gratifying to the most venerated traditional orchestral cannon while tight and controlled for the amplified genres.

Sounds of Success

Audience and performer response has been overwhelmingly positive to the venues and the building as a whole, measured in terms of skyrocketing enrollment in the music school, performer expressions of

delight in the experience on stage, burgeoning interest from booking agents, and full houses of happy audiences.

The viability of mass timber for performance spaces, at least of this scale, is demonstrated by the two halls, but there is much room to further explore the efficient use of materials, alternative forms, and how the acoustic isolation challenges of much noisier urban sites might be met. Still, in the quiet countryside of Massachusetts's Nashoba Valley live two new halls that embrace their audiences in architecture defined by a forthright wooden structure that is there to be both seen and heard.■

Project Team

Owner: Groton Hill Music

Architect: Epstein Joslin Architects, Inc.

Structural Engineer: Odeh Engineers

Acoustic/AV Consultant:Threshold Acoustics, LK Acoustics Design Studio

Timber Fabricator and Detailer: Unalam

Steel Fabricator: Superior Steel Fabricators

Concrete: Pioneer Valley Concrete

Shotcrete: South Shore Gunite

Construction: Goguen Construction, Inc.

David J. Odeh is a principal at Odeh Engineers, now a member of WSP, and was the principal in charge of all structural design for the Groton Hill Music Center.

Carl Giegold is a Partner at Threshold Acoustics and served as an acoustic/AV consultant for the Groton Hill Music Center.

Alan Joslin is a Principal at Epstein Joslin + Picardy Architects and served as the principal in charge of design for the Groton Hill Music Center.

Figure 8 View of the Sketchup model imported into ODEON and used for the geometrical analysis of the room overall.

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INNOVATION

Innovative Solutions of Crane Beams

Presentation

of an already implemented in practice cross sections of beams with a non-standard shape.

By Robert Kocur M. Eng., Agnieszka Kocur M. Eng., Julia Kocur B.

This article discusses the design of custom beams for the support of traveling cranes and is based on twenty years of experience by the author in designing steel structures and presents proven solutions related to the design of beams for traveling cranes. These crane beams are useful for cranes with lifting capacities from two to thirty tons.

The current EN 1993-6 Standard, Eurocode 3 – Design of Steel Structures – Part 6: Crane Supporting Structures, suggests checking the crane beam using two different models. The first model is a full section subject to bi-directional bending and torsion. The second model is an equivalent section subject to bi-directional bending only. This article is limited to the first model only as it is more authoritative and because it more accurately reflects the operation of the crane beam.

Each stage of the crane beam design exercises discussed herein, from manufacturing in the workshop to final assembly, was supervised by the author. This provided an opportunity to fully verify the design solutions used.

Designing the cross-section of a crane beam for cranes with a lifting capacity of 10 to 30 tons

The most common cross-sectional shapes of crane beams are shown in Figure 1. The H steel beam types (HEA/HEB profiles) are wide flange profiles, which according to American standards, are H profiles. Figure 1 shows the most common cross-section. To ensure its greater resistance to torsional buckling and bending in relation to the weaker axis Z it must have wide and thick flanges.

Creating a crane beam from such a cross-section is very simple. Namely, holes are first drilled into the beam’s bottom flanges so that the beams can connect to the column support brackets. Then, the face plates are welded to these beams resulting in finished crane beams. Unfortunately, despite their wide flanges, it is often necessary to reinforce them in the horizontal direction to control deflection and bending in the horizontal plane. Then, more developed cross-sections should be used according to Figure 2, which eventually turn out to be lighter than the most common cross-section shown in Figure 1.

Welding channels or angles improve the section’s resistance in the horizontal direction. It is difficult to construct properly because when wings (angles or channels) are welded to the beam, the welding stresses may cause deformation of the wings. To ensure that the channels or angles do not deform after welding, it is recommended

to use fins to maintain the shape of the section. This, of course, is labor intensive and increases cost. If it is acceptable to the structural engineer, fins can be welded to the web only and not to the bottom of the top flange of the beam.

An alternative to the sections with “wings,” angles with unequal legs can be welded to the bottom of the beam’s top flanges, as shown in Figure 3. This forms a double box at the compression flange, which significantly improves the section’s torsional characteristics and increases the lateral buckling resistance. Figure 3 shows the fillet welding of angle sections to I-sections. It is always possible to choose a suitable angle with unequal legs with a selected I-beam profile. The angle is connected to the I-beam by fillet welds, as shown in Figure 3. The example calculation helps to demonstrate the advantages of the above solution. The analysis was carried out for Eurocode 3 Load Combination Group 5 (crane passage), which is a group of reactions from the crane during its passage with full load when the load is in the most unfavorable position. In addition, in this situation, a lateral force is introduced, which is the result of crane bridge skews. The passage of the crane along a curved, not straight, track, see Figure 6.

Example Calculation #1

Assumptions:

•Overhead crane with a lifting capacity of 25 tons

•Single-span beam

•V = 190 kN [42.7 Kip] (maximum load from 1 crane wheel)

•HT = 24 kN [5.39 Kip] (bevel force of the crane bridge)

•a = 3.0 m [3.28 yd] (crane wheel spacing)

•L = 7.5 m [8.2 yd] (span of the freely supported beam)

•Steel: S235 (equivalent to A570 Gr. 36 in USA), rail – 40 × 50 [1.57 × 1.96 in] bar square section welded directly to the top flange of the beam.

• Internal forces: My,Ed = 635 kNm [155 Kipyd], Mz,Ed = 58 kNm [14.16 Kipyd] For determination of these internal forces, see Figure 4.

Figure 1

Figure 2

Figure 3

Five different cross-sections were analyzed:

•Section 01 - HEB 550

•Section 02 - HEA 550 + wings with L65 × 65 × 9

•Section 03 - welded plate girder [top chord Bl. 400 × 30, web Bl. 15 × 440, bottom chord Bl.300 × 30]

•Section 04 - HEA 550 + 2 × Bl.25 × 100 – (This cross-section is shown on the Figure 5)

•Section 05 - HEA 550 (+ 2 L150 × 75 × 9)

Profile dimensions:

•HEA 550 h=540 mm [21.25 in], bf= 300 mm [11,8 in], tw = 12.5 mm, tf = 24 mm

•HEB 550 h=550 mm [21.25 in], bf = 300 mm [11,8 in], tw = 15.0 mm, tf = 29 mm

Vertical and horizontal deflection criteria:

•dV < L/600 = 7500 / 600 = 12.5 mm [0,5 in]

•dH < L/600 = 7500 / 600 = 12.5 mm [0,5 in]

The calculation results are summarized in Table 1.

Conclusions – Example #1

Only Section 05 meets all the conditions. The stress in the tension flange does not exceed < fy = 235 MPa (Point 2 of the cross-section) and the load-bearing capacity with stability (see calculations below) is also not exceeded (Point 1 of the cross-section). Section 05 has a torsional moment of inertia more than seven times greater than the average torsional moment of inertia of Sections 01 to 04. This results in strains from deplaning (see Figure 7) due to torsion being approximately two times lower than the other sections. This also affects the critical moment of lateral buckling, which is more than 1.5 to 2 times greater than other sections. Therefore, the reduction in carrying capacity due to overall stability (here lateral buckling) is small.

In Row No. 4 of Table 1, a check of the so-called “wide flange effect” according to [2] is included. Only Sections 01 and 05 do not require a reduction in flange width. The analysis omits the flange reductions for Sections 02, 03 and 04, as these sections do not meet the load-bearing conditions and the load-bearing violations would be even greater.

Please note, Section 05 is one of the lightest sections making it an economical and efficient steel section choice. Section 05 is also very compact. The total width does not exceed the width of the beam flanges. As a result, the support bracket widths for the

crane beams can be shorter which will reduce the column’s bending moment.

Figures 8 through 9 show crane beam Section 05 located in the paint shop hall of the PROMUS Katowice company where there are two cranes with a lifting capacity of 5 tons each, operating on a track.

Load-bearing capacity with stability is described by the following formula

-Maximum bending moments of a beam cross-section with respect to its axis Y-Y i Z-Z

- Maximum bi-moment acting on the beam, resulting from restrained torsion

- Torsional buckling coefficient.

Figure 4

Figure 5

Figure 6

Table 1

My,Rk = W y fy - Bending capacity of the cross-section with respect to its axis Y-Y

Mz,Rk = W z fy - Bending capacity of the cross-section with respect to its axis. Z-Z

Bx,Rk = W �� fy - Bi moment capacity of the cross-section.

��M1 - Material safety factor = 1.0 or 1.1

Designing the cross-section of a crane beam for cranes with a lifting capacity of 2 to 10 tons

For lightweight crane beams, hot-rolled IPE sections are the most common base profile. According to American standards, these are W profiles. To improve their deflection and bending resistance in the horizontal direction, wings created by welded angles [similar to wide-flange I-beam sections] (see Figure 11) or

channels are used (usually the largest available from wholesalers are C300, see Figure 10.

Figure 11 shows additional stiffeners that prevent deformation of the “wings” after welding.

An alternative to the above sections is to combine the IPE profile with “blanks” cut from 250x100x8 rectangular pipe (RP) and welded to an IPE 450 I-beam (flange width 190 mm [7,48 in]), see Figure 12. The relationship is as follows: (designations as in Figure 13)

a =~ [f − (bf − tw) /2 ] / 2, a = [250 − (190 − 9) / 2 ] /2 = ~80 mm [3,15 in], hence total section width b = 190 + 2 × 80 = 350 mm [13,8 in]

Also, instead of a “blank”, it is possible to use bent sheet metal (8 to 12 mm thick [0,31-0,47 in]). (See Figure 13) In order to optimize the reinforcement effect in the horizontal direction, it is recommended that the following relationship is fulfilled: Bmin > 1.3 × A.

