STRUCTURE magazine | July 2017

July 2017 Wind/Seismic

Inside: Navy Pier, Chicago

A Joint Publication of NCSEA | CASE | SEI

STRUCTURE ®

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CONTENTS Columns and Departments

Cover Feature

34 Chicago’s Navy Pier Centennial Wheel

EDITORIAL

7 Determining the Indeterminate

By John E. Sheridan, Ken Maschke, P.E., S.E., and Jared E. Brewe, Ph.D., P.E., S.E. Fitting the impressive Centennial Wheel at Chicago’s Navy Pier was a monumental task. Constructing the Ferris wheel on the roof of an operating garage located atop a 100-year-old pier took both structural engineering ingenuity and critical evaluation of construction picks and crane placements.

By Corey M. Matsuoka, P.E. STRUCTURAL PERFORMANCE

10 The Road to Resiliency

By Scott Adan, Ph.D., P.E., S.E., SECB STRUCTURAL SUSTAINABILITY

14 Hurricane-Driven Building Code Enhancements

LEGAL PERSPECTIVES

52 Key Concerns in Consent to Assignments

By John W. Knezevich, P.E. and Ronald Knezevich

By Gail S. Kelley, P.E., Esq.

STRUCTURAL ANALYSIS

HISTORIC STRUCTURES

18 Lateral Analysis – Part l

54 Lake Champlain Bridge

By Samuel M. Rubenzer, P.E., S.E.

By Frank Griggs, Jr., D.Eng., P.E.

STRUCTURAL DESIGN

SPOTLIGHT

23 Challenges in Cladding Design

59 21st Century Seismic Design Saves an Architectural Landmark

By Steven Judd, S.E.

By Bryan Seamer, S.E.

STRUCTURAL PRACTICES

26 Observing Deficient Façade Repairs

By Dan Eschenasy, P.E., SECB

64 Designing for Tornados

By Roy Denoon, Ph.D.

TECHNOLOGY

31 Equilibrium Finite Elements for RC Slab Design

STRUCTURAL FORUM

67 4 Ways to Empower Your Team

By Angus Ramsay, M.Eng., Ph.D., C.Eng.

By Solomon Ives, P.E.

and Edward Maunder, C.Eng., Ph.D. EDUCATION ISSUES

49 Higher Education That Includes Timber Engineering

By Uchenna Okoye, P.E., Michelle KamBiron, P.E., S.E., SECB, Brent Perkins, P.E., S.E. and Craig Barnes, P.E., SECB

INSIGHTS

IN EVERY ISSUE 8 Advertiser Index 56 Resource Guide – Concrete Products 60 NCSEA News 62 SEI Structural Columns

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement

by NCSEA, CASE, SEI, C 3 Ink, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

STRUCTURE magazine

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July 2017

Features 38 REVAMPED FERRY MAINTENANCE FACILITY

By Percy R. James, P.E. Designing a new fiveslip layout constrained by property limits and requirements for withstanding the effects of wave action, currents, and wind events were the challenges faced by structural engineers at the Port Bolivar Ferry Facility in Galveston.

43 FANEUIL HALL MARKETPLACE

By Rimas Veitas, P.E., Derek Simpson, P.E., and Michael Cronenberger, P.E. A popular location in Boston’s urban market place experience, the Faneuil Hall revitalization project encountered several challenges. Not only a sensitive historic area, the project presented limited access issues and difficult subsurface soil conditions.

46 MASS TIMBER AS STRUCTURE AND FINISH

By Alan Organschi Common Ground High School in New Haven, CT, envisioned a new facility that embodied sustainability and sensitivity to the environment. Mass timber components would become the project’s most innovative feature.

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Editorial

Determining the Indeterminate

new trends, new techniques and current industry issues

Develop a Risk Management Plan

By Corey M. Matsuoka, P.E., Chair CASE Executive Committee

S

and impact for each risk. The design team can also organize the risks by category, timeline (how soon a risk is expected), etc.

Those risks with a high probability of occurrence or high impact to the project are deemed a significant risk and are forwarded to step three of the risk management plan: risk response planning. In this section, a response plan is developed for every significant risk.

For negative risks, these response plans fall in one of four categories: • Avoidance: Change the project to eliminate the threat posed by an adverse risk.

• Transference: Shift the negative impact of a threat, along with the ownership of the response, to a third party. This is often done financially through insurance contracts or operationally through outsourcing an activity.

• Mitigation: Reduce the probability and impacts of an adverse risk to an acceptable threshold through intermediate steps.

• Accept: Live with the possibility of a negative impact.

Developing response plans reduces the possibility of negative impacts to your project or prepares you to respond if they occur.

The final step in the risk management plan is to monitor and control the risk. Be on the lookout for risk triggers and review the plan regularly to recognize and acknowledge new risks or modify previously identified ones. The risk management plan should be a living document that is revised as the project moves forward.

To summarize, a formal or informal risk management plan should be developed for every project. The plan should also be updated as more information is learned. The process of creating a plan is: 1) Identify risks 2) Qualitative and/or quantitative risk analysis 3) Risk response plans 4) Monitor and control To learn more about Risk Management, attend our 2017 CASE Risk Management Seminar: Time-Tested Techniques for Managing Your Firm’s Risk. It will be a full day program on August 4, 2017, in Chicago, Illinois. Sessions are geared toward project managers to principles and include the always popular Professional Liability Case Studies. We also couldn’t be happier, as the session kicks off the night before with dinner speaker Ashraf Habibullah, President of Computers and Structures, Inc. (CSI).

For more detailed information about the sessions and how to sign up, contact Katie Goodman at KGoodman@acec.org or visit www.acec.org/education/seminars. For more information about risk management plans and other resources for claims prevention, visit the CASE website at www.acec.org/CASE. By all means, keep finding definition in indeterminate frames. It is what makes the project challenging and exciting, but remember to mitigate the risks that come with it.▪

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STRUCTURE

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tructural engineers are a weird bunch. The vast majority of them are comfortable determining the forces and moments on an indeterminate structure. This, by the very definition of “indeterminate,” shouldn’t be possible. And yet, structural engineers find a way to do it. The more complicated and intricate a project is, the bigger the challenge and the more excited we get.

However, sometimes we get so excited that we forget about the risks associated with that project. After all, engineers are generally an optimistic bunch – what could go wrong?

Unfortunately, structural engineering is a fairly risky business.

Structural engineers have the highest claims-to-revenue ratio among practitioners in the Architectural/Engineering field. This is not to say that structural engineers have the most claims against them. On the contrary, they do not. What they do have is the highest dollar amount per claim among all architects and engineers. If you think about it, when a structure fails, there are significant consequences.

We are talking about significant property damage and threats to life safety. So what do we do about this?

One answer is for every project to have a risk management plan.

Wikipedia defines risk management as the “identification, assessment, and prioritization of risk followed by coordinated and economical application of resources to minimize, monitor, and control the probability and impact of unfortunate events or to maximize the realization of opportunities.” Through experience (i.e., getting burned a few times), seasoned engineers and project managers can do this intuitively. For everyone else, a conscious effort should be made to practice risk management. One way this is accomplished is through the process of creating a risk management plan for each project. For smaller, less complicated projects, the risk management plan can be simple and informal. However, for larger, more complicated projects, a formal plan should be written down and shared with the entire project team.

A good risk management plan should contain an analysis of potential risks as well as a strategy to mitigate those risks should they manifest themselves. The first step in the development of a risk management plan is to identify the risks. A risk is defined as an uncertain event or condition that, if it occurs, has a positive or negative effect on at least one project objective (Project Management Institute).

The project team should determine which risks might affect the project and document their characteristics. The characteristics of risks could include root causes, risk category, and risk triggers (early warning signs that the risk will occur).

The second step in the development of a risk management plan STRUCTURAL is to qualitatively and/or quantiENGINEERING tatively analyze each risk. In this INSTITUTE process, the design team assesses the probability of occurrence STRUCTURE magazine

Corey M. Matsuoka is the Executive Vice-President of SSFM International, Inc. in Honolulu, Hawaii. He is the chair of the CASE Executive Committee. He can be reached at cmatsuoka@ssfm.com.

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July 2017

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Erratum The byline of an article in the May 2017 issue of STRUCTURE magazine, Cruise Terminal Expansion at the Port of Galveston (page 20), was inadvertently missing an author. Ashish Patel, P.E., provides structural engineering services for a variety of projects at LAN including municipal buildings, water and wastewater treatment plants, hospitals, institutional buildings, and marine ports and terminals. Mr. Patel’s name and contact information (APatel@lan-inc.com) have been added to the online version of this article at www.STRUCTUREmag.org.

STRUCTURE

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MARKETING & ADVERTISING SALES sales@STRUCTUREmag.org Joe Murphy jmurphy@STRUCTUREmag.org; Tel: 203-254-9595 Denis O’Malley domalley@STRUCTUREmag.org; Tel: 203-356-9694, ext. 13

EDITORIAL STAFF Executive Editor Alfred Spada aspada@ncsea.com Editor Christine M. Sloat, P.E. publisher@STRUCTUREmag.org Associate Editor Nikki Alger publisher@STRUCTUREmag.org Creative Director Tara Smith graphics@STRUCTUREmag.org

EDITORIAL BOARD Chair Barry K. Arnold, P.E., S.E., SECB ARW Engineers, Ogden, UT chair@structuremag.org Jeremy L. Achter, S.E., LEED AP ARW Engineers, Ogden, UT Erin Conaway, P.E. SidePlate Systems, Phoenix, AZ

Your Opinion Makes a Difference!

On the first of every month, a link to the Editorial Board’s current Issue Survey is posted on the homepage of STRUCTUREmag.org. This survey provides valuable information to the Board on what types of articles readers prefer. Please take 5 minutes to send us your anonymous feedback on the current issue of STRUCTURE!

John A. Dal Pino, S.E. FTF Engineering, Inc., San Francisco, CA Linda M. Kaplan, P.E. TRC, Pittsburgh, PA

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Greg Schindler, P.E., S.E. KPFF Consulting Engineers, Seattle, WA Stephen P. Schneider, Ph.D., P.E., S.E. BergerABAM, Vancouver, WA John “Buddy” Showalter, P.E. American Wood Council, Leesburg, VA C3 Ink, Publishers A Division of Copper Creek Companies, Inc. 148 Vine St., Reedsburg WI 53959 Phone 608-524-1397 Fax 608-524-4432 publisher@structuremag.org July 2017, Volume 24, Number 7 ISSN 1536-4283. Publications Agreement No. 40675118. Owned by the National Council of Structural Engineers Associations and published in cooperation with CASE and SEI monthly by C3 Ink. The publication is distributed free of charge to members of NCSEA, CASE and SEI; the nonmember subscription rate is $75/yr domestic; $40/yr student; $90/yr Canada; $60/yr Canadian student; $135/yr foreign; $90/yr foreign student. For change of address or duplicate copies, contact your member organization(s) or email subscriptions@STRUCTUREmag.org. Note that if you do not notify your member organization, your address will revert back with their next database submittal. Any opinions expressed in STRUCTURE magazine are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, C3 Ink, or the STRUCTURE Editorial Board.

STRUCTURE® is a registered trademark of National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

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Structural Performance performance issues relative to extreme events

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lmost three years have passed since the 2014 magnitude-6.0 South Napa earthquake. The event was the largest in the San Francisco Bay Area since the 1989 Loma Prieta earthquake. Ground shaking was significant with accompanying maximum Modified Mercalli Intensities between VIII and IX. Following the quake, many historical unreinforced masonry (URM) buildings in the Napa downtown area sustained damage. Some of these buildings are now restored or are currently undergoing restoration efforts. Others are still awaiting restoration or are in limbo.

Following the earthquake, the Post-Disaster Performance Observation Committee (PDPOC) of the Structural Engineers Association of California (SEAOC) deployed key personnel as part of its Earthquake Performance Evaluation Program (EPEP) to survey select affected areas. The PDPOC, a subcommittee of SEAOC’s Existing Building Committee, was established in 2006 to gather building performance information that can be directly correlated to measured ground motions. Following an earthquake, all buildings within a “pod” that registered minimum ground motion intensity triggers are surveyed. A pod includes buildings and structures within a 5001000 feet radius of a Strong Motion Instrument Program (SMIP) station.

The Road to Resiliency 2014 South Napa Earthquake Retrospective

By Scott Adan, Ph.D., P.E., S.E., SECB

2014 South Napa Earthquake

Dr. Scott Adan is Director of Seismic Assessment Services with CBRE and Principal with Adan Engineering in Santa Monica, California. He serves on the PostDisaster Performance Observation Committee (PDPOC) for the Structural Engineers Association of California (SEAOC). He was deployed to Napa following the earthquake and has subsequently tracked the progress of earthquake recovery efforts. He can be reached at scott.adan@cbre.com.

Post-Disaster Deployment Following the earthquake, EPEP’s Phase 1 deployment focus was on the downtown Napa pod.

Within the pod, there are a significant number of commercial and civic buildings, both historic and modern. It includes a significant number of URM buildings. The objective of the initial stage, or Phase 1 of the deployment, was to collect pertinent data for as many buildings within the pod boundaries as possible. A secondary follow-up deployment occurred on May 23, 2016.

Of the 42 buildings surveyed, URM structures made up almost half of the inventory. Six of those are highlighted below to illustrate representative development beyond the initial EPEP Phase 1 survey. They illustrate the complicated nature and timeframes associated with restoring URM buildings to fully operational status.

1219 1st Street (Goodman Library) Built in 1901, the Goodman Library is a two-story retrofitted stone masonry building. A partial retrofit of the building was completed in 1975 and a more comprehensive one in 2004, following damage sustained in the 2000 Yountville earthquake. The federally funded 2004 project was designed in accordance with the 1997 Uniform Code for Building Conservation (UCBC). The retrofit included the installation of tension and shear ties, parapet strengthening, and the installation of a concrete shear wall behind the front façade.

Reportedly, work on the stone masonry turret was limited to avoid compromising the building’s historical integrity.