An example calculation was prepared to illustrate the advantages of the above cross-section. The analysis was carried out for Eurocode 3 Load Combination Group 5 [crane passage] as in Example #1.

Assumptions:

•Overhead crane with a lifting capacity of 8 tons

•Single-span beam.

•V = 75 kN [16.8 Kip](maximum pressure of 1 crane wheel)

•HT = 15 kN [3.35 Kip](bevel force of the crane bridge)

•a= 3.0 m [3.28 yd] (crane wheel spacing) and L = 7.5 m [8.2 yd] (span of the freely supported beam

•Steel: S235 (equivalent to A570 Gr. 36 in USA), rail – 40 × 50 [1.57 × 1.96 in] bar square section welded directly to the beam top flange.

•Internal forces: My,Ed = 257 kNm [62.7 Kipyd], Mz,Ed = 36 kNm [8.8 Kipyd] based on [1].

Figure 8

Figure 9

Figure 7

Figure 10

Figure 11 Figure 12

Four different cross-sections were analyzed:

•Section 06 – IPE 450 + U300

•Section 07- IPE 450 + wings with L75 × 75 × 10

•Section 08 - HEA 400

•Section 09 - IPE 450 + RP 250 × 100 × 10

Profiles dimensions:

•HEA 400 h=390 mm [15.35 in], bf = 300 mm [11,8 in], tw = 11 mm, tf = 19 mm

•IPE 450 h=450 mm [17.7 in], bf = 190 mm [7,48 in], tw = 9,4 mm, tf = 14,6mm

•C 300 h = 300 mm [11.8 in], bf = 100 mm [3,94 in], tw = 10 mm, tf = 16 mm

Vertical and horizontal deflection criteria:

•dy < L/600 = 7500 / 600 = 12.5 mm [0,5 in]

•dH < L/600 = 7500 / 600 = 12.5 mm [0,5 in]

The calculation results are summarized in Table 2.

Conclusions – Example #2

Only cross-section 09 meets all the conditions. The stresses in the tension flange are not exceeded < fy = 235 MPa (Point 2 of the cross-section) and the load-bearing capacity with stability (Point 1 of the cross-section) is also not exceeded. Section 09 has a torsional moment of inertia more than 25 times greater than the average torsional moment of inertia of Sections 06 to

08. This results in strains from deplaning (see Figure 7) due to torsion being approximately four times lower than the other sections. This also affects the critical moment of torsion, which is more than four times greater than the other sections. Therefore, the reduction in carrying capacity due to overall stability (here lateral buckling) is small.

Row 4 in Table 2 includes a check of the so-called “wide flange effect.” A 13% violation causes virtually no flange reduction.

Please note, Section 09 is one of the lightest sections making it an economical and efficient steel section choice.

Final conclusions

The I-section, reinforced in such a way that two “boxes” are formed at the compression flange, is an economical and efficient steel section that provides excellent strength and stiffness resistance. It is recommended for use at crane beams, overhead cranes and suspended cranes and hoists. In hall spaces with cranes, the spacing of the load bearing system determines the span of the crane beam. By using the proposed beam cross-sections, the spacing can be more efficient if the load bearing system can be spaced every 7 to 7.5 m instead of the standard 6 m spacing. The calculation of the cross-sectional indices (basic and sectional) was performed in Autodesk Robot Structural Analysis 2014. Critical moments of lateral buckling and cross-sectional indices were calculated in Idea StatiCa Beam 10.1■

Figure 13

Robert Kocur M. Eng., the author of this article, works in the Polish company PROMUS Katowice Sp. z o.o. (kocur@promus.katowice.pl)

Full references are included in the online version of the article at STRUCTUREmag.org

Climate Change and the Structural Engineer

How it may affect the profession. Part 3

By Gail S. Kelley

Parts 1 and 2 of this series were previously published in the May and July issues of STRUCTURE. Part 1 provided a general overview of how climate change could impact an engineer’s services; Part 2 looked at the engineer’s professional standard of care. This final article looks at contractual provisions that could help to protect engineers against allegations that they failed to address climate change in their designs. While the provisions listed below are generally appropriate for any design agreement, they can be of significant value with respect to an engineer’s liability for alleged failure to adapt its designs for climate change:

•Limitation of Liability

•Waiver of Consequential Damages

• Reliance on Owner-Provided Information or Service

•Owner’s Obligation to Analyze Possible Changes

•Owner’s Assumption of the Risk

Limitation of Liability

Limitation of liability is often a difficult negotiation (and could be the subject of an entire article by itself). If an owner will not agree to an appropriate limitation of liability for potential claims due to climate change or qualifies the limitation of liability with significant carve-outs (i.e., the limitation does not apply to claims covered by insurance or does not cover claims against the owner by third parties), the engineer should consider proposing a specific limitation on climate-related claims such as the following:

Engineer’s liability to Owner and anyone claiming through Owner for costs, losses, or damages resulting from changes in the environment or site that are not identified at the time of the design as design parameters shall be limited to the amount of Engineer’s Fee. Engineer’s liability will be based on actual damages to the extent caused by Engineer’s failure to design to existing and applicable codes.

Waiver of Consequential Damages

Consequential damages are economic damages such as lost rent; they are typically due to a delay in completion or inability to use the project as planned. Ideally, the Owner will agree to a complete waiver of consequential damages. If the Owner does not agree to a complete waiver, the engineer should propose a waiver of consequential damages arising from climate-related events:

Owner waives all consequential damages caused by climatic conditions that are not identified in the contract or existing codes and standards as needing to be addressed by the Engineer. Consequential damages include but are not limited to loss of use, income, profit, financing, business, or reputation.

Reliance on Owner-Provided Information or Services

All design agreements should contain a provision that explicitly states that the engineer is entitled to rely on information provided by the owner, with language such as that found in AIA B101, Standard Form of Agreement Between Owner and Architect:

§ 3.1.2 … The Architect shall be entitled to rely on, and shall not be responsible for, the accuracy, completeness, and timeliness of services and information furnished by the Owner and the Owner’s consultants.

This can help defend against a claim that climate change issues should have been addressed in the design. Other standard wording that is often used includes:

Engineer shall not be liable for claims arising from errors or omissions in the services and information furnished by Owner and Owner’s consultants except to the extent Engineer knew or should have known of such errors or omissions.

The engineer might want to expand on this with a contract provision that explicitly states that the owner is responsible for analyzing the potential effects of climate change.

Engineer shall be entitled to reasonably rely on, and shall not be responsible for, the accuracy or completeness of services and information furnished by Owner and Owner’s consultants. The Owner acknowledges and agrees that Engineer cannot anticipate the potential effects of climate change on the Project’s site or environment unless Owner has had these effects analyzed by Owner’s own consultant and contractually requires

that these effects be considered during the design stage of the Project. Alternatively, the engineer may require that the contract specifically disclaim its responsibility for climate adaptation:

Engineer will exercise engineering judgment consistent with generally accepted scientific, industry, and governmental information concerning environmental, atmospheric, and geotechnical conditions and developments to provide a work product that complies with applicable regulations and codes in accordance with the professional standard of care. However, Engineer expressly disclaims any obligation to design for the effects of climate change. The Owner assumes the risk of any damages or injury due to climate change, even if such damages or injury are foreseeable.

Can an Owner Stipulate that the Design Not Anticipate Risks Due to Climate Change?

Costs are a concern on virtually every project, and resistance to designing for climate change is likely to arise because of the additional costs for both investigation and design. Provided the design complies with applicable codes and current design practices, the engineer would probably not be liable to the owner if the design agreement explicitly stated that the owner does not want the engineer to address the effects of climate change and agrees to indemnify the engineer against any claims.

However, unless the owner also agrees to indemnify the engineer against third-party claims, the engineer could still face liability.

Sample language for indemnification against third-party claims is: Owner agrees to indemnify, defend, and hold harmless Engineer and its sub-consultants against any third-party claims alleging harm by Engineers or any sub-consultant’s failure to design for climatic events or conditions not addressed by existing codes and standards.

Conclusion

Engineers should review their contracts with an appropriately qualified legal professional to ensure that their contractual obligations are correctly stated. In addition, while these articles have focused on the claims and liability risks that could arise from an engineer’s failure to adapt to climate change, reputational risks and strategic risks are also a consideration. Whether or not an engineer is ultimately found liable for a claim, involvement in a project that did not meet the owner’s expectations or resulted in alleged damages to third parties (however slight) can damage the engineer’s reputation. Likewise, an engineer who cannot demonstrate a clear understanding of climate adaptation options is likely to lose out on business opportunities.■

•

Gail S. Kelley is a professional engineer, as well as a LEED AP and a licensed attorney in Massachusetts, Maryland, and DC. Ms. Kelley's work is focused on representing design professionals in both contract negotiations and claims management.

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CODES and STANDARDS

Temporary Structure Design

Management in the entertainment industry.

By Kyle Kusmer, P.E.

The development of the International Building Code 2024 for temporary structures goes a long way toward defining the criteria that engineers should use in design. It provides design loads and operational management plans that both the building official and structural engineer can agree on and acknowledges that performance-based criteria for a structure with a limited duration can be achieved safely. The proposed Code has also referenced the operational management plan and inspection criteria specified in ANSI E1.21-2020, “Temporary Structures Used for Technical Production of Outdoor Entertainment Events.” While it is the best of both worlds, it is only one piece of the puzzle.