During the earthquake, as shown in Figure 1, the building’s turret was significantly cracked and damaged. Also, it was reported that excessive outward movement and damage occurred at all four building corners. The building was posted “Unsafe.”

The magnitude-6.0 South Napa Earthquake occurred on August 24, 2014, at 3:20 am (PDT) with an epicenter located approximately 8.0 km (5.0 miles) south-southwest of Napa. The quake was felt widely throughout the northern San Francisco Bay Area, with the most severe shaking recorded in the cities of Napa, American Canyon, and Vallejo. The hypocenter was located at a depth of 10.0 km (6.3 miles).

Significant shaking lasted less than 10 seconds, depending on location. In downtown Napa, the SMIP station recorded approximately 6 seconds of strong ground motion. The recorded peak ground accelerations (PGA) were 0.61g (north-south) and 0.32g (east-west). Total damage in the southern Napa Valley and Vallejo areas was in the range of $363 million to 1 billion. One person was killed and 200 injured.

Figure 1. Goodman Library turret and front façade damage.

10 July 2017

the building’s placard had been revised to “Restricted Use” to allow access for repairs. Repair work began this spring.

840-844 Brown Street (Alexandria Square)

Figure 2. Alexandria Square following the 2014 South Napa earthquake.

After the damage, the building was targeted for repair and restoration. California’s State Mitigation Assessment Review Team (SMART) in collaboration with the Office of Emergency Services (Cal OES) conducted an assessment of the retrofit and damage.

In the interim, the building has been fitted with temporary scaffolding on the north, south, and west sides. Additionally, at the time of the EPEP follow-up survey,

Built in 1910, the Alexandria Square, also known as Plaza Hotel and Annex, is a three-story retrofi tted unreinforced brick masonry building.

Th e retrofi t reportedly occurred circa 1984 to 1986, and included the installation of steel moment and braced frames within the building’s interior and the installation of tension anchors at the diaphragm-to-wall connections.

During the earthquake, as shown in Figure 2, a portion of the third-story cupola collapsed onto the sidewalk below. At these high cupola locations, steel strongbacks were anchored to the walls. However, as shown in Figure 3, the anchorage and

strongback bracing was ineff ective at this level. Other portions of the building were relatively undamaged. The building was initially posted “Unsafe.” However, shortly after the earthquake, the less heavily damaged portions of the building were reopened, and the placard re-posted to “Restricted Use.” Subsequently, the building was immediately targeted for repairs and restoration.

At the time of the EPEP follow-up survey, the building was essentially restored and rebuilt. Alexandria Square officially reopened in June of 2016. Reportedly, repair costs exceeded $1 million. continued on next page

Figure 3. Ineffective steel strongback bracing in the third-story cupola at Alexandria Square.

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Figure 4. Napa Law Center following the 2014 South Napa earthquake.

816 Brown Street (Napa Law Center Building) Built in 1904, the Napa Law Center Building is a two-story, unreinforced brick masonry building. Prior to the earthquake, the building’s owner had been considering a retrofit as mandated by the city (Napa Municipal Code).

However, at the time of the earthquake, no retrofit upgrades were constructed. During the earthquake, a portion of the second story south wall separated from the roof diaphragm and collapsed onto a car in the adjacent parking lot (Figure 4). Immediately following the earthquake, the instabilities associated with the URM walls forced the temporary closure of neighboring businesses. The building was posted “Unsafe.” Following the erection of wooden protection barriers, the surrounding buildings were re-posted “Inspected.”

In subsequent dealings with city officials, the building’s owner was permitted to remove the front façade and redevelop the site. The plan is to number and inventory the stone blocks and then reuse the facade as part of a new five-story reconstruction on the same site. At the time of the EPEP follow-up survey, the reconstruction process was progressing.

The adjacent York Building, once home to the Napa City Council chamber, was demolished as part of the redevelopment plan.

Reportedly, construction will require at least two years and cost approximately $20 million.

1352 2nd Street (USPS Napa Franklin Station) Built in 1933, the U.S. Post Office (USPS) Napa Franklin Station is a single-story steel frame with unreinforced brick masonry infill.

It has a partial mezzanine and a basement. As

Figure 5. USPS Napa Franklin Station following the 2014 South Napa earthquake.

shown in Figure 5, during the earthquake, the massive brick masonry piers on both ends of the building ruptured and shifted laterally.

The large inclined cracks exposed damage to the underlying steel frame columns. A number of the building’s windows were also damaged. The building was posted “Unsafe.”

Following the earthquake, temporary bracing was constructed around the damaged exterior piers, windows were boarded up, and the location has since remained closed. In July of 2015, the USPS proposed demolishing the building. The agency estimated that quake repairs would cost $8 million, while demolition would cost only $500,000. Following a public outcry, the agency decided to place the building on the market. It was reported in March 2017 that a local developer and broker purchased the property for $2 million, with the intention of restoring it into a boutique hotel with a possible residential component.

occurred through many of the window spandrels and, at the southeast corner, a portion of the façade collapsed. The building was posted “Unsafe” and has since remained closed.

Temporary bracing and a moisture barrier were constructed around the building’s exterior.

After the damage, the building was targeted for a significant repair and restoration effort.

The Federal Emergency Management Agency (FEMA) has tentatively agreed to pay more than $500,000 toward the $635,000 building’s earthquake insurance deductible. The Napa County Board of Supervisors approved a $3.3 million contract to design and manage the restoration project. This is in addition to the previous $2.4 million contract to assess and stabilize the damaged building. Some of the walls are so severely damaged they are slated to be disassembled and reassembled, brickby-brick. The repair work is scheduled to be complete in July of 2018.

825 Brown Street (Napa County Superior Court building) Built in 1878, the Napa County Superior Court is a two-story, unreinforced brick masonry building. In 1977, the building was partially retrofitted. The retrofit included installation of tension anchors around the diaphragm-towall connections.

During the earthquake, the building’s URM façade sustained significant damage Figure 6. Napa County Superior Court building following the (Figure 6). Diagonal cracking 2014 South Napa earthquake.

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1130 1st Street (Gordon Building) Built in 1928, the Gordon Building is a two-story unreinforced brick masonry structure. The building has not been retrofitted.

During the earthquake, the building’s rear brick masonry bearing wall sustained significant damage (Figure 7). Diagonal cracking occurred through the supporting brick piers and lower wall. Horizontal cracking occurred at the base of the roof parapet wall and above several of the window spandrels. The building was posted “Unsafe” and has since remained closed. Temporary vertical bracing, parapet bracing, and tension anchors at diaphragmto-wall connections were constructed to stabilize the structure.

The building changed ownership in October 2013, just one year before the earthquake. As part of the new owner’s due diligence efforts, a 3D laser scan of the building was performed in January of 2014. After the earthquake, the building was targeted for a unique follow-up investigation – using a supplemental 3D laser scan to compare the before and after building condition. The rescan optimized the ability to measure cracking and damage.

The scanning investigation concluded that the earthquake motions expanded areas of previously known weaknesses and also created new ones. However, at the time of this writing, there were no known further investigation or restoration plans associated with returning the building to operational status.

The Road to Resiliency Figure 7. Two-story Gordon Building following the

Given their complicated 2014 South Napa earthquake. nature and the construction processes associated with their restoration (i.e. downtown Napa have returned to fully operabrick-by-brick reconstruction), a significant tional status. However, as illustrated here, amount of time can pass prior to earthquake some restoration efforts are either ongoing damaged URM buildings returning to fully or uncertain.▪ operational status. In some cases, public versus private sector designations can affect restoration times. The ongoing restoration The online version of this article contains efforts in Napa exemplify the challenges and detailed references, which expand on much unique avenues employed. In terms of overall of the information reported here. Please restoration efforts, most of the buildings in visit www.STRUCTUREmag.org.

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www.buysuperstud.com/DeflectionClips or call 800.477.7883 US Patent 6213679; other patents pending

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Structural

SuStainability sustainability and preservation as they pertain to structural engineering

D

esigning structures to resist extreme wind events is critical to providing sustainable structures that perform as needed throughout their design life. While the primary structural system is usually sufficiently designed by the structural engineer, the building envelope is composed of a variety of elements with minimal input from the structural engineer. Due to the wide range of systems and elements that compose the building envelope, the building code provides one of the strongest tools available to communities to implement a sustainable building stock that functions after extreme wind events.

A review of the building code enhancements over the past 20 years shows a focus on exterior envelope components and provides insight for engineers practicing in regions with less rigorous code requirements. Many of these provisions are gaining traction in tornado-prone regions as jurisdictions look for ways to mitigate tornado

conservative requirements for the resistance to windborne debris and slightly higher factors of safety on tested components. Because this standard is relatively new and not commonly used, the extent of systems and products available to comply with these provisions is limited.

The wind mitigation provisions adopted in hurricane-prone regions, especially the High Velocity Hurricane Zone of the Florida Building Code utilized in South Florida, are similar to the provisions specified in ICC 500, but with numerous economic building envelope systems already developed to comply.

Historical Code Review A review of the evolution of the codes utilized in South Florida is provided so one may understand the history and basis for these provisions.

This provides an engineer a background to evaluate the measures that may be based on scientific principles rather than political ones so that one may implement the provisions appropriate for a given project.

South Florida has led the country in hurricane-related building code enhancements since the devastating impacts of Hurricane Andrew in 1992, and by Katrina and Wilma in 2005. Before Hurricane Andrew, South Florida did not require the use of storm shutters to protect glazing. The only building code provision addressing hurricane protection stated that if storm shutters were used, deflection should be limited to avoid breaking the glass behind them.

In August of 1992, Hurricane Andrew exposed the shortcomings of the existing South Florida Building Code. The storm struck a less densely populated area of South Florida but was more than three times as costly as any other previous hurricane, causing over forty-five billion dollars in damage (adjusted to 2010 dollars). Of interest, the destruction differed significantly among neighborhoods. In some neighborhoods, all the houses were severely damaged, while in other adjacent developments the damage was minimal.

What made the difference in these neighborhoods

Hurricane-Driven Building Code Enhancements

By John W. Knezevich, P.E. and Ronald Knezevich

John W. Knezevich is President of Knezevich Consulting, LLC. Mr. Knezevich is a member of the American Association for Wind Engineering (AAWE) and a member of FIU’s Wall of Wind Technical Advisory Committee.

He may be reached at jwk@knezevich.com.

Ronald Knezevich is a freshman at the Georgia Institute of Technology and a summer research assistant at Florida International University’s Wall of Wind research facility.

He may be reached at rwk@knezevich.com.

damage. Whether applied to residential or commercial projects, utilizing existing building code provisions from hurricane-prone regions provides an economic methodology for designing structures that are more resistant to extreme wind events.

For a structural engineer to provide leadership on a project, one should be able to discuss the advantages of designing beyond the “minimum” building code requirements. Owners or developers of critical facilities may be especially interested in understanding the economy of available options.

Designing for Extreme Winds To appreciate the basis for utilizing design provisions greater than those specified in the local building codes, an understanding of the relationship between design winds for hurricanes and design winds for tornados is useful. In South Florida, the basic wind velocity required when utilizing ASCE 7-10 is 175 mph for Category II structures and 186 mph for Category III and IV structures. These wind speeds correlate to an EF4 tornado as shown in Table 1. Design wind speeds for most tornado-prone regions is 115 mph. In ASCE 7-10 and the Enhanced Fujita Scale, wind speeds are based on a 3-second gust and commonly termed “ultimate” wind velocities.

While the ICC/NSSA Standard for the Design and Construction of Storm Shelters (ICC 500) addresses the design of structures to resist specific tornado conditions, it is intended for storm shelters. Many of the provisions of ICC 500 are similar to those used in hurricane-prone regions, but with more

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Table 1. Enhanced Fujita Scale wind speeds.

Figure 1. Hurricane Andrew damage in four neighborhoods. Courtesy of Aerial Cartographics of America, Inc.

was the design and construction methods used to build the homes (Figure 1).

Recognizing the differences in performance levels, Miami-Dade and Broward Counties created and adopted a new South Florida Building Code that incorporated many of the observed hurricane resistant features and added entirely new requirements. The most significant change was the requirement that the building envelope resists wind-borne debris. As ASCE 7 internal pressure coefficients indicate, allowing wind to build up inside a structure increases the negative pressures the envelope components must resist; thus, breaches in the building envelope lower the failure threshold. The use of large and small impact testing for all exterior envelope components provides a methodology to address breaches caused by flying debris. Large missile resistance was required for all exterior envelope elements in the first 30 feet above grade. Small missile resistance was required for all elements above 30 feet.

In the large missile test, a 9-pound, 2x4 stud (approximately 8 feet long) is fired at the window, shutter, or wall system at a speed of 34 miles per hour. For the small missile testing, ten 5⁄16-inch-diameter steel balls (initially small gravel was specified) are shot at ADVERTISEMENT–For Advertiser Information, visit www.STRUCTUREmag.org

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the test specimen at 50 miles per hour. In both the small and large missile tests, the test specimen must resist cyclic forces similar to hurricane winds for over 3 hours, and this occurs after two or more impacts to the specimen. Furthermore, the test must be repeated on three samples. While the glass may crack during testing, there can be no cracks more than 1⁄16-inch wide and 5 inches long through which air can pass. While this test procedure was rigorous and unique at the time, it is now standardized as ASTM E1886 and ASTM E1996, and codified in the International Building Code (IBC). An abundance of building envelope products exist that comply with this standard.

South Florida’s Unique Code Provisions There is one important distinction between the impact requirements of the IBC versus those used in South Florida. The impact requirement in the IBC is based on ASCE 7’s statement that glazed openings must be protected in wind-borne debris regions. In South Florida, all building envelope components must be impact resistant, including exterior wall systems. Glazing is not a relevant

factor. For buildings utilizing popular Exterior landfall in Florida. The following year, Once torn, the system loses its design resisInsulation and Finish Systems (EIFS), it is Hurricanes Katrina and Wilma also made tance as wind gets under the system and the worth noting that the major manufacturers landfall in Florida. This increased storm activ- failure propagates. Proper design and detailhave developed impact resistant solutions for ity initiated more changes in the building ing of rooftop equipment is critical to the this requirement. code. Since there had been significant con- sustainability of a building when subjected While glazed openings may present a more struction under the new codes, researchers to extreme winds. Depending on the project obvious susceptibility to breaches from wind- were able to study the damage from these and the size of the specific piece of equipborne debris, lightweight exterior wall systems storms and evaluate the effectiveness of the ment, rooftop dunnage may be engineered consistently fail during high wind events. The new codes, identifying additional mitigation by the Engineer of Record or placed by a use of impact resistant wall systems provides measures that were needed. subcontractor and never actually engineered not only a level of protection from windborne Due to observed uplift failures of roof tiles, to resist wind loads. The extent of damage debris but also provides a higher level of con- especially at the eaves, eave tiles now require from rooftop equipment led to more research fidence in the actual wind resistance of the a metal clip tying the leading edge to the roof to address the issue. Subsequently, ASCE 7-05 system since it has passed a battery of wind deck. Excessive failures were also observed introduced increased horizontal wind loads and impact tests. at hip and ridge tiles; thus, special adhesives for rooftop equipment based on research in Other significant changes were incorpo- or mechanical fastening are now required the wind engineering community. Further rated into the building code after Hurricane rather than simply using a bed of mortar. research provided the data for ASCE 7-10 to Andrew. While many are only applicable to Field testing of the installed hip and ridge tiles address uplift loads on rooftop equipment, as residential construction, the requirement for became mandatory after installation to assure this was not previously addressed. minimum 5⁄8-inch plywood roof decks and the minimum uplift requirements were satisfied.

Unfortunately, while ASCE now addresses prohibition of using oriented strand board Glass debris during Hurricane Wilma these equipment loads, their use is limited (OSB) affects many commercial projects prompted a change to the glazing require- in practice. Rooftop equipment must be as well. Plywood roof deck connections are ments. The sacrificial exterior lite of glass in properly fastened to engineered supporting also required to use nails; the use of staples some small missile resistant systems had no elements and the equipment housing must is forbidden. specific requirements, but breakage of this be designed to remain intact, or the equipSince the proper design and bracing of wood glass created large and damaging debris. The ment itself will be tossed about the roof gable ends above masonry walls seems to be code was modified to require the use of safety causing the same damage. Applying proper misunderstood by many in the residential glass for this exterior lite to minimize the design requirements to the manufacturers’ marketplace, gable ends of masonry walls effect of any resulting debris. housing caused such a lack of equipment are simply prescriptively required to be full availability in Florida, the FBC has removed Roof Top Mechanical height. The common practice of stopping the the requirement for the higher loads on the masonry wall at the low eave height and filling equipment itself. For reliable performance Equipment the gable with wood framing and nominal during and after an extreme wind event, roof bracing is prohibited.

Building breaches not only cause increased top equipment must be properly addressed by Many of these changes were initially resisted wind pressures on cladding elements, but the Engineer of Record through appropriate by the building industry due to costs, but they also allow water infiltration which sig- delegation and review procedures. the most important provision, the missile nificantly affects losses in terms of dollars impact testing requirements, were adopted and business interruption. Another primary Building Code Evolution

by Palm Beach County in 1995 and Monroe source of building breaches occurs at the County in 1997. Despite the continued resis- roof level with damaged roofing systems. Building codes addressing the design of tance from builders, the new Florida Building Often, the roof damage is a result of poorly hurricane resistant structures have signifiCode (FBC) adopted these impact provisions supported mechanical equipment tumbling cantly changed and been enhanced over the in the first edition published in 2001. This across a roof and tearing the roofing system. past few decades due to continued research. new FBC incorporated missile impact testing for all Florida counties in windborne debris regions, while separately specifying all South Florida’s original changes for Miami-Dade and Broward County. The Miami-Dade and Broward County provisions were officially named the High Velocity Hurricane Zone (HVHZ) provisions and are applicable only to those two counties. The HVHZ continues to be the only region that requires impact resistance of the entire building envelope, not just glazed openings.

In 2004, Hurricanes Charley, 1Wind-borne debris regions are areas of high wind velocities, defined specifically in ASCE-7-10, Section 26.10.3.1.

Frances, and Jeanne all made Table 2. Code comparison of building envelope protection.

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sustainable structures. An understanding of the various code provisions available that provide criteria for resisting extreme wind events is essential. Table 2 provides a simplified summary showing the significant levels of wind mitigation provided in several readily available code documents.

While enhanced design specifications are not always appropriate for a project, the structural engineer is best positioned to evaluate a client’s objectives and determine whether they should be considered. While

most of these advancements in wind mitigation can address building elements outside the primary structural frame, the structural engineering of these elements is crucial to a building sustaining an extreme wind event with minimal damage. In some regions, it may be an expansion of scope, but structural engineers bring added value to building owners and architects if they introduce wind sustainability concepts that exceed minimum building code requirements with minimal economic impact.▪

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Unfortunately, much of this research is conducted in the field during the aftermath of a hurricane or tornado. Because building codes are a balance between science, engineering, economics, and politics, shortcomings in codes are often realized but not addressed until the damage is repeated or particularly catastrophic.

A more recent example of a local building code being strengthened after extreme wind events is in Moore, Oklahoma. A devastating EF5 tornado strike in 1999 raised awareness, but it took EF4 strikes in 2003 and 2010, then devastating EF4 and EF5 strikes in 2013, before the political will existed to strengthen the local building codes. While the Moore building code enhancements are significant, they are simply prescriptive requirements for residential construction addressing some common failure mechanisms and fall short of requiring that homes be engineered to resist extreme winds. In South Florida, a professional engineer (or Registered Architect) is required to structurally design all buildings, including residential homes, a unique requirement in the residential construction industry.

For commercial or residential projects, simply engineering the structures to resist wind speeds similar to those used in South Florida provides a resilient structure with a much lower probability of failure during an extreme wind event. Designers in tornado-prone regions have significant code provisions that may be referenced and utilized to provide clients and communities with buildings that are more sustainable than would otherwise be provided using only local building codes.

Enhancing the building code is an evolutionary process. In the coming years, more communities will adopt specific residential design features and maybe even higher design wind velocities for all buildings, but more wind events will show that the changes are not enough. Once building structures and roofs are able to resist a reasonably significant wind event, we will then see less excessive damage from building breaches and cladding failures. Eventually, higher design wind speeds and impact protection for exterior cladding elements will seem appropriate for many buildings in tornado-prone regions.

Until then, structural engineers need to lead the way in designing more

Structural analySiS discussing problems, solutions, idiosyncrasies, and applications of various analysis methods

Lateral Analysis Part 1: Right Way, Wrong Way with Software

By Samuel M. Rubenzer, P.E., S.E.

Samuel M. Rubenzer is the founder of FORSE. He has also been the structural engineering consultant to Structural Masonry Coalitions in several states.

He can be reached at sam@forseconsulting.com.

T

he increasing ease of performing a Lateral Analysis of a structure is becoming a double-edged sword: there are many benefits, but it can also be quite dangerous. A fair share of presentations and articles from seasoned engineers warn about colleagues losing a sense of the real behavior of structures, or that engineers today simply do not know how to design structures without a computer.

To rationalize such comments, one may remember that the engineering curriculum remains similar to the education that engineers received 10, 20, or even 30 years ago. Many engineering programs focus on statics, dynamics, and material properties courses. So what explains this current perspective that today’s engineers are not as grounded as their predecessors in their understanding of structural engineering design?

One explanation is rooted in the thought that structural engineers are overly reliant on software programs and that software processes are replacing engineering judgment.

Software programs should make better engineers, not worse. They are tools and should be treated as such. Each program has unique and varying abilities to create a representation of the real structure, some with more features and options than others. However, it is better to view all structural engineering software programs as incredible graphic user interfaces, able to solve complex sets of equations and run predefined formulae. It should not be assumed that software programs can understand the complexity of the structures and the loads that need to be applied, or be able to devise and create unique solutions.

These tools provide the ability to solve problems very efficiently and iterate design options until the best possible alternative is developed, provided the problem has been accurately identified. Accurately defining the problem is the main issue. Also, the isolated manner in which engineers work leads to few people thoroughly examining the design problem to ensure it has been defined accurately. Individuals are then left to complete the work with no one watching, in an ever-increasing budget-constrained, schedulecramped environment. This is an opportunity for the shortcomings of using software to take root.

The May 2016 issue of STRUCTURE magazine included an article on the use of the Finite Element Method (FEM) for masonry, and perhaps future articles will feature the specific uses of other materials with FEM. This article, however, is on the broader topic of lateral analysis, and the right way and wrong way to use FEM software programs. Therefore, this piece does not include material-specific recommendations; instead, it discusses lateral load generation, element (beam,

18 July 2017

column, wall) properties, diaphragm properties, loads on diaphragms, element connection to diaphragms, types of lateral analysis, and methods for quality assurance of the lateral analysis.

Brief Review of FEM Basics The FEM is the process of simplifying a real life structure, generally a continuum with infinite degrees of freedom, to a finite number of elements with unique material properties.

The FEM is broken up into three steps: 1) Modeling: Pre-processing step where a user defines elements of the model, element connectivity, support conditions, and forces to represent various loading conditions.

2) Analysis: Processing step that requires little input from the user – users establish a few important parameters and then allow the software to solve vast sets of equations based on modeling.

3) Validation and Design: Post-processing, the step of interpreting and verifying the results of the analysis and then designing elements based on parameters determined by the material codes one uses.

Part 1 of this article examines the first and most important step for FEM, model generation.

Modeling In defining a model, users establish one-dimensional line elements (straight or curved) with two end nodes, and/or two-dimensional plate elements (square, rectangular, or triangular planar shape) with nodes at each corner of the element.

In the process of defining the elements and end nodes, some nodes need to be identified as supports. Others remain as free nodes able to translate in three dimensional (X, Y, and Z) degrees of freedom (DOF), and rotate about the three axes (RX, RY, & RZ).

Support nodes generally have translational and/ or rotational DOFs fixed. Remember that, for the actual condition the model represents, most support conditions are less than the idealized fixed condition. In nearly all cases, it would be more accurate to specify the support as a resistance over a potential displaced distance; in other words, a spring support. This is true for both translational and rotational degrees of freedom. Small differences in these support conditions may have a significant impact on the lateral resistance of the assigned members.

Not that all foundations should be spring supported, but their true behavior should be considered, especially when a building has dissimilar lateral resisting elements such as moment frames and walls where lateral load distribution may be greatly affected by soil stiffness. However, when

dealing with similar resisting elements, such as a building with equally sized and uniformly distributed braced frames, fixed supports or spring supports may have little effect on the outcome of the analysis. Here, a column’s foundation can be modeled as a pinned support (free rotational DOF) since rotational stiffness may have less impact compared to vertical and horizontal translational stiffness.

Modeling Line and Plate Elements

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Section properties and member elasticity for line and plate elements need to be defined.

Many of the software programs are based on a linear elastic analysis, which is sufficient for members that remain elastic under static loads such as dead load, live load, snow load, and even idealized wind loads on a model.

However, additional material properties must be considered when an inelastic response is expected, such as for concrete elements that respond in a non-linear manner when the concrete cracks under tensile stress and engages reinforcement. Inelastic response is also expected for members designed to resist seismic dynamic loading with a response modification factor that is, in part, based on an inelastic response.

This inelastic behavior of concrete is accounted for by reducing the stiffness with an element reduction factor. It is very important to realize that software tools do not make this modification automatically, and require user input through several iterations of modeling and analysis (loads to the member change as stiffness changes and may require further member modification).

Many software programs allow the users to define the geometric boundaries of entire slab elements or wall panels and discretize those large geometries into smaller finite elements by a process called meshing.

Sometimes meshing is a manual process, and other times programs offer automatic meshing. To a certain extent, the finer the mesh (smaller the elements), the better the finite element method can approximate the result. There is a point of diminishing return in which finer meshes only result in a small percentage change in the results. It should be noted that finer meshes also produce significantly increased processing times for the finite element analysis. Having a mesh with nodes closer together than the thickness of the element (this is especially relevant with concrete materials) is unnecessary in most cases, as it would be unreasonable to have differential movement between nodes spaced that close together.