Temporary structures and their production, installation, and design involve multiple parties and organizations to put on a show. On-site personnel and engineers must work together so that operational procedures and loading on the structure have been addressed. We rely upon structural design theory and communication by on-site personnel for the public’s safety around these structures.

Types of Temporary Structures

Stages take many forms and sizes, from large steel structures to lightweight aluminum box truss structures that can be required to support 30 to 50 tons of rigging equipment.

Mobile Stages

Mobile stages can offer a unique design-build type strategy that can drive right to the venue and unfold before your eyes (See Figure 1). These mobile stages are pulled behind a truck or semi and open up into stages that can be 50 feet wide and up to 40 feet tall. Hydraulics are typically used to open the stages with guy wires and locking pins installed after deployment. The chassis and structure self-weight, along with additional ballast, if necessary, are used to resist sliding and overturning loads.

Concert Stages

Temporary structures for concert venues can be erected and disassembled every 1 to 2 weeks throughout the spring, summer, and fall. These structures are assembled, raised into position, lowered, and placed on a truck to travel to the next show.

Large stage roofs are typically erected on the ground and hoisted into place using a pulley system attached to the tops of the truss columns. The motor attaches to the structure’s base, and the lift line wraps

over two pulleys at the top of the truss. It then connects to the fully assembled roof structure. The roof structure, supporting only its selfweight, is lifted into place, locked out, and ready to support the show loads (See Figure 2).

ANSI E1.21 Operation Management Plans (OMP)

An operation management plan (OMP) is a key tool required for each show. The OMP is the engineer’s way of communicating safety-related action items to the on-site personnel for environmental loading. To reduce environmental loading on the structure, operation management plans, defined by ANSI E1.21, allow for mitigating actions to be taken by the designated qualified person on site. Weather with high wind speeds exceeding 40 to 50 miles per hour (mph) does not occur frequently, but the potential for winds of this magnitude can be forecasted ahead of time. The updated IBC recognizes that these gusts can occur during a weather event, and a plan can be implemented to reduce wind loads on these structures. Therefore, it makes sense to provide performance-based criteria that allow for a reduction in design loads during the normal operation of a show. Have you ever been in a storm that generated 40 to 50-mph winds? If you have, hopefully, you were near a permanent structure. It would be impractical for an outdoor show to go on, and the OMP allows for this practical event scenario. People can be evacuated from the area, or vertical walls can be lowered to reduce wind loads on the structure.

Figure 1 Back side of mobile stage. The walls of the trailer unfold to create the roof to support rigging loads.

Environmental Factors

Engineering documents are required to include the allowable loads assumed for the structure and the environmental factors and loads the structure can support. Environmental factors, such as snow, ice, wind, and flooding, must be monitored on-site so that the design parameters of the structure are not exceeded. Weather events that can produce undesirable effects will have an associated action that the qualified person on-site must take to maintain a secure structure. In the event of extreme weather where parameters could be exceeded, the public around the structure can be evacuated. Production personnel during the show have their eyes on the weather. During the show, on-site monitoring of wind speeds is required using methods such as an on-site anemometer or real-time monitoring with a local weather agency.

For wind specifically, the minimum stability threshold defined by ANSI E1.21 is a 40-mph service level wind. But what if a storm approaches or winds are expected to exceed this threshold? The OMP can be implemented to reduce wind loads on these structures. Some practical methods to reduce wind loading can be as follows:

• Removing scrim curtains and backdrops

• Lowering rigging equipment and light-emitting diode (LED) panels

• Using bungee cords to allow for increased deflection in scrim curtains and reducing sag in members, therefore reducing axial tension from catenary elements

• Using quick-link and pull mechanisms that remove backdrops by a single pull

• Using Keeter beam systems (See Figure 3). (Keeter beams are used for large structures and consist of aluminum members with a pulley system inside that allow large wind walls to be lowered to the stage level if a wind event occurs.) In some cases, large storms generating high winds and tornadic conditions will cause the public to be evacuated from the venue to find shelter. The show and event can even be canceled for the day or weekend since public safety is paramount. While hurricanes can give more notice about an imminent wind event, they can be somewhat unpredictable weeks before landfall, leaving us without a clear answer when constructing multiple temporary structures, stages, and grandstands necessary for larger events. In rare circumstances, the structures can be built or partially built prior to the hurricane, sometimes without enough time to deconstruct the structures. In these instances, the structure may have to remain in place, but access to the site is restricted. Unfortunately, while the structure or a portion of the structure may be damaged, a properly implemented OMP means that the public and people that could have been injured are not present. Similar measures are already widely used for permanent structures. Weather personnel and media will inform the public about possible high wind events and encourage people to take shelter and find a safer location within a permanent structure. The methods and actions in temporary events are no different, just at a lower threshold. Also, due to temporary

Figure 3 Typical keeter beam system that slides the wind walls up and down on a rail to lower panels during high wind events.

Figure 2 Roof erection of temporary stage with chain motors on top of columns to lift structure into place.

structures being, well, temporary, the probability that it will experience a significant wind event over a shorter period of time is also lower.

Inspections

Inspections must occur frequently, and records must be maintained to confirm that no component has been damaged during transit or construction. The ANSI E1.21 standard includes a load reduction factor of 0.85 to the strength of re-used systems to account for minor damage that may occur, specifically in aluminum structures. Still, it can be applied to other materials depending on their use. The combination of inspections before each show and this reduction factor contributes to the reliability of these structures. ANSI E1.21 includes information on inspection requirements and repairs or removal of damaged products. Inspections are also required after an extreme weather event to ensure that the structure has not been adversely affected before guests can enter that area.

Communication Is Key

Temporary structures, by their very nature, are dynamic. The structures can change in shape and size from show to show, and the rigging and loads can even vary depending on the changes the artists want to make. Structural engineers collaborate with many parties throughout the design process. In typical permanent building design, structural engineers work with civil engineers, architects, mechanical engineers, and electrical engineers. The collaborative team looks a little different for the entertainment industry and temporary structures. The artist, set designer, riggers, production company, stage managers, and structural engineers must communicate in the weeks prior to a show. Gravity loads from LED walls, lighting, and scenery change from venue to venue,

and the structural engineer must have the most up-to-date information to analyze the structure. In some cases, the stage structure could be overstressed, and the structural engineer will have options to consider. Can those heavy LED walls move upstage or downstage to alleviate stresses on trusses? Will some of the lighting or rigging components need to be removed completely? You may call for an additional truss element above the roof to stiffen or increase the strength of a particular member. It is up to the structural engineer to communicate the structure’s capabilities to the production team. Once the documents are complete, the rigging points from the roof can be laid out for the stage by the rigging crew, and the support motors can be dropped into the location from the roof structure above (See Figure 4).

Aluminum Roof Structures

Aluminum shapes and trusses are used extensively in the entertainment industry. The balance of strength and lightweight material lowers shipping costs and reduces the weight required to move the element for workers. However, the lightweight nature of these systems also requires more ballast or weight at the base of the structure to resist lateral loads. Aluminum box trusses and structures, sometimes 40 feet x 40 feet x 30 feet tall, can be used as an erector set to support rigging loads. Sticks of trusses, typically 10 feet in length, are bolted together to form large roof systems supporting the show's rigging. These unique structures have challenges that the structural engineer should consider. In some cases, engineering judgment needs to be used to determine structural capacity when code criteria are not specific. Box trusses, ranging in depth from 8 to 48 inches, typically look like a conventional truss when looking at the side of the member – they have two chord members with vertical and diagonal web members. However, when looking from the top and bottom, most trusses act as Vierendeel trusses for lateral loads such as wind –and distribute these loads in a much different way.

In outdoor environments exposed to wind, one must consider the unique truss action in the lateral direction. Truss manufacturers of these truss systems have charts for vertical loading only, but lateral loading is typically excluded. Bending moments at the intersection of the chords and horizontal members due to the Vierendeel truss action must be considered during outdoor use (See Figure 5). This can become a challenge as the chord members are typically two-inch diameter members with a small section modulus, and the weld-affected zones (areas in the aluminum members with reduced capacity due to the welding process) will need to be considered. As a result, bending moment design capacity ratios in the chords can be significant when compared to the tension and compression capacity of the members.

Typical truss action and failure modes, such as round tube to round tube connections and gapped "K" connections in the vertical direction of the truss, do not exist in current aluminum codes but may be present in upcoming additions. One would have to assume that tube-to-tube connections could be a failure mode in these systems. A conservative approach, using engineering judgment, maybe to use the American Institute of Steel Construction's (AISC) equations for these failure modes, but should weld-affected properties be used to determine strength? Are these failure modes likely to happen using the weld-affected regions and lower allowable stress in the analysis? If so, the possibility of these failure modes could control designs since weldaffected regions require a reduction in the strength of the member.

Engineering judgment must be used when evaluating the size and span of these systems. Box trusses, unbraced for 50-plus foot lengths, have the potential for lateral torsional buckling. The stability of these

Figure 4 Photo showing rigging points located on the stage in chalk while motors are dropped into their exact position from the trusses above.

trusses depends on the horizontal truss action described above and is a consideration in longer spans.

Structural design in the entertainment industry is an ever-changing, dynamic process that involves effective communication and coordination. Current and new standards have begun to close the gap between the design and operation of large temporary events. Next time you are at a show, look up, and you will see a lot more than just lights and sound.■

Kyle Kusmer, P.E., is a principal at Schaefer in Cincinnati, Ohio. He’s a licensed structural engineer with nearly two decades of experience designing entertainment projects across the country. Kyle is a member of ESTA participating on the Rigging Working Group and a member of the E1.21 – Temporary Structures Used for Technical Production of Outdoor Entertainment Events.

Figure 5 Aluminum truss system. Note vertical trusses and Vierendeel truss action on the top and sides of the members.

Old Bones - New Purpose

The repurposing of One Milk Street, Boston.

By Michael J. Richard, Ph.D., P.E., Paul E. Kassabian, P.E., P. Eng., CEng

Located in Boston’s Historic Newspaper Row District and at the location of Benjamin Franklin’s birth, One Milk Street is composed of three connected buildings: the Boston Transcript Building (circa 1873), the Boston Post Building (circa 1874), and a reinforced concrete connector building from the 1930s (Figures 1 and 2). These three buildings, as with many mass masonry and timber buildings of this age, have had numerous modifications over their 100+ years. The current rehabilitation project encompassed a full building renovation with structural alterations to create mixeduse upper-floor offices and retail space on the ground floor. As Engineer of Record (EOR) for the renovations, Simpson Gumpertz & Heger (SGH) performed a broad range of scope including, but not limited to, evaluation of the existing structural components of the building to determine the level of code-triggered structural upgrades; development of a procedure for in-situ proof load testing of the existing heavy timber floors; and design of new structural components such as new elevator cores, egress stairs, a two-story mechanical penthouse, a new entrance canopy, and new concrete sidewalk vaults. This multifaceted project highlights the challenges inherent in structurally retrofitting 19th-century buildings.

Project Overview

While medium-scale in square footage, the project saw large-scale challenges both from existing conditions as well as scope changes during design and construction. SGH first visited the building in June 2015 and was involved in the final tenant fit-out in Fall 2020; throughout SGH as EOR provided project continuity for building owner Midwood Investment and Development (Midwood) as the project team saw substitutions for both the project architect firm and general contractor. The long timeline highlights the challenges and design iterations encountered as the team progressed from the feasibility stage to the immediate repairs and floor upgrades phase, followed by the base building retrofit and final tenant fit-out.

In 2015, the project team envisioned a building retrofit focused on mechanical system upgrades with limited structural modifications. SGH’s preliminary scope included the following:

• Structural International Existing Building Code (IEBC) Feasibility Study

• Building Façade and Roofing Visual Survey

• New 6-Story Elevator Shaft

Figure 1 Historical (1960s) and Post-Renovation Images of One Milk St., Boston.

• New Multi-Story Rooftop Penthouse

• New Rooftop Mechanical Dunnage

• New 2-Story Basement Elevator

• New Retail Stairs

• New Entrance Canopy

• Miscellaneous Mechanical Penetrations

• New Retail Entrance Ramp

By the end of construction, SGH’s scope had expanded to also include:

• Structural Replacement of Sidewalk Vaults

• Removal of the Existing Basement Mezzanine Floor

• Structural Evaluation of Existing Floor Structures

• Development of an On-Site Adhesive Anchor Testing Program

• Visual Survey, Timber Grading, and Strengthening of the Existing Roof Structure

• In-Situ Load Testing of Existing Timber Floors

• Removal of an Existing Basement Column

• Visual Survey of Existing Mass Masonry Elevator Shafts and Bearing Walls

• Extensive Engineering Support for owner, contractor, and design team throughout construction

This article focuses on four of the major scope items associated with the project: the structural feasibility study, the removal of an existing basement column, the replacement of existing deteriorated sidewalk vaults, and the in-situ load testing of existing timber floors.

Feasibility Study

SGH was brought onto the project by Architect #1 to conduct a feasibility study of the renovation. Given owner budget constraints, Architect #1 sought to renovate the building within the limitations of the existing building structure (i.e., limit the extent of structural modifications, focus on updating architectural finishes, improve stair and elevator access, and upgrade mechanical systems). SGH’s 2015 feasibility report stressed that

structural modifications should be limited in scope to avoid triggering a building-wide structural seismic evaluation for current building code loads required by the IEBC for substantial structural modifications encompassing more than 30% of the building structure. Given the age of the building, SGH noted that major seismic upgrades would be required if the design team triggered the IEBC requirement.

After a 6-month pause, SGH was contacted by Architect #2, who took over the project and proposed an expanded scope to meet the owner’s needs. In feasibility study #2, SGH used visual plan markups to illustrate how proposed scope changes triggered substantial structural modifications and code-required seismic upgrades. These markups enabled the project team to refine a project scope that met the owner’s expectations without triggering substantial seismic upgrades.

Column Removal

The owner desired a new 2-story basement elevator to connect the double-height basement space (after the removal of an existing basement

Figure 2 Typical plan layout of One Milk St.

Figure 3 Existing Building Column and Proposed Elevator.

mezzanine level) with the ground floor retail space of the Transcript Building (Figure 3). Given the tight existing building layout and required egress paths on both floors, the architect could only place the new elevator at an existing building column location. This column supported all building levels. SGH was tasked with developing a solution that allowed for the elevator shaft to be installed while also supporting the existing building above.

SGH designed the reinforced CMU elevator shaft as the new “column” of the building. Steel transfer beams at the top of the shaft (Figure 3) transfer load from the existing steel column above to the shaft walls. A new reinforced concrete mat foundation/pit supports the full structure above.

However, two challenges arose from this approach: the transfer of existing load from the existing column to the new shaft and the design of the new mat foundation. Both issues highlight the need for collaboration between the EOR’s final design and the General Contractor (GC)’s construction means and methods.

First, SGH specified the GC to submit a shoring and jacking sequence for the demolition of the existing column and construction of a new shaft. While SGH had designed the shaft to support the building loads in the final design condition, the transfer of loads from the old to the new structure needed to be an engineered sequence designed by the contractor. The GC’s scaffolding subcontractor initially proposed two scaffolding towers be constructed on either side of the proposed elevator, with transfer beams supporting the existing column above at Level 2. Under this sequencing, the GC would construct the temporary scaffolding towers with transfer beams, connect the transfer beams to the existing column, cut and remove the existing column and foundation, and then construct the new foundation and CMU shaft up tight to the steel transfer beams.

SGH stressed that the contractor’s sequencing did not provide a controlled transfer of loads from the existing column into the transfer beams. Since building forces already existed within the column, the GC needed to jack the load into the transfer beams prior to cutting the existing column. Simply cutting the column without jacking would risk an uncontrolled, unsafe transfer of load between the column and the beams. Furthermore, the proximity of scaffolding towers to the new pit excavation would undermine the scaffolding.

Given the constrained floor plan around the proposed elevator, SGH worked with the GC and the shoring contractor to develop a solution to switch out the existing column for the new masonry shaft while maintaining the building’s structural load path. The team’s solution was to use the shaft itself as the column shoring prior to the column’s removal. Specifically, the sequence entailed:

• Excavating the proposed elevator pit around the existing column footing

• Doweling the new elevator foundation to the existing column footing to allow the new CMU shaft to be erected with the existing column in place

• Attaching steel transfer beams to the existing steel column at Level 2

• Erecting the masonry shaft

• Erecting scaffolding within the shaft to allow for the load jacking sequence

• Jacking the existing building load out of the column and into the transfer beams

• Shimming the top of the CMU walls tight to the transfer beams

• Releasing the jacks to transfer the load into the CMU shaft

• Demolishing the existing steel column from within the elevator shaft

Sidewalk Vaults

Basement vaults below city sidewalks are common in older cities like Boston. However, poorly maintained vault structures can pose a risk to pedestrians if no condition assessment is performed and water ingress is a factor.

One Milk proved to be one such case study. SGH’s initial survey found sidewalk structural conditions ranging from original granite stone slabs supported on heavily corroded steel to brick masonry arches, previous concrete on metal deck repairs, and exterior stone foundation walls (Figure 4). The city historical commission required that the original granite slabs remain in place, while the city building official noted that the sidewalks must support an equivalent H20 truck wheel load (16 kips per AASHTO).

The team initially sought to repair and rehabilitate the sidewalk in place with retrofit work from the basement vault below. The contractor would install new steel framing to support existing granite slabs while also repairing and sealing cracks. This scheme sought to minimize disruption at the street level, manage construction costs, and retain the historic character of the original sidewalk.

However, this scheme could only mitigate rather than prevent future water ingress and would require ongoing maintenance. In addition, it would require significant structural steel for new loads and still leave the owner with a patchwork sidewalk. The GC separately advised repairing the sidewalk at street level, as getting new steel into the tight basement vaults would be difficult and costly.

SGH and the team subsequently developed a topside sidewalk replacement where existing granite slabs were removed, and a new reinforced concrete structural slab was cast to span between existing

Figure 4 Existing deteriorated sidewalk vault conditions.

stone foundations at the street and new steel framing at the building face. The architect used the new concrete slab as a clean substrate for a waterproofing layer. To meet the historical commission requirements, ,which focused only on maintaining the original granite visual, the original large granite slabs were shipped off-site, cut, and reinstalled as a wide curb on top of the new concrete slab, while a layer of high-strength rigid foam and a non-structural concrete topping was installed to provide a top walking surface.

In-Situ Floor Testing

The building’s ground floor level was classified for existing retail (one hundred psf occupancy load) with a mixture of structural systems (concrete on a metal deck, corrugated metal arches supporting concrete, and brick masonry arches). The upper floors were classified for existing office space (50 psf occupancy load) and were composed primarily of heavy timber framing. The design team intended for the upper floors to remain office space. To lease the space, Midwood sought to install a cementitious topping on the floors to meet flatness and levelness requirements for a floor finish substrate. In spring 2017, Midwood had SGH structurally evaluate existing floors for an additional standard one-inch-thick cementitious floor material. However, in late 2018, an independent survey of the existing upper floors by the prospective tenant determined that the original floors were out of tolerance for flatness and levelness by as much as four inches. The tenant, therefore, required thicker floor prep material to support their proposed finishes, resulting in a higher design load. With new ceiling finishes already installed in the building, Midwood asked SGH what the team could do to meet the tenant’s loading and levelness requirements, keep the lease on track, and avoid replacing all the completed upper-level floors. Due to newly installed ceiling finishes, SGH could not survey the wood floor structure. Therefore, SGH recommended on-site proof load testing of the floors to evaluate the timber floor’s reserve capacity.