With respect to the elements of the finite moment conditions, it may “collect” rela- diaphragm level to another. Rigid diaphragms element model, the last critical piece of infor- tively large loads that need to be addressed. should not be specified to include isolated diamation is connectivity. This can be simply Another element restraint that is often over- phragms supported at multiple levels. Elements defined as pinned (translational movement is looked and used in a nonconservative manner should be modeled that transition diaphragm shared between elements that share the same for the purpose of eliminating instability warn- forces as load transitions from one diaphragm node) or fixed (translations and rotational ings is torsional restraint (or rotational stiffness level to the other. degrees of freedom are maintained between along the member length). This may not be an Consider the element properties of semi-rigid the elements that share the node). Similar to issue for sections such as concrete or closed steel diaphragms similar to other modeled elements. the nodal degrees of freedom, it is important sections (HSS), but open steel sections do not Settings for in-plane axial and shear stiffness, to note that idealized connections between resist torsion well. This is an example of how and out-of-plane shear and bending stiffness, members are more accurately represented by a simple error in modeling may result in the need to take into account either the elastic elastic springs. Some deformation can occur collection of loads that are not being checked or potentially inelastic behavior of semi-rigid between the two elements at a joint otherwise during design. diaphragm elements. considered as rigid, just as some rotational stiffness occurs for most connections specified as Modeling Diaphragms Modeling Element Stiffness pinned. Just as with nodes, not all joints need to be connected with springs, but consider the A very important criterion of lateral load- All of these options for nodes, one and twoimplication for each element end to deform ing for buildings is the types of diaphragms dimensional members, and diaphragms changes independently compared to the idealized con- that are defined. With nodes, line element the lateral load distribution. The more strength dition and model accordingly. For example, it columns and beams, and plate element walls, and stiffness that is represented by an area of the would be challenging to have a forty-inch-deep many programs offer the ability to define a model, the more the lateral load is distributed steel beam with thirteen rows of bolts act as a diaphragm constraint instead of requiring to that area. It is important in the modeling truly pinned end condition. plate element slabs to be modeled. Both rigid phase to define actual properties. Far too often, Another option for one-dimensional mem- and non-rigid diaphragm types are idealized users take shortcuts such as defining idealized bers in most software programs is the ability to simplify analysis. Rigid diaphragms fix the support conditions, not defining section modifito shorten members using cation factors because it takes rigid offsets, based on the too much time, or defining There is no such thing as conservative modeling; all efforts dimension of the member material properties that are should be made to be as accurate as reasonably possible into which it is framing. arbitrarily low in an effort to This allows a user to create when defining the model. The design step is the appropriate be conservative. Not only are a model that reflects the each of these and other modtime for implementing conservative principles. actual joint size and should eling shortcuts incorrect, but be considered since all memthey lead to inaccurate lateral bers have a physical dimension (width and translation of all nodes of a similar elevation load distribution. This results in some areas depth). Users define members using centerline relative to one another, while non-rigid dia- being assigned too much load (regions with too modeling, but then the program recognizes phragms allow free horizontal translation of much stiffness) and other areas being assigned the member lengths to be the elastic portion one node relative to another. too little load, leading to unconservative designs between rigid links. This creates a stiffer model, Is this idealized modeling necessary? In fact, (regions with too little stiffness). wherein the elastic portion of the element is this approach of trying to capture the true We all would like to think of ourselves as being shorter and reactions are at the face of the joint. diaphragm behavior is even named semi-rigid progressive in our industry by using FEA softWhen building a finite element model, one diaphragm modeling. Geometric irregularities, ware. Much more progress should be made. It can over-restrain members to nodes. For lateral resisting elements with different mate- would be wrong to blindly use software tools example, if there are two-line elements that rials and different types (walls and frames), without fully understanding them, and also share a common node, they are connected, or diaphragms with relatively large openings, wrong to not fully utilize the tools. As author and one can choose each members’ connec- should be defined as semi-rigid. Semi-rigid C.S. Lewis states, tivity at the node. If the rotational restraint diaphragms complicate the stiffness of the We all want progress. But progress means getting for each member is released, one will likely finite element model by requiring many plate nearer to the place where you want to be. And receive a warning of a local instability. A element slabs be defined, which again leads if you have taken a wrong turning, then to go node needs to be elastically attached to one to increased time in analysis. Not every diaforward does not get you any nearer. If you are member or the other. However, to avoid phragm needs to be semi-rigid. For example, on the wrong road, progress means doing an the error, engineers fix both axes (not one diaphragms with similar and regular vertical about-turn and walking back to the right road; or the other), possibly resulting in an over- lateral resistance elements and diaphragms with in that case the man who turns back soonest is connected model. Often, at the end of a uniform and consistent slabs with few slab the most progressive man. steel beam, only the strong-axis moment is penetrations can likely be considered rigid.

If you find yourself on the wrong path, recondesigned as a moment-resisting connection.

When it comes to diaphragm action, cau- sider your approach to modeling with FEA The modeled weak-axis moment rigidity tion must be used when there is a step in the software. The next article (Part 2) discusses participates in the models’ resistance to lat- diaphragm. Not only does diaphragm behavior completing the modeling step by offering sugeral loads. Although in most situations this require a vertical element (short wall, braces) gestions regarding applying loads to the model, may be relatively small, it can make a differ- to transfer load, but the diaphragm chords comments on analyzing the model and when ence in load distribution. In certain models (generally at edges or extreme stress locations) to review the results, and finally a discussion where large sections have modeled weak axis must be adequate locally or transition from one regarding the design of members.▪ STRUCTURE magazine

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S

tructural engineers occupy a unique position in building design in that they are the sole resource for most of the information that is critical to the building cladding design. The Structural Engineer of Record (SEOR) is responsible for designing the lateral and gravity force resisting systems of buildings to perform within certain code prescribed seismic (and wind) drift limits. They also provide the design for roof and floor spandrel elements (for multi-story buildings) to perform within specific deflection criteria, based on the exterior cladding system and function.

The proper design of the building enclosure – often by a Specialty Structural Engineer (SSE) – cannot be performed without two specific and clearly defined structural criteria: story drift and spandrel deflections. Project documents should specifically identify real (calculated) building displacements, not just regurgitated code limits. For instance, in a very stiff shear wall system, seismic drift displacements may be a fraction of the code prescribed limits. There is an unnecessary cost impact to the project if only the code limits are defined. Real (calculated) displacements are best, which should include torsional displacements of the floor and roof diaphragms, where they occur.

Slab Edge/Spandrel Deflections Spandrel beam deflections require careful consideration. Live load deflections of spandrel beams should be based on something other than a proportional span limit, like L/360 or L/600.

It is best to provide a specific deflection amount based on the performance expectations of the cladding system. For example, assume a 30-foot, simple span spandrel beam supporting one floor of exterior enclosure framing in a multi-story building. The default rule of thumb of L/360 would produce a deflection of 1 inch. If the dynamic horizontal joint between floors is a sealed joint using a material that has a warrantable movement potential of 50% (silicone or urethane), then the actual joint size would want to be at least 2 inches.

That is 2 inches for only the prescribed live load beam deflection.

In addition to live load deflections, the joint must also allow for superimposed loads added after the exterior wall system is installed, as well as thermal-gradient-driven volumetric changes in the cladding materials, long-term volumetric changes, and creep. A joint larger than 2 inches is a very big joint. A much more visually acceptable joint would be ½- to ¾-inch, which would require the spandrel live load deflections to be less than ¼- or 3 ⁄8-inch, respectively, regardless of span length. For the same 30-foot simple span spandrel beam, a deflection limit of 3⁄8-inch would be L/960, much stiffer than the rule of thumb would prescribe.

It would be best to move away from proportional span deflection limits in relation to spandrel framing and the associated cladding design and move toward a rational, realistic stiffness or deflection criteria based on aesthetics and performance.

The real expectations of the spandrel stiffness and cladding implications should be discussed with the architect and owner early in the design stages of a project.

Structural DeSign design issues for structural engineers

Joint Sizing All foreseeable deflections should be cataloged by the SEOR to determine the proper joint sizes in the exterior cladding systems. For example, for concrete construction, long-term shrinkage and creep should be identified, not only for the slab edge or spandrel elements but the vertical load bearing system – columns or walls – as well. All contributors to floor shortening should be identified. For posttensioned floors, the floor plate shrinks over time, reducing both the width and breadth of the building. Composite construction spandrel beams (concrete deck carried by steel beams linked together via shear studs) will “creep” due to the shrinkage and creep of the topping concrete, even though there is a steel beam present.

The creep effects of composite construction are frequently overlooked in domestic (U.S.) practice.

If vertical or horizontal joints in the cladding system are not sized properly, the exterior cladding can be forced to resist axial loads until failures or yielding occurs. If the cladding material includes brick or other clay-based materials, long term irreversible volumetric expansion should be identified. Brick absorbs environmental moisture and increases in size, after it is first cooled from the firing process, throughout its service life.

It is generally accepted that the building jointing system should function and remain as an uncompromised weather barrier before, during, and after lateral displacement of the building in response to wind and seismic forces that generate an elastic response of the lateral force resisting system. It is also generally accepted that once the building’s lateral force resisting system enters the energy dissipation mode of inelastic behavior, it is acceptable for the cladding system to breach the weather barrier capacity while avoiding damaging contact and the potential of creating falling hazards. Most performance testing of wall systems demands compliance to this protocol. This two-stage performance design concept should be discussed with the owner/developer during the early design phases of the project.

STRUCTURE magazine

Challenges in Cladding Design

continued on next page

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By Steven Judd, S.E.

Steven Judd is the Director of Engineering for KEPCO+, an architectural cladding systems designbuild subcontractor. He can be reached at sjudd@kepcoplus.com.

In-Plane Forces In-plane forces imparted to or generated within the cladding system must be considered. The cladding system must be designed to resist those “internal” forces. The effects of the in-plane shear and any overturning moments that may be induced in a wall panel should be resolved and developed into the building superstructure. In-plane lateral seismic forces for tall, narrow elements, like column covers, may increase vertical shear demand on the stud support fastenings several times greater than the shear gravity load demand alone, especially for heavy cladding materials.

Multi-Story Walls All buildings drift when resisting lateral forces due to wind or seismic events. Buildings that utilize moment frames for the LFRS tend to drift the most, and member sizes in the LFR frames are often controlled by drift limits/ lateral displacement restrictions rather than strength. If exterior walls are multiple stories tall, bypassing the floors, and intended to remain rectilinear during their service life, then the accumulated drift over the multiple floors must be accounted for in the connecting hardware and the interface with other finishes, for example where the roofing membrane bridges across the shear plane at the back of the parapet. Accumulated drift may be more than 8 inches for a 3-story wall.

There is some debate as to whether one should algebraically sum the drifts at multiple floors or take some reduction to account for the real phenomenon that all floors do not experience their maximum drift simultaneously.

This is especially true where a higher mode of vibration controls the generation of the seismic forces. Either way, accumulation of drift should not be ignored where wall systems are continuous across multiple floors.

Building Corners and Seismic (Story) Drift Almost all buildings have corners where adjacent exterior walls converge. Dealing with exterior cladding wall corners is a challenge due to the bi-directional movement required – movement parallel to each of the two orthogonal sides of the building that meet at the corner. Often, the issues of the corners are ignored and the two intersecting walls do not have any accommodation for bi-directional drift. This means that the corner area will most likely fail during the inelastic lateral displacement of the building (Figure 1a). If the cladding material is heavy,

such as stone or brick, then significant hazards may exist without proper corner detailing.

The simplest approach to resolving bidirectional movement at the building corners is to terminate one wall or the other at, or very near, the corner and provide a vertical dynamic joint of sufficient size to allow for the full drift movement of each wall without a clash of the cladding materials (Figure 1b).

This may create a very large joint that is difficult to seal.

A third option is to move the joints away from the corner and frame the “L” shaped corner element as a rigid unit braced to the floor below. Smaller joints can be used as they function primarily in a “shearing” mode (Figure 1c).

Keep in mind, all exterior framing joint scenarios have interior finish ramifications. The interior finishes are largely outside the purview of the SEOR and/or the SSE, but a valued team partner would direct some discussion toward those issues so that they can be discussed during the development of the design.

The more options that are available to resolve corner challenges, the more valuable the services of the engineer become.

drift drives wall tilting

significant corner distress

Figure 1a. With no joint provided at the corner, when walls tilt out of plumb - driven by upper level story drift - the corner wall will fail. joint sized for unencumbered movement

Other Situations of Note Openings in floors adjacent to the exterior wall, such as for stairs, mechanical shafts, and elevator shafts, can create challenges. The supporting slab edge may be nonexistent. Interruptions in the typical exterior wall framing scheme create challenges for continuity of dynamic joints and, specifically, story drift joints. Stairs, in particular, often require framing at intermediate levels (for the landings) that can be difficult to resolve if the support of the studs shift half a floor up or down.

Stacking single story walls that are tied to each floor, but bear on each other without a horizontal deflection joint, can create unintended load bearing walls which are not designed to carry floor loads. The stacked walls are much stiffer than the deflecting spandrel beam, and the floor loads tend to follow the stiffest load path until something fails.

Vertical steps in continuous horizontal drift joints should be avoided wherever possible.

Whenever a vertical transition occurs, the width of the vertical portion of the joint at the transition must resolve the story drift movement without inducing a clash. This may be several inches in moderate to high seismic zones.

Framing at or near egress doors and openings can be a challenge. If a wall deforms in-plane (racking, for instance) and causes egress doors to bind and become non-functional due to poor detailing, then the required means to evacuate the building may be compromised.

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July 2017

Figure 1b. A joint of sufficient size each wall to tilt unencumbered when driven by upper level story drift. joints sized for shear movement

Figure 1c. Rigid “L” corner with minimal width joint (shearing action, not extension and compression). “L” corner needs sufficient bracing and stiffness.

Avoidance of Clashes When detailing cladding joints to accommodate building movements, the designer is required to interpret and apply the appropriate code provisions. For example, ASCE 7-10, Chapter 12.12.3, uses the terms “structure” and “structural(ly)” when defining certain performance issues relative to separations and

12.12.3 Structural Separation

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Conclusion During the process of developing the design of a building, the SEOR and/or SSR should, to the best of their abilities, catalog and clearly convey all foreseeable building movements for the cladding designer’s use. It is also important to inform the client of the necessities of proper jointing in the exterior cladding system for the cladding system to accommodate all relevant building movements. In some cases, the primary structural elements can be made stiffer (extra cost) to reduce displacement, and reduce joint sizes (reducing joint cost). However, more commonly, the cost-benefit ratio of changing the structure versus changing the joint width drives the solution to the cladding joints.

Wide joints may be maintenance challenges, so smaller joints may be preferred.

However, small joints must function to allow for the necessary building movements while avoiding clashes with cladding materials and creating public safety hazards from falling debris. For a complete, thorough design, in-plane forces in the cladding system must be resolved via a complete load path to the supporting elements.

Cladding design can be every bit as challenging, in some circumstances, as the design of the primary structure. It should be approached as an important, necessary, and integral part of a complete building system.▪

STRUCTURE magazine

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“All portions of the structure shall be designed and constructed to act as an integral unit in resisting seismic forces unless separated structurally by a distance sufficient to avoid damaging contact as set forth in this section.

Separations shall allow for the maximum inelastic response displacement (δM). δM shall be determined at critical locations with consideration for translational and torsional displacements, where applicable, using the following equation: δM =Cd*δmax /Ie (Eqn. 12.12-1) Where δmax = maximum elastic displacement at the critical location.”

Adjacent structures on the same property shall be separated by at least δMT, determined as follows: δMT = (δM12 + δM22)½(Eqn. 12.12-2) where δM1 and δM2 are the maximum inelastic response displacements of the adjacent structures at their adjacent edges.

Where a structure adjoins a property line not common to a public way, the structure shall be set back from the property line by at least the displacement δM of that structure.”