Using Section 1708 of the 2015 IBC along with ASTM E196-06 (Standard Practice for Gravity Load Testing of Floors and Low Slope Roofs), SGH developed a testing procedure for the wood-framed upper floors. A representative floor area was designated on Level 5 of both the Transcript Building and the Post Building. To avoid

damaging newly installed finishes, the team elected to use standard CMU blocks instead of typical water-filled drums for the test weights. The contractor removed ceiling finishes from the representative test areas, allowing SGH to conduct a wood microscopy analysis and in-situ visual stress grading.

The SGH-developed testing procedure entailed:

• The GC incrementally added CMU block to the floor in four increments (Figure 5)

• The SGH and surveying team measured floor deflection on the floor below for each load increment. Loading was halted for 1 hour before the addition of the next CMU increment

• The team left the fully loaded floor in place for 24 hours to monitor deflections due to creep

• After the 24-hour loading period, the contractor incrementally removed the CMU

• The team measured floor deflection rebound after each unloading stage with a 1-hour wait between increments

• The team monitored the unloaded floor for another 24-hour period to measure floor deflection rebounding

The team determined that both test areas met the IBC deflection and recovery criteria for a successful test and could support an additional load of ninety-two psf, satisfying the tenant requirements and allowing Midwood to proceed with their tenant lease agreement.

Through collaboration, the design team was able to preserve major portions of the historic character, façade, and structure of the building while also integrating modern egress paths, retail businesses, and office space. As EOR, SGH was able to provide continuity on the project team for the owner and help Midwood understand the multiple structural systems in the building. SGH, the architect, and the GC were then able to navigate the challenges of these varying existing building conditions to integrate new structural components and load-test existing floors to keep the project moving toward successful completion.■

Michael Richard, Ph.D., P.E., is a Senior Consulting Engineer with Simpson Gumpertz & Heger, Inc. (mjrichard@sgh.com)

Paul Kassabian, P.E., P. Eng., CEng, is a Principal with Simpson Gumpertz & Heger Inc. (pekassabian@sgh.com)

Figure 5 In-Situ Load Testing Sequence.

structural ADHESIVES

Enhancing Light-Frame Shear Wall Performance with Elastomeric Adhesives

A test program study.

By Bilal Alhawamdeh, Ph.D., A.M.ASCE, and Xiaoyun Shao, Ph.D., P.E., M.ASCE

Modern elastomeric adhesives can potentially transform the realm of light-frame wood (LFW) construction, offering a cost-effective solution to increase strength, stiffness, and energy dissipation under lateral loads induced by earthquakes and wind. LFW shear walls are integral to the lateral force-resisting system, providing a primary source of stiffness and strength to the structure by transferring loads to the foundation. The current model of shear walls dissipates energy through plastic deformation of the sheathingframe connections, resulting in nail yielding, nail withdrawal, and sheathing edge tearing. Investigators found that conventional adhesives, including water-based, solvent-based, and polyurethane-based (PU), can significantly improve shear walls’ strength and stiffness. However, concerns about volatile emissions, lack of durability, and

brittleness limit their application in LFW structures. On the other hand, silyl-modified polyether (SMP) are modern adhesives gaining interest in construction for their moisture-curing, isocyanate-free, UV-stable, chemically resistant, and high flexibility properties. This article demonstrates the efficiency and cost-effectiveness of SMP adhesive in improving the seismic performance of shear walls through an experimental program.

Materials and Methods

The researchers fabricated three configurations of 8 × 8 feet LFW shear wall specimens (see Figure 1), which consist of 2 × 4 nominal Douglas-Fir frames and 3-ply plywood sheathing of 4 × 8 feet and 3/8 inch thickness. The reference configuration (R) used the minimum standard of nailing specified in the Special Design Provisions for Wind and Seismic (SDPWS). The construction applied an adhesive bead thickness of approximately 0.25 inches of PU and SMP adhesives at the sheathing-frame connections to make the two adhesive configurations. Table 1 shows the mechanical properties of these two adhesives. The adhesive cost in each specimen

Table 1 Mechanical properties of the used adhesives

Figure 1 Schematic of the test specimen

was less than 21 US dollars, corresponding to a 15% increase in the total material cost. To evaluate the seismic capacity performance, the researchers tested the specimens under lateral cyclic loading simulating ordinary ground motions.

Results and Discussion

Force-displacement relationships

The researchers developed hysteresis loops and corresponding envelope curves from the cyclic loading tests (see Figure 2). The R configuration initially demonstrated linear loops but ultimately exhibited a nonlinear response due to nail deformation and hysteretic pinching. Pinching occurs when the hysteretic cycles pass closer to the horizontal axis due to nail slippage. The PU configuration showed a linear response of all the loops. The drop in the load capacity at a small displacement explains its brittleness. The SMP configuration had linear loops at the early primary cycles. Then the nonlinear loops dominated yet showed no signs of degradation at higher displacement relative to the PU configuration. The shear resistance of the continuous adhesive bonding between the sheathing and framing decreases the pinching in the hysteresis responses in both PU and SMP configurations.

Performance analysis

The performance parameters, including maximum load strength, elastic shear stiffness, and energy dissipation, show that the SMP adhesive increases the energy dissipation by 100%, strength by 150%, and stiffness by 15% compared to the R configuration (see Figure 3). The higher elongation of the SMP adhesive allowed the sheathing to translate and rotate with less constraint and resulted in a cohesion failure in the wall specimen. Comparatively, the relatively high shear strength and low elongation of conventional PU adhesive caused a premature failure in the substrate and reduced energy dissipation.

Comparison to design standards

Shear wall design shall comply with the allowable story drift (Δ) specified in the ASCE 7 standards based on the seismic risk category to maintain structural integrity and life safety. For LFW residential housing, the drift limit of an 8 feet wall is 2.5% of the wall height translating to 2.4 inches. The displacement corresponding to the first drop below 80% of the maximum load determines the ultimate displacement (Δu). The R configuration had a Δu of 3.2 inches, exceeding 2.4 inches of the SMP configurations and 1.1 inches of the PU configuration. Thus, increasing the strength and stiffness of shear walls using the SMP adhesive slightly reduced the seismic deformation, which reduced damage while still leading to the most energy dissipation (see Figure 3) among the three configurations. On the other hand, the R configuration would require heavy construction (i.e., additional wood materials of ~20 US dollars based on the current market) to achieve an equivalent strength (i.e., 12,000 lbs) of the SMP configuration.

Conclusion

This experimental program demonstrated that modern elastomeric adhesives like SMP significantly improve light-frame shear Figure 2

Force-displacement hysteresis

walls’ wind and seismic performance by providing a constant source of strength, stiffness, and energy dissipation. Using SMP adhesive can lead to cost savings and efficiently reduce the risk of damage during seismic and high wind events. The

continuous adhesive bond between wood members significantly reduces pinching in the hysteresis responses of shear walls. It also decreases drift to meet the allowable story drift criteria, thus benefiting nonstructural components by mitigating damage caused by large displacements. SMP adhesive results in twice the energy dissipation of the conventional nail shear walls. The more ductile failure observed in the SMP specimens proves its ability to change the brittle failure observed in conventional adhesive research, making SMP adhesive a cost-effective solution to enhance the seismic performance of LFW structures. The promising experimental results motivate the researchers to further quantify the building system’s performance and response parameters following the Federal Emergency Management Agency (FEMA P-695) methodology. The quantification will establish global seismic performance factors for the SMP application as a new seismic force-resisting system proposed for inclusion in model building codes. Overall, the SMP adhesive represents a promising solution for construction professionals seeking to optimize project safety and efficiency.■

Bilal Alhawamdeh, Ph.D., A.M.ASCE, is a Sr. Research Associate at the Bronco Construction Research Center, Western Michigan University (bilal.alhawamdeh@wmich.edu).

Xiaoyun Shao, Ph.D., P.E., M.ASCE, is a Structural Engineering Professor at the Department of Civil and Construction Engineering, Western Michigan University (xiaoyun.shao@wmich.edu).

Figure 3 Normalized results of the reference configuration

structural SYSTEMS

Modular Composite Core Wall System

A modified speedcore system.

By Ahmad Rahimian, Ph.D., P.E.,

The composite plate shear wall-concrete filled (C-PSW/CF) system, also referred to as the SpeedCore system, is a modular construction system where wall modules are composed of two steel faceplates, concrete infill, and tie bars connecting the face plates. While the system is somewhat similar to the system previously introduced by Corus (now TATA steel) in the early-mid 2000s and marketed in the United Kingdom (UK) as CoreFast, the SpeedCore system has significant differences from CoreFast. The key feature of the CoreFast system is that the elements are shop-fabricated using a patented friction stir welding process to connect the tie bars to the inside of the face plates, whereas SpeedCore is a nonproprietary system (Huber et al., 2021). Although CoreFast was not widely used outside of the UK, the SpeedCore system is emerging as an alternative to conventional reinforced concrete (RC) shear walls in the US. This is due to both constructability advantages for steel buildings, research undertaken by the nuclear industry, and, more recently, the research led by Purdue University and the University at Buffalo. The SpeedCore system is now addressed in the American Institute of Steel Construction’s AISC 360-22 Specification for Structural Steel Buildings, AISC 341-22 Seismic Provisions for Structural Steel Buildings.” and AISC Design Guide 38 “SpeedCore Systems for Steel Structures.” This article presents alternative approaches to connecting Speedcore modules to simplify construction.

Studies for the Rainier Square project in Seattle, WA, performed by contractors, indicated that the use of SpeedCore reduced construction time. As a possible substitute for reinforced concrete shear walls, SpeedCore walls resist gravity, lateral wind, and seismic loads. An advantage of SpeedCore is in commercial high-rise buildings where steel is generally the primary structural material, and either steel bracing or reinforced concrete shear walls around the building core are used for building stability. SpeedCore provides the lateral strength and stiffness associated with reinforced concrete shear walls. Although this system might utilize more steel material than conventional reinforced-concrete wall systems, it has the benefit of eliminating conventional formwork and steel reinforcement bars. The construction schedule is also improved since all steel construction moves ahead in unison, and only the concrete infill operation follows a few days behind. This sequence is similar to the concrete-on-composite deck construction method. With

SpeedCore, there is no traditional shear wall construction to sequence before or after steel framing installation. Such steel-centric construction methods could simplify construction coordination, scheduling, and logistics. The fabrication, construction methods, and tolerances are generally the same for SpeedCore walls and adjacent structural steel floor framing. The SpeedCore system has been used in two projects: the Rainier Square Building in Seattle, WA, and the 200 Park Ave Building in San Jose, CA. For these projects, a ductile lateral force-resisting system was required for seismic design even though seismic forces did not necessarily govern the Rainier Square building member sizes. As reported for these buildings (Morgen et al., 2018), the SpeedCore system had a competitive advantage over other systems that are commonly used in high-seismic areas. For the 200 Park Ave building, recent improvements and refinements of the system were implemented through additional research at Purdue University and the University at Buffalo. To further increase the advantage of the SpeedCore system, especially for areas of lower ductility demand, this article proposes an alternative, proprietary method of detailing the SpeedCore system that reduces the amount of field welding and simplifies the connections at the interfaces between individual wall modules (Rahimian et al., 2022, U.S. Patent 11,352,786 B2). This alternative detailing approach is based on AISC and the American Concrete Institute (ACI) standards, as applicable, with experimental verification currently planned. The proposed alternative approach is not part of the AISC Design Guide 38, “SpeedCore Systems for Steel Structures.”

Figure 1 Modularization options

S.E., F. ASCE, Konstantin Udilovich, P.E., Ilya Shleykov, Ph.D., P.E., LEED AP BD+C, and Jeffrey Smilow, P.E., F. ASCE

Module Sizing

Maximizing offsite fabrication and reducing the piece count of field-connected elements is key to achieving economy for any prefabricated system, including SpeedCore. Therefore, using the largest prefabricated wall module size that can be efficiently produced and shipped is preferable. Currently, the widest plate commonly rolled in the United States is limited to a 10-foot width, while the maximum shippable length is typically 50 feet; the shipping length and width can be increased subject to special permitting by local authorities. Domestic facilities with the capability of rolling wider plates are coming online. However, only a limited number of domestic suppliers have this capability. Whereas previous designs of the SpeedCore system used horizontally oriented modules with horizontal joints approximately at every floor (Fig 1a), the proposed alternative orients the modules long-side-vertical, extending upward for a height of two to three floors with a standard module width of approximately 10 feet, or wider when possible. The proposed arrangement (see Figure 1b) also allows increased flexibility to place SpeedCore wall modules around elevator shafts, stairs, and other building core elements. The primary intent of this modular arrangement is fewer horizontal joints, which are structurally more demanding than vertical joints. Furthermore, it offers steel contractors the option of fabricating narrower modules, giving them more control over fabrication, transportation, and erection in projects where wide modules may present challenges.

Joints

Recent seismic applications of SpeedCore featured field joints between steel plates using field welding to join adjacent wall modules. If the yield strength of the plates is fully utilized or if high levels of inelastic deformation capacity are needed, field-welded joints, including complete joint penetration (CJP) welds, are typically needed. However, even in a high-ductility system, full continuity of the faceplates should be only required in certain sections, and not all, of the structure. Using performance-based design principles, areas with large inelastic strains (e.g., plastic hinge zones) can be identified where faceplate continuity is needed. Outside such areas, simpler joining methods can be utilized. In traditional SpeedCore construction, the horizontal joints between modules can be highly stressed due to axial, bending, and shearing stresses. In contrast, vertical joints between adjacent wall submodules are subjected to mostly in-plane vertical shear forces that result from in-plane flexure of wall segments and are very different from horizontal joints. Detailing strategies that eliminate field welding should lower the overall cost of the system.

Steel module plate thickness may be governed by considerations other than strength, and therefore plate material might not be fully utilized in strength calculations of module joints. Less than full utilization of the faceplate may be a result of plate thickness being governed by crosstie spacing and practical aspects of plate thickness selection. The spacing of the ties is related to plate thickness based on two criteria: 1) the pre-composite stiffness of the wall module for constructability, and 2) plate buckling between ties in the composites state for design loads. The pre-composite stiffness of the empty modules is important for transportation, shipping, and handling. Plate local buckling between ties may occur under a full load if the tie spacing is too large and the

plate slenderness ratio is too high. The AISC Design Guide seeks to eliminate plate elastic local buckling. Plate thicknesses of less than 3/8 inches are often impractical since thinner plates may be difficult to handle in fabrication and result in modules that are too flexible for assembly, installation, and concrete placement. Faceplate thickness will typically range from 3/8 inches to 5/8 inches. Greater thickness may be used in zones of high seismic demand. It is generally more economical to have fewer ties, even if the plate thickness is increased slightly to meet stiffness and slenderness limits, since installing them is labor intensive. Consequently, in many cases, the development of the full tensile strength of a plate in the joint is not required unless the wall element is in a demand-critical location where plastic hinging is expected. Although field welding using CJP joints is a viable way to connect all wall modules, alternative ways to achieve the needed joint performance in regions of low demand should be considered. (In principle, identical to detailing transverse reinforcement in special reinforced concrete shear walls.)

Horizontal Joints

Depending on the utilization of the steel plate in strength calculations, the full strength of a faceplate at a horizontal joint may not be required. Some of the potential alternatives for the horizontal joint details are shown in Figure 2.

Figures 2b and 2c offer alternatives to CJP and PJP welds, namely, fillet welds and bolted splices. Such connections, possibly to be used where strength demands are low, require less onerous inspection. Option (a) represents the current standard practice for welded joints, and option (b) has been studied by Ramesh (2013, 2014). Research and testing are underway at the University at Buffalo on the use of bolted splices such as option (c) and other bolted alternates (Huber, 2022). These alternatives also provide greater tolerance for field adjustment since the gap between the faceplates can be larger than the maximum root opening for a PJP or CJP weld. Since access to the back side of the faceplate is not possible after the module is placed, a bolted option would most likely require a proprietary product: either a blind bolt or a nut retention system. Bolted connections are often preferred by steel contractors due to speed and easier quality control but require additional allowances for splice thickness.

Vertical Joints

Vertical joints between wall modules must have adequate strength to ensure that the adjacent modules are connected to act as a single structural element. These vertical joints will likely not need to develop the full strength of the faceplate outside plastic hinge zones. Eliminating field welding in vertical joints offers opportunities for speedier construction and significant cost

Figure 2 Horizontal joint options

reduction. These alternatives are presented by Rahimian et al. (2022, “Constructing Buildings with Modular Wall Structure,” U.S. Patent 11,352,786 B2).

The shear transfer across the vertical interface between adjacent wall modules may be achieved through concrete and headed stud anchors attached to stiffener plates welded to the faceplates, as shown in Figure 3. Similar force transfer mechanisms in composite structural elements have been used in multiple projects (Tuchman,1988,Viest,1997) and are covered by the AISC 360 specification; Chapter I. Figures 3, 4, and 5 show the proposed approaches for shear transfer through concrete. As shown in Figure 3, a fiber-reinforced concrete or ultrahigh-performance concrete

(UHPC) mix can be deposited into the cavity between the stiffener plates, potentially with no need for additional reinforcement. Normal or high-performance concrete, along with limited steel bar reinforcement, could be used, as shown in Figures 4 and 5. (The workability of UHPC would need to be evaluated depending on the module size and construction details.)

The vertical stiffener plates shown in these figures stiffen the individual wall modules, facilitating shipping and lifting without temporary bracing, and allowing empty 3-story high units to resist wind and construction loads in the pre-composite state. Adjacent modules may be held in place temporarily during erection using bolted clip angles. If these angles cannot be concealed within the wall finish construction, they can be removed after the infill concrete gains the required strength.