Examine the last paragraph. It is logical to assume that the reason a “structure” should be set back from the property line by the amount of the calculated inelastic drift is so that, under the code prescribed circumstances, the “structure” doesn’t sway across the property line and crash into adjacent buildings or other constructions.

It can be interpreted that the term “structure” in this usage means “building,” not “primary structure.” It makes little sense to allow appurtenances, overhangs, awnings, canopies, cornices, and other architectural projections to cross the property lines and create clashes with adjacent constructions. If the reason for the set back is, in fact, clash avoidance, then the “structure” is anything and everything attached to the “entity” we call a complete building. Borrowing from the first line of the quoted text, “all portions of the structure” would appear to mean everything attached to the building.

The first paragraph of the quoted code text above appears to require the building to function “without damaging contact” for “all portions of the structure” at separation joints. Joint separations allow for maximum inelastic response displacement. If the intent of the code were solely focused on avoiding “structural” damage, then it would be satisfactory for architectural features such as fins, cornices, windows, canopies, and awnings to crash together and create falling hazards,

preserving solely the “primary structure.” One can reasonably deduce that falling hazards are precisely what the code is intending to preclude and separations should be sufficient to avoid damaging contact (clashes) with any and all portions of the building.

For lightweight cladding (EIFS, wood, and perhaps stucco,) the consequences of damaging contact are not nearly as severe as damaging contact of dimensional stone slabs, brick veneer, or even CMU or manufactured concrete or stone veneer. However, damaging contact is just that – damaging contact – which is required to be avoided by code.

Picture yourself sitting on the witness stand in a trial where someone got hurt due to falling debris. Consider, would your design decisions regarding building separation or the width of the dynamic joints be appropriate? How would the lay person in the jury interpret “all portions of the structure” knowing the code’s main purpose is to protect life safety? Does the code mean only the structural elements are to avoid damaging contact? If in doubt, it would be prudent for the local building official having jurisdiction over the project to make an official informed and considered interpretation of the intent of the code.

© 2017 ClarkDietrich Building Systems

damaging contact. These terms may lead one to think along the lines of “primary structure,” but that may be an incorrect interpretation of the intent of the code. Here is what the code says (ASCE 7-10; bold added for emphasis):

clarkdietrich.com

July 2017 CD17070_Clips_Structure_ThirdPg_VertAd_June17.indd 15/11/17 11:58 AM

Structural PracticeS practical knowledge beyond the textbook

T

he present condition of a façade is often the result of its aging as well as of the period when it was constructed, as façade solutions are generally specific to periods of time and the profession’s position on the learning curve. The author argues that repair and remediation solutions, evident in any existing façade, are also marked by the level of knowledge prevalent at the time of the repair.

Repair methods have evolved over time. Many first attempts were not adequate, and there have been several cases where repairs failed shortly after completion. The issue of prior repairs creates a dilemma for façade inspectors. This article illustrates this dilemma using examples of repair solutions for exterior corners of transitional masonry façades in New York City.

New Façade Solutions The advent of skeleton construction led to the abandonment of bearing wall structural systems and the development of new façade designs for

Observing Deficient Façade Repairs

By Dan Eschenasy, P.E., SECB, F.SEI

Dan Eschenasy is the New York City Buildings Department Chief Structural Engineer. He is an Honorary Member of SEAoNY and a Member of the ASCE Structural Assessment of Buildings Committee.

incorporating steel into the façade fabric; the painting of steel, required by the building code, had proved to offer poor long-term protection against corrosion.

The introduction of a new solution without fully understanding its long-term reliability was not particular to the transitional façades. It was just the result of the fact that the evaluation of the assemblies was being done primarily by observing “in service” performance.

For a couple of years following their introduction, new solutions might give the impression of success, but many serious defects would take years to surface and be recognized as systemic.

Deficiencies related to weathering, especially corrosion, take a long time to develop; the formation of noticeable scale from corrosion might take 30 years. This time lag allows some systems to take hold for long periods of time.

Occasionally, during these periods, the industry would discover weaknesses and would develop improved solutions.

For example, around 1900, the use of terra cotta in façades became very popular. The industry published a detailing manual, Terra Cotta Standard Construction in 1914. At that time, terra cotta installations were believed to be highly impervious to water. Following a number of failures due to water penetration and subsequent corrosion of terra-cotta anchorages, the standard was revised in 1927 with the authors noting that the changes were the result of a “careful study of the behavior and weathering properties of exterior building materials.”

When inspecting a terra-cotta façade element, it is thus useful to know whether it was installed incorporating the more prudent details of the 1927 manual. The recognition of corrosion of perimeter steel in transitional façades followed

high-rise buildings. During the period between 1900 and 1950, the typical high-rise façade design used the perimeter steel frame to support the weight of the façade at each floor. These façades, referred today as “transitional,” were allowed to be thinner because the masonry had lost its vertical load bearing function. They still had to resist out of plane wind pressures from floor to floor and, in many cases, they were relied upon to contribute to the building’s stability as shear walls. However, fire protection concerns set a limit to the reduction in thickness. These walls were, in many cases, assemblies that consisted of face brick furred (backed) with terracotta blocks. With the incorporation of several additional materials, such as steel structural shapes and anchors, bricks, terracotta blocks, and more, the façade led to the loss of tried and trusted details of construction and the diminution of the traditional role of masonry artisans. The typical details for the transitional façades were disseminated mostly by means of manufacturer brochures and trade pamphlets. It took many years to fully understand Figure 1. Cavity wall from the early 1960s. Different brick colors indicate the problems resulting from various repair campaigns.

26 July 2017

SUBSEQUENT CRACK

CUT JOINT

Figure 2. Cavity brick face collapse. The corroded tie does not engage the face brick anymore.

Figure 3. Ineffective repair. The vertical joint was cut without cleaning steel – a new crack appeared.

Figure 4. Bearing masonry corner repair.

the same pattern and timeline. The significant length of time can also be explained by the fact that the first generation skeleton frames used wrought iron and some types of steel alloys that were less prone to corrosion.

Around the 1950s, cavity walls started to replace transitional envelopes in high-rise buildings. The idea of the cavity wall was derived from hollow wall systems, a solution that had proven successful over time.

Following some positive fire and structural tests near the end of the 1950s, most local codes permitted such construction types.

The first generation lacked sufficient joints in the face brick. It took almost twenty years to understand the serious problems posed by the thermal expansion of the face brick or by the water evacuation from the cavity. Almost all façades built during that period had to undergo a major retrofit that included cutting joints and adding or reconstructing flashing (Figure 1).

The lessons learned from the first cavity walls helped form the detailing practice of later generations. The identification of some of the various systemic defects of the cavity systems – such as flashing details, maintenance of weep-holes, joints in shelf angles, frame/façade movement compatibility – spread over many years and came in increments. From time to time, trade associations and manufacturers updated their recommendations and produced improved details that eliminated the newly identified deficiencies. Any present day evaluation and repair may benefit when it considers the cavity wall condition in light of the specific stage of understanding existing at the time of the original installation and the subsequent repair.

Several recent collapses of face brick have revealed the lack of sufficient metal ties anchoring the brick veneer. Missing or corroded ties created problems and had been reported by forensic engineers since the 1990s, but they were considered rare or unique cases (Figure 2). Because this type of deficiency had not been identified as specific to the first generation cavity walls, remedy efforts may have been limited. Several recent accidents and pressure from the New York City Department of Buildings (NYC DOB) should result in a retrofit campaign to identify and repair this type of defect.

(Figure 3). In contrast to the jagged cracks that form randomly and that may lead to pieces of brickwork falling, a joint is, in essence, a man-made crack that can be managed. Of course, the technology of joint caulking is far from perfect, and a joint itself might become a source of water penetration. However, cutting joints in the outside wythe of cavity walls proved to be an effective solution for releasing the stresses produced by restraints on the tendency of the face brick to move.

Cutting joints without ensuring that the now separated element is anchored to a backup can lead to accidents. On one occasion, the author investigated a case where the face brick of the cavity wall collapsed soon after joints were cut because metal ties were missing. Similarly, cutting joints close to the corners of unreinforced masonry parapets avoids the formation of dangerous diagonal cracks at the corner. However, such solutions might not take into account the loss of lateral support that is offered by the parapet return. Reinforcing the parapet (with or without creating a joint) might be a more reliable solution.

Over-Extending Lessons Learned Aside from a few historic preservation specialists, there were only a very limited number of professionals with a solid knowledge of past façade systems when the New York City Local Law 10 for periodic façades inspections was enacted in 1980. Repair of façades was the domain of contractors. At that time, the cavity wall was the most common solution for new façades, and the thermal movement of face brick was already accepted as a principal cause of distress. Cutting joints in the face brick was a typical retrofit prescription.

Engineers and architects that entered the field as a result of the local law had few resources from which to gain specialized knowledge. With a lack of better solutions, they adopted and extended to other types of façades the use of joints as a repair method.

This might have been a period of over-reliance on joint cutting solutions. Cases exist where façades had horizontal expansion joints cut along steel spandrels or the building’s corners

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July 2017

Corners Cracks at Transitional Façades Old bricklaying manuals indicate that builders were well aware that masonry corners provided “the main strength of brick buildings.” Vertical cracks at bearing masonry corners are rare. When they do occur, they produce a weakening of the load path and might also allow the occurrence of geometric instabilities in the intersecting walls. These potentially dangerous conditions require

Figure 5. Typical corner crack. Cavity wall (left); Transitional façade (right).

immediate restoration of continuity. The most common solution uses steel plates anchored into the intersecting walls. This solution is not expensive but is liable to fail (Figure 4, page 27). The proper repair should entail reconstruction of the masonry at the corner.

Vertical cracks around building corners represent a typical distress of transitional façades (Figure 5). They have occurred in many different face materials: brick, terra cotta, and even stone. In the case of brick faced buildings, long vertical cracks rarely occur at other locations. Masonry cracks at corners of skeleton construction do not necessarily affect the structural stability of the building but might constitute a danger to pedestrians.

The first method that was used for crack repair involved plugging the cracks, with mortar (Figure 6). This avoided the danger of brick separating and falling, and reduced the rate of water penetration. However, the

Figure 7. Collapse at corner. Masonry fill along cut joints.

Figure 6. Ineffective crack repairs at corners. Elastic compound fill (left); Mortar fill (right).

cracks started to open again. Placing elastic compounds in the crack was not more effective. These solutions did nothing to arrest the corrosion of the underlying steel.

Studies on cavity walls indicated that corner cracking was the result of the convergence of movement of masonry on both sides of the building corners. An explanation was provided that seemed pertinent to corner cracks at transitional façades. Under the influence of these studies, a typical 1990s corner repair design involved full height vertical cuts placed close to the corner. The masonry was removed to allow cleaning and rust proofing of the exposed sides of the steel column. The brick was replaced in a manner that created joints along the vertical cuts on both sides.

These new joints served the intent to avoid the damaging combination of movement of the perpendicular fronts, but they severed the horizontal continuity of the masonry.

Collapses occurred in cases when the repair did not include anchoring the new vertical corner stack to the steel (Figure 7). On the positive side, joints on both sides of the corner allowed the new brick to expand when exposed to the weather, without being restrained by the old masonry that had already exhausted its ceramic expansion.

Later studies and reports specific to transitional façades added several other possible causes for cracks, such as the irreversible expansion of fired clay products when exposed to humidity and steel columns restraining the thermal movement of the brick infill. Other causes include incompatibility between the movement of the structural frame and the masonry, and steel spandrel members deflecting and imparting a vertical load to the masonry below.

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These explanations are all valid, but they only describe the original cause of cracking. Today, almost one hundred years later, the visible cracks are the expression of steel column corrosion. Under the conditions of the New York City climate, a crack will enlarge under repeated cycles of icing and the masonry will not revert to its initial shape whatever the initial cause. The cracks facilitate increased water penetration and, when present, the subsequent corrosion of the steel. Once steel begins to corrode, rust jacking takes over and becomes the main culprit for the evolution of cracks.

While not contesting the validity of any of the wall movement explanations, one needs to consider that the level of façade deterioration is in direct relation to the capacity of exterior water to penetrate the skin and induce steel corrosion. Brick is a porous material that, even in the absence of cracks, allows some water to migrate towards the interior. Obviously, the rate and frequency of this migration are dependent on wall thickness. Numerous visible horizontal cracks, typical in transitional façades, are the consequence of beams and spandrels being protected by only one brick wythe. The fact that the corner steel columns have two sides facing the exterior is a particular liability and might be a major contributor to cracking. When corrosion forms on one of the column’s faces, it will lead to the cracking of the perpendicular masonry wall that provides some restraint.

In short, issues of expansion and movement might play a role in the initial crack formation and consequent steel corrosion, but the repair solutions should focus on the corrosion protection of the steel. As further proof, one should consider the cases where

We’re all about superior support.

Figure 8. Corner repair without cut joints – before and after.

Conclusions This article described how some façade systems and repair solutions have evolved over time. In its beginning, the façade repair industry might have used solutions that were insufficient or created unintended consequences (e.g. the creation of joints destroyed the continuity and weakened the attachment of the façade to the structure). This article also illustrates a particular difficulty faced by façade inspectors that often need to assess the effectiveness of work performed in a previous cycle by a different team of contractors and engineers. In

the absence of construction documents and limited to visual data, it may be challenging to differentiate between a retrofit, a repair, or a temporary fix. Looking at a corner that had vertical joints created on both sides, how can one determine if the new corner stack was anchored to the steel column?

Every façade has its particularities, but, in general, a façade is the result of the fashion during the period of construction, the ability of engineers to detail them, and contractors to execute them appropriately. The present condition of the façade also reflects the effect of weathering, and past repair and maintenance efforts. The identification of the façade’s unique conditions and the root causes of its distress is the main responsibility of the inspector. As the façade inspection/repair profession has progressed, today one can expect a specialist in this field to be knowledgeable of the different historical systems and be capable of properly selecting appropriate repair recommendations from a variety of solutions.

There has been a steady increase in the number of published articles related to façade issues, but such piecemeal dissemination of knowledge is still far from satisfactory, especially for those entering the profession. To contribute to the development of the façade inspection specialty, the NYC Department of Buildings has published, online, Façade Conditions – An Illustrated Glossary of Visual Symptoms. This is the result of the collaboration of a group of respected local practitioners.