Shear transfer across vertical joints between adjacent wall modules can be achieved using steel bar reinforcement using the shear-friction provisions of Chapter 22 of ACI 318-19. Steel bar reinforcement placement may require a separate trade, so the benefits of reducing field welding versus involving another construction trade must be considered on a case-by-case basis. Kurt et al. and Seo et al. experimentally and analytically investigated the connection of steel-concrete composite walls to reinforced concrete members via noncontact lap splices of reinforcement bars and steel face plates. The same approach could be used for shear transfer in vertical joints. A variation of this approach may be to use stiffener plates with headed stud anchors on each side of the joint, forming a secondary vertical mini-joint cell at the joint between adjacent wall modules with reinforcement bars in the joint area. Shear-friction bars crossing the vertical plane of the joint may be positioned in the mini-joint cell after adjacent modules are held in place by temporary bolting. Shear-friction reinforcement that crosses the vertical plane between adjacent modules does not need to be placed as horizontal bars that are uniformly distributed along the wall module height. Instead, these bars can be arranged in different ways to simplify the placement of that reinforcement during module erection. The reinforcement could be strategically concentrated in zones of easier

Figure 3 Vertical joint with headed stud anchors

Figure 4 Vertical joint with clustered horizontal shear-friction reinforcement

Figure 5 Vertical joint with shear-friction reinforcement cages

access (Figure 4). Alternatively, reinforcement can be detailed to intersect the shear friction plane at multiple locations as a series of bent bars. Such bars may be pre-bent and lowered into the cavity between faceplates as a prefabricated reinforcement cage (Figure 5). Furthermore, concentrated horizontal bars may be placed at the top or bottom of the modules (Rahimian et al., 2022), further simplifying the installation of shear-friction reinforcement and speeding the joining of adjacent wall modules.

A series of numerical studies were performed for the proposed joint cell under shear loading using the VecTor2 nonlinear finite element software (Wong et al., 2002). The analysis considers steel and concrete material nonlinearity (Vecchio, F.J., Collins, M.P., 1986). This analysis was performed for monotonically increasing loads. The study simulated “pure shear” conditions in the joint for a 24-inch thick wall with 8000 psi concrete fill and ½-inch thick face plates, with a height of 28 feet. (Further analysis under reversed cyclic loading is planned in conjunction with a testing program.) Numerical studies were performed for the proposed zigzag and clustered reinforcement options shown in Figures 4 and 5; uniformly distributed horizontally was used as a benchmark. The reinforcement area was the same for all options. Although the sloped segments of zigzag bars (Fig. 5) may be arranged at any angle, only 0° and 45° were considered here. The clustered and zigzag reinforcement arrangements are much easier to install in sequence with steel module erection than the benchmark uniformly distributed reinforcement. Figure 6 shows the load-deformation behavior under monotonic loading for all four reinforcement configurations. The total area of reinforcement crossing the vertical plane of the joint cell is the same in all four cases. The study indicates that both the angle of reinforcement and the spacing of the bars in the cell affect joint shear strength. To enable a comparison, the figure also shows the ACI 318 shear friction strength using coefficients of friction of 1.0 and 1.4. Figure 6 indicates that the proposed rebar arrangements provide strength similar to that calculated using the ACI 318 shear friction model. Figure 7 shows the VecTor2 crack patterns for the

zigzag option at the peak load.

Link Beams

Link beams (also called coupling beams) that connect adjacent wall segments in the areas around door openings in tall buildings are some of the most challenging elements to design and detail. The overall system ductility, fabrication, and erection need to be

Figure 6 Joint shear monotonic load-displacement curves for various reinforcement arrangements

Figure 7 VecTor2 crack pattern at vertical joint between wall modules reinforced with zigzag bars

considered. AISC Design Guide 38 addresses the coupling beam design, recognizing composite concrete-filled and steel-only beams. Bruneau and Varma et al., 2019 studied SpeedCore concrete-filled link-beam box sections and connections between box-section beams and SpeedCore walls. That study considered coupling beam connections using continuous web plates and interrupted wall flange plates, as well as connections using lapped web plates and continuous wall flange plates and face plates from outside. In the first option, wall flange plates are interrupted and welded to the beam flange plate, where the beam flange plates penetrate the wall and are welded to the wall web plates. In the second option, beam flange plates are welded to uninterrupted wall flange plates. According to Bruneau and Varma et al., box-beam welds, used to connect the coupling beam flanges and web plates to the composite wall steel plates, are designated as demand critical and, therefore, they are required to meet the additional requirements in AWS D1.8. Furthermore, concrete placement in such composite link beams requires a mix that can flow in and fill the box section without forming cavities due to air pockets. Special holes in the beam top flange would be needed, and these holes should be accounted for in the box beam design. Although AISC Design Guide 38 discusses steel-only link beams as an alternative for non-seismic applications, steel I-beams embedded (rolled or built-up) in concrete walls are a recognized seismic lateral force-resisting system per AISC 341 (Section H4). I-shape link beams can be incorporated into the proposed SpeedCore system while meeting the intent of AISC 341 provisions. Even though AISC Design Guide 38 describes the concrete fill inside the composite walls as "plain" concrete, the AISC 360 specification and a substantial body of research recognize the ductility of the composite wall elements and the fact that steel plates act as "reinforcement" for the concrete fill in a manner similar to conventional reinforcement in cast-in-place concrete walls. In addition, the AISC Design Guide 38 references the fundamental mechanics-based model (MBM) for the in-plane shear behavior of composite walls. The shear strength provisions in the guide and in the AISC 360-22 specification are based on MBM principles. The mechanics-based model developed by Varma et al. (2014) and Seo et al. (2016) explains how the shear strength of the composite section is mobilized through a compressive stress field in the concrete and a simultaneous tensile stress field in the face plates. This behavior is characteristic of reinforced concrete membrane elements, as studied by Rahimian (2019). It is, therefore, consistent with the AISC 341

specification to utilize the beam embedment provisions for composite shear walls (Section H4) with SpeedCore walls. Openings may be provided in the vertically oriented wall module flange plates to receive the steel link beams. The beams may be preset within the wall module during shipping and slide, into position after the wall modules are erected. If the beam embedment length or the opening width is too large, a bolted splice may be provided at the mid-span of the link beam (see Figure 8).

Conclusions

The SpeedCore wall system is a new development in the construction of steel buildings. It is considered a ductile system and suitable for use in regions of high seismic hazard. The alternative construction proposed in this article is intended to reduce the cost of the SpeedCore system by reorienting the wall modules, reducing the field welding of horizontal joints, and replacing vertical welds with composite connections. The proposed joint solution could be widely used for buildings in regions of low seismic hazard with small inelastic deformation demands and in segments of buildings in regions of high seismic hazard where plastic hinging is not expected. The additional benefit of the proposed joining method is increased tolerance in vertical joints between modules due to the elimination of field welding.

The proposed modified link-beam design based on conventional steel I-shape members could help simplify the coordination, fabrication, and construction of coupling beams.

Increased construction speed through offsite prefabrication of major structural elements has the potential to minimize risks, improve quality and safety, and mitigate disruptions in urban areas. New innovative systems, such as those presented here, are key to making the composite concrete-filled wall system an economical and viable alternative to conventional reinforced concrete shear walls.■

Ahmad Rahimian, Ph.D., P.E., S.E., F.ASCE, is a Director of Building Structures at WSP USA, a global multidisciplinary engineering firm. Ahmad’s more than forty years of experience in structural engineering includes the design of numerous notable projects worldwide, ranging from super-tall commercial and residential towers to sports facilities.

Jeffrey Smilow, P.E., F.ASCE, is the Managing Director of Building Structures at WSP USA. He has over forty-five years of experience in the structural engineering of high-rise towers, with an extensive background in the design of a wide range of steel and concrete structures.

Konstantin Udilovich, P.E., is an Associate at WSP USA Building Structures division. With over twenty years of experience, Konstantin has performed advanced engineering contributing to a wide range of building projects in steel and concrete.

Ilya Shleykov, Ph.D., P.E., is a Vice President of Building Structures at WSP USA. Ilya has decades of experience in structural engineering, specializing in advanced analysis of complex and special structures and performancebased seismic design.

Figure 8 Detail of link-beam with embedded I-section

Full references and acknowledgements are included in the online version of the article at STRUCTUREmag.org.

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structural DESIGN

The Devil in the Details

Structural design considerations for rooftop equipment platforms.

By John P. Phelan, P.E. and Michael F. Hughes, P.E., S.E.

Supporting rooftop mechanical or electrical units is a simple problem with a complicated solution. While some units are small and light enough to bear on the roof with minimal additional support, many require supplemental structural roof framing. Depending on the building structural system below, the units may be supported on a pad or curbs, sleepers or loose dunnage, or raised platforms. Typically, the term dunnage has been used interchangeably with raised platforms Still, it is more accurate to say that dunnage consists of loose structural members that distribute the unit’s load over the existing roof without penetrating the roof membrane. These members are not connected to existing roof structural framing.

Raised mechanical platforms are steel-framed structures added to the roofs of new and existing buildings. These platforms provide support for the units and maintenance walkways between each unit. They are supported on posts bearing on the base building structure. These raised platforms are sometimes not considered part of the base building structural system and are often added while upgrading or updating the building mechanical equipment. This article reviews the design procedure for raised mechanical platforms.

Geometry

The design of raised mechanical platforms is frequently driven by geometric constraints of both the building below and the units themselves. It is best to install platforms away from building edges and roof drains, if possible, to avoid complications with facade attachments. However, if avoiding a roof drain is not feasible, the roof drain may need to be moved to accommodate the platform post.

Existing roofs typically do not have the reserve capacity to support large mechanical unit loads added after the original construction. Therefore it is common to support the units on new elevated equipment frames, which typically post down to the top of existing columns. Columns tend to have additional capacity compared to horizontal roof framing. The platform must be tall enough to accommodate potential reroofing and any ducts that run below it or other equipment that may be anchored directly to the roof. The National Roofing Contractors Association (NRCA) Roofing Manual for Membrane Roof Systems recommends the following clearances under the platform to allow for roofing:

Where ducts run below the edge of the platform, the kickers that provide lateral load resistance must be clear of the duct, which may necessitate a further increase in height. If there are no ducts or the duct height is minimal, the platform should be raised high enough to ensure that the bottom of the steel is above the top of the flat roof snow height so that snow will not drift against the platform structure. Starting with the 2016 edition of ASCE 7, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, Section 7.8 specifies that drift loads may be ignored only when the bottom of the platform or other projection is at least 2 feet above the balanced snow load height. The platform plan geometry and framing are primarily driven by the equipment on the platform and require close coordination between the design team and Mechanical Contractor. Framing must be provided under the perimeter of the equipment and any intermediate supports. Some equipment, such as cooling towers, generators, and air-handling units, have fixed projections below the equipment that framing must avoid, like piping, conduits, and ducts.