Professionals that consider entering this field need more than a couple of technical papers or presentations – they need college courses with specialized tracks.▪

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July 2017

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improper rust-proofing led to new cracks forming, despite the introduction of joints.

By 2000, the solution of joints on both sides of the corner of transitional brick walls had run its course; corner repairs involving a single joint or no joint at all were being implemented. These solutions respect the original architectural intent but risk the potential cracking created by the pressure exerted by the new brick’s tendency to expand and press on the old bricks. One can avoid such cracks by using soft/low strength mortar and sequencing the toothing between the old and new bricks.

In 2001, the author designed the repair of a 1940 transitional façade that displayed severe cracks, especially at its exterior corners. The design involved removing sufficient bricks to allow the cleaning and rustproofing of the corner steel columns. The removed bricks were replaced with new bricks bonded to the remaining bricks. After 15 years, no cracks can be observed around the corner (Figure 8).

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This indispensable publication provides the designer with comprehensive and efficient procedures for the safe design of both architectural and structural precast and prestressed concrete products. The design information and recommendations contained in the eighth edition of the PCI Design Handbook are based not only on the latest research but also on a consensus of engineers in practice. Every designer needs to have this indispensable publication at their fingertips for quick reference.

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E i g h t h Edition

S

tructural engineers are primarily concerned with the design and assessment of load bearing structures. Of paramount interest is ensuring sufficient structural capacity (strength) to withstand the loads (demand). It is also important, in an increasingly energy-conscious world, not to waste material unnecessarily. In the design of reinforced concrete (RC) slabs, the engineer may use limit analysis to assess the flexural strength of a design but, until recently, this has involved a laborious hand calculation using either the yield line technique or the strip method. Modern limit analysis tools that automate these approaches are now available and were presented in the September 2015 edition of STRUCTURE magazine (Modern Limit Analysis Tools for Reinforced Concrete Slabs, Ramsay et.al). Regarding the assessment of RC slabs, the two approaches are equally efficient at capturing accurate collapse loads. However, when it comes to design, the lower bound approach offered through the equilibrium finite element (EFE) software provides a distinct advantage in that, in addition to providing a safe prediction of the collapse load, the load path taking the loads to the supports is available through principal moment trajectories. With a knowledge of the load path, the engineer can tailor the reinforcement layout and size to minimize material usage. This article presents a simple, but common example where using this approach reduced the reinforcement requirement by 50%.

on the FE formulation adopted, as indicated in the Table where pure FE formulations, i.e., ones that satisfy the constitutive relations exactly and only weaken one of the other essential conditions, are considered.

Commercial FE systems usually use a conforming (CFE) formulation that weakens equilibrium conditions at the expense of strong compatibility. This runs counter to what the practicing engineer requires, as stated by Edward Wilson: Equilibrium is Essential – Compatibility is Optional. http://bit.ly/2qJFWrN Equilibrium is essential because, if the FE stresses are not in strong equilibrium with the applied loads, the engineer cannot be certain that his/ her reinforcement is sufficient to withstand the applied loads. This statement is derived from the lower bound theorem of plasticity, which is nicely expressed as: The only reason why structural designers sleep soundly is the second [lower bound] theorem of plas-

Technology information and updates on the impact of technology on structural engineering

Equilibrium Finite Elements for RC Slab Design

The Importance of Equilibrium In the design or assessment of simple structural members, the practicing engineer can utilize an extensive library of strength of material solutions offering known theoretically exact solutions to problems where the geometry, material, boundary, and loading conditions are simple. With more complex structural forms, the theoretical solution is unknown, and the engineer must use numerical simulation techniques, such as the finite element (FE) method, to achieve an approximation of the solution.

The FE method produces stresses/moments that approximate the theoretical solution, and the solution normally converges with mesh refinement.

The nature of the approximation is dependent

Strength of Materials Conforming (CFE) Equilibrium (EFE)

ticity theory. This theorem says that no matter how I designed my structures, they are safe because everything was in equilibrium, nowhere [were] the stresses … too large, and I used ductile components and joints.

(From the CIE4150 Plastic Analysis of Structures course description, Delft University of Technology, Netherlands) http://bit.ly/2rsvJx3.

With CFE models, the engineer may not appeal to the lower bound theorem unless he/she is sure that the model is sufficiently refined to provide a decent approximation to a strong equilibrium solution. The EFE formulation, on the other hand, does provide solutions that, even for the coarsest mesh, offer the practicing engineer a strong equilibrium solution with which a safe design may be determined. The tapered cantilever problem of Figure 1 (page 32) demonstrates these ideas by comparing sectional stress results from coarse CFE and EFE models.

The CFE elements, which, respectively, can approximate constant and linear stress fields for the four and eight noded elements, recover continuous displacements (and hence strain/displacement compatibility) at the expense of strong equilibrium.

The equilibrium element satisfies equilibrium exactly at the expense of strong compatibility,

Statics

Constitutive

Kinematics

Strong Weak Strong

Strong Strong

Strong Strong

Strong

Weak

Table of conditions satisfied weakly or strongly for ‘pure’ finite element formulations.

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By Angus Ramsay, M.Eng., Ph.D., C.Eng., FIMechE, and Edward Maunder, C.Eng., MA, DIC, Ph.D., FIStructE Angus Ramsay is the owner of Ramsay Maunder Associates, an engineering consultancy based in the UK. He is a member of the NAFEMS Education & Training Working Group and acts as an Independent Technical Editor to the NAFEMS Benchmark Challenge. He can be contacted at angus_ramsay@ramsaymaunder.co.uk.

Edward Maunder is a consultant to Ramsay Maunder Associates and an Honorary Fellow of the University of Exeter in the UK. He is a member of the Academic Qualifications Panel of the Institution of Structural Engineers. He acts a reviewer for several international journals, such as the International Journal for Numerical Methods in Engineering, Computers and Structures, and Engineering Structures. He can be contacted at e.a.w.maunder@exeter.co.uk.

Displacement (mm)

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Figure 1. Tapered cantilever problem (linear elastic).

as illustrated by the discontinuous edge displacements shown for the constant stress field element (p=0). The displacements from the equilibrium element model are, however, rather accurate in an average sense, as shown in Figure 1’s accompanying table (vertical displacement at point A), and are more than adequate for assessment of serviceability conditions such as that of maximum displacement.

The exact stress results on the section are easily determined from statics. The FE stresses for both formulations are discontinuous across the section. For the CFE model, this leads to different stress results on either side of the section, neither of which are in equilibrium with the applied load. With the EFE model, however, even though the stresses may be discontinuous, the sections are equal and opposite, and, most importantly, are in equilibrium with the applied load.

Figure 1 also shows a plot of principal stress trajectories from the EFE model. These form an orthogonal net of lines and provide information about the direction and magnitude (color) of the principal stresses at a point. Such trajectories clearly show the way in which the force is transmitted from the applied loads to the supports. Because equilibrium is always satisfied, they do not change significantly with mesh refinement. Principal stress trajectories may be used to guide strut and tie representations so that the reinforcement is optimized in terms of layout and size.

In addition to linear elastic analysis, EFE may be used to conduct limit analysis to determine

Figure 2. Constant torsional moment field in an RC slab.

the plastic limit load. In contrast to the yield line technique, which is an upper bound technique providing potentially unsafe predictions of the collapse load, EFE provides lower bound solutions which are safe. The constant torsional moment field shown in Figure 2 provides a simple example of these ideas. This moment field would be observed in a square slab supported on three corners and with a point load at the fourth corner.

The principal moments are determined from Mohr’s circle. The principal moment trajectories are shown as red (hogging) and blue (sagging) lines. This problem was adopted by the IStructE and used in the magazine Structural Engineer as an “And Finally …” question in the August 2016 edition.

The constant torsional moment field is transformed into principal moments (principal moment trajectories are shown in Figure 2).

The trajectories cross the slab at 45 degrees, with hogging trajectories shown in red and sagging trajectories in blue. The magnitude of the principal moments is identical everywhere in the slab. Whereas such a slab might be reinforced with an orthogonal mesh placed both in the top and bottom and parallel to the sides of the slab, it is evident from the principal moment trajectories that an optimal form of reinforcement would require it to be placed parallel to the trajectories. Further, it is seen that, rather than reinforcing both the top and bottom in two orthogonal directions, only one direction is truly required. Thus, a knowledge of the moment field within the

(a)

(b)

Figure 3. Slab configuration and reinforcement layouts; a) Initial layout, b) ‘Optimized’ layout.

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slab leads to a potential reduction in reinforcement steel of 50%!

Principal moment trajectories provide a total picture of the statics within the slab from which an optimum form of reinforcement may be easily established. This is considered, in the example problem below, for a more common slab configuration where the moment fields are more complex.

A Slab Design Problem The reinforcement for the 2- by 1-meter rectangular slab, simply supported on two adjacent sides and with a 25kPa uniform load, as shown in Figure 3, is to be designed. The approach adopted will be design-by-assessment; i.e., a reinforcement layout will be assumed, an assessment performed, and simple scaling of reinforcement size applied to ensure that it can safely carry the load.

EFE will be used for the design but, since this slab possesses no known theoretical solution, the yield line technique will first be used to establish an upper bound to the collapse load. This will be used to verify the EFE solution.

While highly efficient commercial software is now available for yield line analysis, a more conventional approach will be demonstrated in this example. This is because it mimics the more traditional hand approach where likely collapse mechanisms are sought and then geometrically optimized to find the lowest upper bound solution.

Yield line analysis based on a finite element analysis of a refined unstructured mesh of triangular elements provides a good, although fuzzy, image of the collapse mechanism, as seen in Figure 4a. This indicates that the collapse mechanism can be idealized as a single sagging (blue) yield line angled across the slab and initiating at the supported corner. A coarse mesh including this mechanism is then constructed and the optimum termination position of the yield line, along the long free edge, is determined

(a)

(b)

(a)

Figure 4. Upper bound (yield line) solutions; a) Refined unstructured mesh (30kPa), b) Coarse optimized mesh (23.61kPa).

by geometric optimization. Since the yield line analysis provides an upper bound approximation, the position giving the lowest collapse load is taken as the optimum position.

The lower bound solution from EFE, based on a relatively coarse mesh, is shown in Figure 5a and a very tight bound on the collapse load is provided, thus verifying the solution when compared to the yield line solution.

If the safe collapse load provided by EFE is taken, then the collapse load is just under the required 25kPa. The reinforcement size would need to be increased slightly to account for this.

Having achieved a safe design that is not overly conservative, the engineer could stop at this point with the knowledge that he/she has done a sound job. If, however, the engineer wants to whittle down the cost of the slab, then further work is required.

The principal moment trajectories of Figure 5a may be interpreted as an optimal reinforcement layout and sizing. It is evident from this diagram

(b)

Figure 5. Lower bound solutions showing principal moment trajectories and collapse loads; a) Initial layout (23.59kPa), b) Optimized layout (23.30kPa).

that, similar to the case of the constant torsional moment configuration, by the simple expedient of rotating the reinforcement layout through 45 degrees and removing the rebars that serve no function, a reduction in reinforcement of 50% might be achieved. This has been confirmed using EFE, and the result is presented in Figure 5b. The principal moment trajectories are virtually unchanged, as is the predicted collapse load which is reduced by a little over 1%.

Practical Conclusions This article has demonstrated how, using an appropriate software tool, e.g. EFE, that offers up a complete and easily understandable graphical representation of the statics within a slab through principal moment trajectories, the practicing engineer can safely make very significant material/cost savings in the design of an RC slab.

This article has not considered the serviceability limit state (SLS) conditions of

deflection and cracking. However, these could be considered separately and the reinforcement requirement to satisfy these conditions added as further, separate layers of reinforcement.

The equilibrium finite element formulation is not a new one, dating back as it does to the early days of finite element research. However, Equilibrium Finite Element Formulations, published by Wiley, March 2017, has resolved many of the numerical issues that were initially an obstacle to the acceptance of the method. http://bit.ly/2qObMyL EFE, as currently formulated, is for the assessment of slabs but it may be used, as demonstrated, for slab design through a design-by-assessment approach. It is possible also to formulate EFE as a true design tool where the reinforcement layout becomes a variable in the process. This formulation of EFE is a future development that is being considered by the authors.▪

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July 2017

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By John E. Sheridan, Ken Maschke, P.E., S.E., LEED AP, and Jared E. Brewe, Ph.D., P.E., S.E.

O

n May 27, 2016, Navy Pier unveiled Centennial Wheel, the much anticipated 525-ton Ferris wheel erected in celebration of the Pier’s 100th year. Both tourists and Chicago natives alike had been anxiously awaiting the reimagined landmark, which stands an impressive 50 feet taller and weighs nearly twice as much as its predecessor.

Planning for Centennial Wheel began almost two and a half years before its 2016 Memorial Day opening. Structural engineering firm Thornton Tomasetti (TT) worked closely with Navy Pier to select a Ferris wheel that not only offered an impressive guest experience but also would fit within the structural limitations of the iconic Pier.

After all, the wheel was to be constructed on the roof of an operating parking garage atop a 100-year-old pier stretching half a mile into Lake Michigan.

Due in large part to the engineering restrictions inherent in this project, Chicago-based James McHugh Construction Co. (McHugh) started planning for the project more than one year prior to the former wheel’s September 2015 demolition. McHugh engaged Simpson Gumpertz & Heger (SGH) to assist in the evaluation of construction picks and crane placements, crucial elements to ensuring the project’s safe and timely completion. SGH also provided shoring and repair design during construction.

Existing Structure Navy Pier was constructed in 1916 on thousands of 12-inch diameter wood piles. In 1992, this portion of Navy Pier was redeveloped with a precast concrete parking structure supported on concrete caissons spaced at 60 feet in the north/south direction, with 30-foot bays in the east/west direction and a full-width expansion joint east of the original wheel. The upper Pier Park level of Navy Pier houses the main public space which includes the Ferris wheel, swings, carousel, and other amusements, as shown in Figure 1.