Walkway widths should be at least 3.5 feet for access and service. In addition, service clearances for replacing coils or filters are sometimes needed. Typically, these clearances are the width of the unit and occur on one side of the unit. Sometimes removable guards can provide the necessary clearance and limit platform size. These removable guards will require additional maintenance.

Construction/Detailing Considerations

As discussed above, it is best to align platform posts with the building’s columns. While posting off roof framing is possible, the roof framing or connections likely require reinforcement for strength or stiffness. Typically, a cap plate can be added to the top of the column, and the post is either welded or bolted to this plate. The post base connection detail and post lengths must account for the roof pitch where the roof framing slopes. In open web steel joist construction, the joists and joist girders typically bear on top of the column, complicating adding a post above the column. The authors have successfully used the detail shown in Figure 1 to bolster the column in joist construction.

To attach to the framing below, the posts must penetrate the existing roof to support the platform. Round or rectangular posts allow for better flashing of the post into the roof membrane. Wide flange posts are significantly more difficult to flash and should be avoided unless round or rectangular posts are not available or feasible due to other constraints. If wide flange posts are needed, plates can be welded between the flanges to create a rectangular base that allows flashing. In addition to waterproofing, the design team must consider the thermal bridge created when the platform posts penetrate the roof insulation. The architect, envelope consultant, and mechanical engineer should review the impacts of this bridge and determine whether a thermal break should be provided. Thermal breaks may be achieved by adding load-bearing insulation material within the insulation layer

of the roof or adding a thermal coating to the post and the supporting framing near the post. Many thermal breaks will require special detailing to transfer lateral loads into the roof structure. Because mechanical platforms are exposed to weather, the framing is commonly hot-dip galvanized. High-performance coatings are sometimes used for architectural reasons. Field welded connections (i.e., moment connections) should be avoided, since they require field touch-up of the galvanized coating with zinc rich paint. This zinc rich paint touch-up is inherently less robust and less abrasion resistant than shop-applied galvanizing and may have a shorter service life. Vent holes in hot-dip galvanized posts must be plugged after galvanizing since unplugged holes are a potential source of water intrusion into the building. All cap plates should be seal welded to avoid leakage Equipment supports are typically bolted to the supporting beams. In general, beams should have at least a 6-inch wide flange for equipment and grating support; however, certain equipment, such as spring isolators, may require a significantly wider supporting flange. When desired, the typical walking surface for a raised platform is metal bar grating. Standard grating is 19-W-4, which is welded bearing bars at 13/16 inch on-center, with cross bars at 4 inches on-center. Use fully banded panels to avoid sharp edges at the ends of bearing bars. The grating should be manufactured in a maximum of 3 foot wide increments to allow it to be removed and adjusted in the future, if necessary. Use bolted saddle clips to facilitate future removal.

The owner may prefer serrated platform grating for improved slip resistance. Serrated grating is manufactured by cutting ¼ inch notches in deeper bearing bars, so 1½ inch serrated grating has similar structural properties to 1¼ inch non-serrated grating. Depending on the project parameters, this can result in around 10% – 20% additional material for serrated grating to have the same structural performance as non-serrated grating. Due to added material cost, the decision of whether to provide serrated grating should be discussed with the owner and end-users of the platform. OSHA does not require serrated grating, and there is little definitive published information on the need for serrated grating on these platforms.

Platforms with grating require access using a ladder or stairs. Stairs are more common because they make bringing large objects onto the platform easier when servicing the units. However, to facilitate future re-roofing, stairs should not bear directly on the roof. Instead, they should be hung from the platform with a gap (typically 3 inches) between

the stair stringer and the top of the roofing. Figure 2 is an example of a hung stair.

Guards are required for most platforms with a walking surface 30 inches above the adjacent surface.

Some platforms and equipment require visual and/or acoustic screening. It is common for these screens to be extended 10 feet or more above the platform, which can induce significant wind load on the platform. For visual screening, the use of perforated screens may slightly reduce the added wind load; however, the Authors recommend designing the platform assuming a solid screen in the event the owner changes the screen in the future. Visual screens do not need to extend below the platform framing since the platform framing is often out of the sight lines from ground level. Set the bottom of the screen elevation to avoid inducing snow drift on the roof framing. Visual screens typically weigh approximately 3 – 5 psf, but this value should be confirmed during design.

Acoustic screens typically extend close to the roofing to mitigate sound projecting from equipment; however, this lower extension will create a snow drift against the screen that must be considered when evaluating the base building structure. Acoustic screens typically weigh approximately 10 – 12 psf to improve their sound impedance.

Bracing for these screens requires close coordination with the project team. Installing braces (sometimes called kickers) back into the platform is an option; however, they may clash with equipment and walkway service clearance. Using vertically cantilevered posts is possible, but this method increases the platform weight. Typically, wide flange posts are needed since round and rectangular posts usually cannot accommodate such large cantilever heights.

Visual and acoustic screening material typically lacks published in-plane strength, making it difficult assume it behaves as a vertical diaphragm for in-plane loads on the screen wall. Therefore it is necessary to provide vertical bracing within the plane of screen walls. Unfortunately, this is sometimes an overlooked detail that significantly affects the in-plane stiffness of the screen.

Figure 1 Bolstered connection detail for joist construction

Figure 2 Hung stairs to a mechanical platform

Special Loading Considerations

Wind loading on equipment platforms can be significant. ASCE 7-16 requires platforms to be evaluated for an amplified wind load due to localized pressures, as calculated in Section 29.4.1. Wind loads act on equipment and the webs of the framing members, causing weak-axis bending of those members.

There are currently no explicit provisions in ASCE 7 to account for shielding of the units from wind load. Neglecting shielding and assuming a full wind load on each unit simultaneously may be overly conservative. Based on research by R.E. Whitbread on the influence of shielding for wind pressures on arrays of lattice frames, similar shielding can be extrapolated for rooftop units, which gives more practical overall wind loads. From Whitbread’s research, generally, if units are grouped so that the space between units is less than or equal to the height of the units, the leading unit can shield up to 50% of the total wind load on the trailing unit. It is important to consider that equipment could be removed or replaced in the future, which would change the loading pattern. Therefore, where shielding is considered for the overall platform design, it is necessary to locally evaluate framing supporting equipment ignoring shielding.

While grated platforms are not solid surfaces, snow can bridge over the gaps in the grating, and snow loading should be assumed to build up on the grating. Figure 3 shows an example of snow bridging gaps on a grated platform. ASCE 7-16 Section 7.13 requires grating to be considered as a solid surface for snow loads because of the cornicing (bridging) between bearing bars.

Design Considerations

Because of the durability issues noted above, the lateral load resisting system (LLRS) for platforms should strive to use bolted connections whenever possible. It is common for lateral systems to consist of bolted knee braces, also called kickers, which are typically double or single angles. To accommodate flashing the posts, the work point of the knee brace is usually 18 in. to 24 inches above the top of the roof structure, allowing for roof insulation depth. Bolted moment connections are an option, but grating can conflict with the bolted connections. Shop welded moment connections or use of continuous framing (i.e., beam-over-column) are also options.

The geometry of knee braces is such that, under both gravity and lateral loads, the knee-braced joints can have significant horizontal forces at the top and bottom of the post (i.e., the post connection to the roof structure below and the post-tobeam connection above). In particular, the post connection to the roof framing and the transfer of horizontal forces into the

roof diaphragm must be checked. In addition, platform beam-to-post connections must be designed for the horizontal axial forces in the joint.

A staged construction approach could be used in which the kickers are installed after equipment and other permanent dead loads to reduce the horizontal demand under gravity loading. However, this method results in a complicated analysis and construction approach and should be avoided for most platforms.

The diaphragm for platforms typically utilizes horizontal plan bracing to resist in-plane lateral loads. Grating cannot act as a diaphragm because it has relatively little in-plane stiffness, is discontinuous, and is not usually permanently fastened. For small platforms, it may be possible to transfer in-plane loads through weak-axis loading of connections; however, this requires careful connection detailing and is typically impractical for platforms with even marginal lateral loads.

Framing that supports equipment is typically designed to more stringent deflection criteria than those in the International Building Code. Most units have operable doors or are assembled from segments in the field. Framing that is too flexible will cause doors to jam or prevent bolting. In the authors’ experience, the maximum framing differential deflection along a unit should not exceed ½ inch. However, more or less stringent requirements should be reviewed and coordinated with the equipment manufacturer. Equipment weight may be modified during design coordination, so consider designing framing to have reserve capacity.

Conclusion

The structural design of equipment platforms is a detail-oriented engineering exercise for these unique structures. A successful design relies on understanding the equipment on the platform and the end use of the platform, which can be achieved through close coordination with the project team and platform end users.■

Full references are included in the online version of the article at STRUCTUREmag.org

Figure 3 Snow on metal grating

John P. Phelan, P.E. is a Senior Consulting Engineer at Simpson Gumpertz & Heger’s office located in Waltham, MA (jpphelan@sgh.com)

Michael F. Hughes, P.E., S.E. is an Associate Principal at Simpson Gumpertz & Heger’s office located in Waltham, MA (mfhughes@sgh.com)

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