The structural framing of the upper Pier Park level consists primarily of 7.5-foot-wide by 40-inch-deep precast concrete double-tees supported by precast concrete inverted tees and columns. The lower parking level consists of similar concrete framing. However, the lower level utilizes 12-foot-wide by 26-inch-deep double-tees due to the lower load carrying capacity necessary for car parking. Figure 2 depicts a cross section of the pier with the original Ferris wheel.

In the original design, the superimposed dead load on the park area varied between 120 and 360 pounds-per-square-foot (psf ) for trees and landscaping, with a design live load of 100 psf for the public spaces. A service drive on the north side of the deck, shown above the wheel in Figure 1, accommodated 250 psf.

Ferris Wheel Selection

Figure 1. Pier Park at Navy Pier. Courtesy Google Earth©.

STRUCTURE magazine

Eight legs supported the original Ferris wheel, all of which landed short of the north/south column grid. A 6-foot-tall, topside steel distribution frame atop the precast structure, shown in Figures 1 and 2, was utilized to transfer design loads to the underlying columns.

The new design sought to eliminate this steel framing so riders could board at Pier Park level without traversing unsightly ramps.

Five different Ferris wheels were initially considered. TT reviewed the gravity and lateral loads imposed on the existing structure, as well as how each wheel’s attachment points interfaced with the Pier structure. The support points for each wheel were overlaid on the existing precast structure. Navy Pier sought to minimize the structural depth at the Pier Park level, the impact of a new structure on drive aisles and parking spots in the garage below, and the number of new foundations required. This meant the closer the support legs were to the east/west girder lines, the better.

Ultimately, TT recommended the DW-60 by Vekoma/Dutch Wheels of the Netherlands, designers of the former Ferris wheel. Six legs

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July 2017

Figure 2. East-west pier cross section.

support the new wheel; two parallel legs on each side of the wheel, and two legs that angle east to support lateral loads. Its 120-foot horizontal spread most closely aligned with the existing structural grid.

Structural Design Gravity loads were provided directly by the wheel designer, but TT had to identify code-required environmental loads, namely wind and ice loads. The team met with the City of Chicago plan examiner early on to address code interpretation, which resulted in the use of ASCE 7-05 to better capture the combined effect of wind blowing on members with ice build-up. The basic lateral wind force was approximately 250 kips. The team estimated 1.25-inch thick ice build-up around all of the wheel lattice members – an additional weight of more than 200 tons, 40 percent of the wheel’s self-weight.

The additional sail area from this build-up resulted in an increase in the lateral force of 65 kips in one direction.

A new structural support system was required to transfer load from the wheel’s six legs to the foundations. Gravity support was provided by two new 18-inch-thick concrete walls installed in the parking garage directly below the four main legs of the wheel. Each leg supports approximately 600 kips. These walls encase the existing precast columns and engage the existing caisson foundations directly below. The walls were designed as deep beams, using the strut and tie method, with eight 290 kip-persquare-inch DYWIDAG bars for the main tie force.

The lateral load from wind is resisted by the splayed legs, which are supported by new, reinforced concrete columns located in the parking garage atop micropiles that extend another 120 feet to bedrock. Each micropile can resist 100 kips of pullout. The horizontal in-plane reactions that occur at the base of each leg exceed the existing structure’s diaphragm capacity; therefore, a new 14-inch-thick concrete mat was designed as a tension tie between the splayed legs and main legs. Highstrength tendons stressed to 150 kips were used to pre-compress the tie mat (Figure 3). Since the splayed legs extend over the existing Figure 3. New concrete mat (high-strength expansion joint, a slip detail tendons shown in yellow). was specified to allow the STRUCTURE magazine

mat to slide over the adjacent structure to an independent support on isolated columns.

There was some concern about shrinkage in the mat and placement of the anchor bolts. If the 120-foot mat spread was continuously poured, the concrete shrinkage could have pulled the anchors more than one inch out of position. The wheel manufacturer required a very tight tolerance of five-thirty-seconds of an inch. Therefore, the design and construction team divided the concrete installation into two pours. The first pour was centered around the anchorage locations and the second pour, which infilled between the first, followed after two days of curing to minimize the net movement of the anchor bolts due to shrinkage.

Another challenge emerged during the transfer of load from the wheel legs into the structure, which required the installation of twenty 42-millimeter anchor bolts. Supplemental reinforcing, in accordance with ACI 318 Appendix D – Anchoring to Concrete, was specified at each location to prevent concrete breakout. The team utilized 3-D modeling of reinforcing steel, anchors, and plates in the congested base shoe areas, which drove decisions simplifying rebar layout for constructability.

Construction Before construction of the Centennial Wheel could begin, McHugh was tasked with dismantling the existing wheel. Several rigging contractors were consulted, and Advantage Industrial Systems (AIS) ultimately was selected to join the team that would work alongside Vekoma to complete this unique project. SGH was asked to evaluate the ability of the existing structure to support the applied construction loading. Available original erection drawings, precast shop drawings, and design calculations were reviewed to determine existing structural member strength. Independent calculations confirmed the strength of the existing structural framing.

At the start of construction, SGH performed a condition assessment of the existing Pier structure. Observations of structural components were limited to the underside due to the existing waterproofing membrane. Isolated locations of deterioration were identified around previous deck penetrations, and SGH provided McHugh repair drawings to address the observed double tee flange distress.

The new Centennial Wheel is not only erected on top of Navy Pier, but also on the roof of the west parking garage. McHugh and SGH worked closely to ascertain material storage locations, construction staging, and crane selection for both dismantling the existing wheel and erecting the new wheel. SGH analyzed the existing structure in consideration of the anticipated sequences, material and equipment staging, crane locations, and construction loadings. Components of the existing and new Ferris wheel were transported to and from the site by truck via the service drive on the north end of the Pier

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July 2017

Figure 5. Shoring to complete installation of new structural support.

Figure 4. Crane outrigger reactions and member checks.

Park level. The maximum individual component weight was the new wheel’s axle at 48,000 pounds.

McHugh worked with Royal Crane Services and Imperial Crane Services to optimize equipment selections to facilitate the necessary lifting. Crane outriggers were placed on heavy timber cribbing to distribute the load to the supporting members. The maximum anticipated crane/outrigger load was 70 kips during dismantling and 118 kips during erection. A summary figure from the construction load analysis is shown in Figure 4, taken from the delegated design submittal. The figure shows the three cranes used during installation (red, blue, and green), the various locations as they moved during the installation process, and the maximum anticipated outrigger reactions. While the double-tee framing is similar throughout, several member checks were performed due to the various combinations of uniform superimposed load and outrigger point loading. Preferred crane/outrigger placement was over the columns or inverted tees.

When locating them on those components was not possible, SGH analyzed the load distribution based on the stiffness of the timber cribbing and the supporting double tees. A parametric study determined the distribution of load between the six double stems below the timber cribbing, and shear demands often dictated permissible crane outrigger locations.

Figure 6. Installed shoring from below.

Before erection of the new wheel could commence, McHugh installed the new structural support system which required removing an existing precast concrete inverted-tee beam and replacing it with a concrete wall. Shoring of the double-tee beams on the upper Pier Park level was required to remove the inverted-tees. The dead load of the upper Pier Park level framing is approximately 100 psf; the design live load of the lower parking level framing was 50 psf.

Accordingly, the analysis determined the lower parking level of Navy Pier was inadequate to support the shoring loads from the level above unless engaging approximately twice the tributary width.

SGH developed a shoring solution that consisted of hanging the precast concrete members from steel beams temporarily installed above the upper Pier level, shown in Figures 5 and 6. The new support walls are depicted in Figure 7.

Construction on Centennial Wheel began in late October 2015.

Months before the start of construction, in partnership with Navy Pier and the Pier’s tenants, McHugh selected two consecutive days to close the Pier to erect the largest elements of Centennial Wheel. Five cranes totaling 1.2 million pounds of lifting capacity were used during assembly.

The wheel was erected by bolting two of the gravity support legs to the base shoe, standing the legs up, and forcing a large diameter pin into the top connection (Figure 8). These two legs remained unstable until a lateral support leg was installed. The main picks were accomplished utilizing two 250-ton cranes and one 100-ton crane operating in unison, instead of what otherwise would have been a single 500-ton crane for a more “typical” job. The center axle, the heaviest crane load installed on the Pier, weighed 48 kips, requiring a critical outrigger reaction (Figure 9). The spokes

Figure 7. New Ferris wheel support walls.

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Figure 8. First set of gravity legs just prior to establishing a connection.

and rim of the wheel were incrementally installed in a clockwise motion, and passenger cabs were connected in groups of three on opposite sides of the wheel to maintain balance.

There was nothing typical or standard about the construction of Centennial Wheel. On the first day of Pier closures, the high temperature in Chicago was only 9° F. The following day was only slightly better with a high of 15° F. With most other job sites across the city closed down, the Centennial Wheel construction team worked 16-hour days.

Centennial Wheel opened in May 2016 to record crowds. While visitors may be most impressed by the new temperature-controlled gondolas, in-cabin video screens, and better views, the Chicagobased teams behind the wheel know the real accomplishment is in the engineering details: a 525-ton, 196-foot-high structure constructed over a fully-functional parking garage on a pier by more than 500 men and women in roughly 25,000 hours. Now, that is impressive.▪

Figure 9. Installation of the axle.

Project Team

John E. Sheridan is a Senior Vice President with James McHugh Construction Co. John can be reached at jsheridan@mchughconstruction.com.

Owner: Navy Pier Inc., Chicago, IL Structural Engineers of Record: Thornton Tomasetti, Chicago, IL (Design) Simpson Gumpertz & Heger, Chicago, IL (Construction) Contractors: Primary – James McHugh Construction Co., Chicago, IL Specialty – Advantage Industrial Systems, LLC, Frankfort, IL Architect of Record: James Corner Field Operations, New York, NY / Gensler, Chicago, IL Fabricators: Vekoma Rides Manufacturing BV, The Netherlands

Ken Maschke, P.E., S.E., LEED AP, is a Vice President at Thornton Tomasetti and can be contacted at KMaschke@ThorntonTomasetti.com.

Jared E. Brewe, Ph.D., P.E., S.E., is a Senior Staff II with Simpson Gumpertz & Heger. Jared can be reached at JEBrewe@sgh.com.

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STRUCTURE magazine

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4/28/17 9:21 AM

Revamped Ferry Maintenance Facility Port Bolivar Ferry Maintenance Operations Facility

T

By Percy R. James, P.E., LEED AP BD+C, M.ASCE

Limits o

f Upland

s

he Texas Department of Transportation (TxDOT), Houston Constraints on Design District, has provided ferry service across Galveston Bay on SH 87 between Galveston Island and the Bolivar Peninsula The design team faced many challenges, including environmental, for over 80 years. This free service represents the only means financial, functional, and operational considerations. One of the key of public transportation across the Bay. Every year, millions of people, engineering challenges was to develop the new five-slip layout within automobiles, and cargo are moved via this ferry route. The 2.7-mile trip the defined limits of the property to avoid any encroachment into takes approximately 18 minutes, saving motorists at least an hour of the Port Bolivar-Galveston Ferry waterway route, other nearby vessel commute time on alternate local roads north of the Bay. This transpor- holding areas to the south, and the U.S. Army Corps of Engineers’ tation link is also critical to the residents of Bolivar Peninsula. When a compound that is located to the north. This, in turn, influenced the storm threatens, it serves as the quickest and primary means of evacua- landside space programming needs for utilities, dock equipment, tion through Galveston to the and other machinery that causeway and the mainland. support each work dock.

The ferry service fleet presThe overall design solution ently has six ferries. Each also had to meet the strict ferry has a six-member crew requirements for permitting Floating Mooring and can carry approximately approval, and the strucStructure (typ.)

500 passengers and 70 vehitures had to be designed to Steel Sheetpile Bulkhead cles. During the non-peak withstand the effects of season, at least three veswave action, current, and sels set out from the ferry wind events. landings from any given Secondly, considering side across the waterway on TxDOT’s annual budget 30- to 60-minute schedules, allocation limitations and Workdock or daily. During the peak season, its mandate to have at least Slip (typ.) additional ferries are used at two of the existing work more frequent intervals. Until docks remain fully opera2011, TXDOT operated five tional during construction, ferries and four work docks at the design accommodated its Ferry Maintenance Facility. up to four phases into When a sixth ferry was added other sections of the existto its fleet, and since at least ing older structures that Figure 1. Panoramic view of Maintenance Facility. Courtesy Google Maps©. one ferry operates a 24-hour will be in use during conschedule, TxDOT recognized the need to expand and enhance the struction. After reviewing different iterations for constructing the Facility, including the provision for a fifth work dock. new works while keeping the Facility opened, it was determined In June 2014, TxDOT contracted with Lockwood, Andrews & that this approach would create the least disruption to TxDOT’s Newnam, Inc. to perform engineering inspections and condition ongoing daily ferry operations. assessments of both waterside and uplands elements at the Facility, The addition of the fifth work dock to moor a fifth ferry year-round including the bulkhead system, berthing and mooring structures, the was paramount to TxDOT’s plans. Currently, the ferries are each bilge containment tank, approach ramps, and other ancillary structures. moored against a floating, rectangular dock. Each occupies about Mechanical and electrical equipment, dredging requirements, distressed 200 feet by 60 feet of water space. The presence of these large moorconcrete-paved uplands, and storm water conveyance were also evalu- ing units complicated the redesigned configuration. How could an ated. An aerial view of the site is shown in Figure 1. A Preliminary additional work dock be accommodated in a limited area of water Engineering Report (PER) described the findings and recommendations surface already near capacity in terms of existing infrastructure? for repairs or replacement of various engineering elements and systems Using a combination of creative design concepts, technical analyrequired to restore the overall function, integrity, and serviceability of the ses, and a series of manipulations of the ferries to generate the most Facility. Subsequently, a contract was awarded to Lockwood, Andrews favorable layout, it was concluded that the existing mooring units & Newnam to perform detailed engineering designs and prepare the should be replaced with a monopile system, without floating docks. construction documents for the redesigned Facility.

The ferries will continue to berth end-on. The use of a monopile STRUCTURE magazine

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system created an additional 90 feet of usable water space with the new berth configuration (Figure 2).

This solution will serve three separate functions. Firstly, each monopile will serve as a “guide-in” dolphin as a ferry approaches the work dock, then as a berthing structure when a ferry is brought to rest at any of the work docks, and finally as a mooring structure for securing a ferry while it is in the dock.

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Figure 2. Structural site plan – new workdocks configuration.

of currents, and even the load status (ballasted or loaded, etc.) of the ferries while they are moored. Therefore, reasonable estimates of extreme environmental conditions were used to direct the design of each element or component. However, what about the combined effects of the environmental loads on the ferries and on the structures that restrain them?

Although as much as 21 feet of storm surge was recorded during the Hurricane Ike storm event, the design team assumed a design wave height of 15 feet as a reasonable criterion because the site is considered to be sheltered from the effects of strong currents. Also, large wave forces that are generated from either passing ships or wind-induced wave action, and the presence of the continuous bulkhead wall along the length of

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C02 LKR

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It is difficult to predict or calculate the maximum environmental forces on these types of maritime structures during their service lives given the unpredictability of wind and wave intensities and directions, the presence

Monopile with “Donut” Fenders (typ.)

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Unlike other fields of civil engineering, there are few standards and no established codes that govern the design of port and harbor berthing and mooring structures. Furthermore, it is well established that the design solutions for these types of structures are a function of facility type and location. For example, the redesign of the TxDOT Facility considered factors such as site bathymetry, geotechnical consideration, vessel characteristics, facility expansion constraints, currents, storm surge, hurricane-induced wind and wave loads, berthing and mooring loads, and utility requirements. The berthing and mooring sequences, and other ferry maintenance-related activities, were observed over multiple weeks to gather information to fully understand TxDOT’s operational needs. Engineers also used this information to help identify potential conflicts during construction and for operational safety considerations.

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UHMW Protection Strips (typ.)

El. ‘22’- 0’ MSL

Fender Element w/Nylon Reinforced Elastomer Skin

Ferry Profile

Ferry Profile

Lower Flotation Element w/ Nylon Reinforced Elastomer Skin (typ.)

60’Dia. Steel Pipe Pile (typ.)

Donut Fenders Normal Operations

Donut Fenders Hurricane Operations

Figure 3. Donut fender and monopile detail.

the property, means that the incident waves are reflected. Given the facility’s configuration and the selected waterside structure types, these wave force effects were considered to be of lesser intensity to those that are generated from direct wind and current forces and therefore did not control the structural design.

For maritime structures, wind forces are based on an industry-standard 30-second design wind velocity as recommended by the World Association for Waterborne Transport Infrastructure (PIANC). This time interval is necessary to allow fender compression or the full tension in mooring lines to develop, respectively. For this project, the design wind speed was estimated from the ASCE-7 intensity maps and adjusted using a wind duration correction factor relative to the 3-second gust.

Modified versions of the velocity pressure equation were used to calculate the forces due to currents and wind. The modified equations account for the design ferry’s shape and its side and underwater projected areas, the effect of shielding from other moored ferries and obstructions, and the orientation of the ferry relative to the flow of the fluids. Wind and current loads were computed assuming various angles of attack between 0 degrees on the bow and 180 degrees on the stern of a ferry. The concurrent wind and current forces acting directly on the structures were also determined.

For mooring purposes, two limiting wind speeds from any direction were established: less than 60 mph winds of 30-second gusts during dock operations and a 130 mph survival wind with a 3-second gust, since TxDOT intends to secure the ferries at the work docks during storm events. Therefore, it was important to design the mooring structures to accommodate at least six additional mooring lines per ferry. These extra lines will be used to batten down a ferry at each work dock during such a storm.

Since the ferries have high windage areas relative to their shallow draft, the effect of strong winds on the ferries are more pronounced than that of currents for this locale, which were estimated at between 1 and 2 knots. The combined effects of the wind and currents loads, coupled with various geometric arrangements of the mooring lines, were studied to estimate the critical forces at each mooring point.

The ideal mooring point locations, given the expected conditions and ferry motions, were identified at each monopile and specific STRUCTURE magazine

locations along the bulkhead. The mooring loads are designed to be distributed through mooring hardware called “mooring rings” that are installed at each monopile or “bollards” at the bulkhead.

The mooring rings are attached to the fender framing that is free to move vertically with the change in water levels. In the case of the monopiles, the line loads were further assumed to act at various levels along each monopile’s shaft and at maximum inclined angles of 25 degrees to simulate line tension given the ferry’s draft condition and the water height relative to datum.

In determining the berthing energy that was absorbed by the fender system, the weight of the ferry first had to be defined. The weight or full load displacement is defined as the total weight of a ferry body, its engine, cargo, fuel, passengers, crew, and other items that are carried. The ferry’s berthing energy is a product of its weight and approach speed, and berthing coefficients. The energy requirements were estimated as 162 kip-feet and 335 kip-feet for side-on and end-on berthing maneuvers, respectively. The coefficients are functions of the relative water depth and configuration of the work dock, the performance characteristics of the fenders, and the design environmental loads.

The response of the fender system to the ferry’s impact follows the proverbial “chicken-and-egg” scenario. On the one hand, at the point of impact when a ferry is brought to rest, the ferry’s energy is absorbed through elastic deformation as the selected fender is compressed and through the lateral deflection of the monopile.

For this project, as much as 85 percent of the energy absorbed is through fender compression. This is directly related to the fender reaction force, which in turn had to be limited by a maximum allowable ferry hull pressure. Once the monopile structure was decided on, the selection of the fender type was influenced primarily by its efficiency. This efficiency was based on the lowest reaction-to-energy ratio so that the body height of the fender had to be dimensioned to ensure that sufficient contact area was provided to mitigate against exceeding the hull pressure requirement.

The Waterside Solution The combo guide-on/berthing/mooring monopile system at each work dock will consist of four five-foot diameter steel pipe piles. The pile size was determined from the energy absorption requirement of the fender and the latter’s related minimum size, the soil lateral resistance, 1

3

2

4

5

A

Cantilevered Steel Combi-Wall Grouted Tie-Back Anchor Location

B

C

60’ Dia Steel Pipe Pile (typ.)

D

E

F

Figure 4. Overall piling plan.

40

July 2017

Anchored Sheetpile (typ.)

and the flexural and shear capacity of the monopile to resist the resultant fender reaction. The fenders will each be a nominal nine-foot diameter “donut” that will be placed over the steel piles (Figure 3).

The selected fender has an energy absorption capacity of 198 feet-kips and associated maximum reaction force of 339 kips.

The “donut” consists of a closed cell polyethylene foam core that is protected by a tough, wear-resistant, reinforced polyurethane skin. The foam cell is attached around a steel core that is lined with low-friction bearings. The bearings allow the fender to rotate freely around the monopiles, and to rise and fall with changes in the water level. The piles were designed to be driven to depths of up to 100 feet below the mud line because of the presence of weak upper soil strata that are estimated to provide low to moderate lateral pile support.

of AZ-19 steel sheet piling. In this solution, the lateral earth pressures are transferred to the bulkhead through arching action of the soil. The use of this bulkhead type eliminates the use of any anchor system that would otherwise have to be installed on the Army Corp’s property.

Protecting the Uplands Infrastructure

Percy R. James, P.E., LEED AP BD+C, M.ASCE, is the Structural Engineering Group Leader at Lockwood, Andrews & Newnam, Inc. He can be reached at PRJames@lan-inc.com.

Conclusion These engineering solutions mean that TxDOT is well on its way to improving its maintenance capabilities and operations. The carefully planned construction phasing will allow TxDOT’s maintenance activities to continue with few disruptions.

Construction started in April 2017 and is expected to take eighteen months.▪

There is evidence of overstressing and some cases of localized failure in the existing bulkhead along the eastern side of the property. To mitigate against further failure along the existing east bulkhead, and to protect the existing uplands infrastructure, a new three-part steel bulkhead was designed, as shown in Figure 4. The first and second parts are conventional anchored steel sheet pile bulkheads with different types of anchor systems. The tie-rod and concrete deadman anchor system were designed for 70% of the roughly 475 feet of east bulkhead. The remaining 30% of bulkhead will be restrained with a grouted tie-back anchor system.

The third bulkhead option was the design of a high strength, cantilevered steel combi-wall with high stiffness properties (large bending resistance). This type of bulkhead structure consists of a series of fourfoot diameter steel pipe piles that are integrated with intermediate pairs

Project Team Owner: Texas Department of Transportation, TxDOT, Houston District Structural Engineer: Lockwood, Andrews & Newnam, Inc. (LAN), Houston, TX Civil and MEP Engineers: LAN, Houston, TX Geotechnical Engineer: Tolunay-Wong Engineers, Inc., Houston, TX General Contractor: Texas Gulf Construction Company, Inc., Galveston, TX

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48 Helical piles proved to be the most suitable and cost-effective piling option.

Faneuil Hall MARKETPLACE

S

By Rimas Veitas, P.E., Derek Simpson, P.E., and Michael Cronenberger, P.E.

ituated near Boston’s original coastline and developed in the 17th and 18th centuries to increase waterfront real estate, Faneuil Hall Marketplace is one of Boston’s central meeting places. This popular location provides both residents and tourists a unique urban marketplace experience. The Marketplace is separated into four locations: Faneuil Hall, Quincy Market, North Market, and South Market.

Over the last decade, revitalization projects within the Marketplace continue, with the most recent being the replacement of a 1970s structure – a well-known large flower kiosk – with a new, one-story retail glass pavilion adjacent to Quincy Market.

Subsurface soil conditions at the site consisted of 9 to 14 feet of unsuitable urban fill (soil mixed with miscellaneous man-made debris) over up to 5 feet of soft organic silt, over thick natural marine deposits (clay, silt, sand), overlying glacial till. The upper portion of the marine layer was relatively stiff/dense and became softer with depth. Glacial till was approximately 60 to 70 feet deep. Remnants of previous structures, such as buried timber wharfs, were present in the fill to further complicate the already challenging subsurface conditions. Groundwater was encountered at 13 to 14 feet below the ground surface.

Geotechnical Design/Build Solution

Project Challenges The project team encountered a number of project challenges including construction in a sensitive, historic area, limited construction access in the dense urban environment, and difficult subsurface soil conditions that are commonly found in Boston’s reclaimed waterfront neighborhoods.

Structural column loads on the project site ranged from about 30 to 60 kips; however, the site provided many challenges.

STRUCTURE magazine

Excavation, disposal, and replacement of the unsuitable fill and organic layers was deemed impractical due to premium costs associated with off-site soil disposal, excavation dewatering, and importing large quantities of structural fill.

The project team explored several piling options in lieu of excavation/replacement, including driven timber piles, drilled micropiles, drilled shafts, ductile iron piles, and helical piles. Driven timber piles were economically viable but were eliminated from consideration

43

July 2017

due to access issues and noise/vibration concerns. Drilled micropiles as all 48 piles were successfully installed at their planned locations. and drilled shafts both offered low-noise and low-vibration solutions Upon completion, all helical piles were cut to their specified elevation, but were too expensive. Ductile iron piles and helical piles were and the interior of the pipe shaft was filled with neat cement grout to both appealing options due provide additional corrosion to ease-of-access, low vibraprotection and pile stifftion, and relatively low cost. ness. All helical piles were The team ultimately selected installed in 8 days, which helical piles as the most included mobilization and suitable and cost-effective load testing. piling option, as it offered Originally, the design of the low-noise and low-vibration helical piles was to achieve where the Ductile Iron Piles the required torsional resiscould not offer low-noise. tance and capacity within Helical piles are deep the upper stiff layers of the foundation elements that marine clays. It is fairly are used to support new common practice in the foundations or underpin Boston marine clay deposexisting foundations. They its to attempt to achieve the generate no vibrations and capacity within the upper can be installed with only stiff layers. During some ini6 feet of overhead space tial probing, it was found and in other limited-access Helical piles support new glass building in Boston. that the marine clays would situations. The pile shafts not allow for the helical pile are made of galvanized steel and are installed in short sections, each to achieve the required 30 kips of pile capacity. In actuality, installers about 5 to 7 feet long. Each pile consists of a lead helical section with were only able to achieve about half the required capacity. Therefore, it welded screw-like helix bearing plates; subsequent straight-shaft sec- was determined that all helical piles should fully penetrate the marine tions are mechanically fastened to the lead section as it is advanced clays and terminate within the glacial till deposit where achieving into the ground. the torque and capacity was not an issue. This was deemed to be the The piles are installed with a skid-steer or an excavator equipped most economical solution to avoid costly redesign of the foundation with a high-power torque head, which is calibrated to correlate torque system to account for additional piles. resistance with axial pile capacity directly. Helical piles can also be installed with hand-held torque motors for locations that are not Quality Assurance and Control accessible with a skid-steer or small excavator.

Helical piles can function as end-bearing or side-friction elements. Pile production included a full-time Quality Control person to overFor an end-bearing pile, the lead section is advanced through the see pile testing and installation. A full-scale compression load test unsuitable soil layers and into an underlying bearing stratum until a was performed on a sacrificial test pile that was loaded to 200% of predetermined design torque value is achieved. For a side-friction pile, the design capacity. The test results showed deflection of less than “digger plates” are added between each pile section to create annular ½-inch at design capacity and less than ½-inch of net deflection space around the steel shaft, and the annulus is filled with grout as the upon completion of the test. During load testing on the sacrificial pile is advanced into the ground. This process creates a grouted bond pile, the interior was not grouted as with production piles. This was with the surrounding soil, resulting in a helical micropile. Similar to done to reuse the test pile at a production location. Therefore, the a Drilled Micropile, a side friction helical micropile is installed to a test pile did not get any benefits from the additional shaft stiffness predetermined design depth. At this point, the helices attached to the through the softer marine clay and organic zones. As a result, no lead section aid in advancing the pile into the ground. However, due shaft buckling was observed during the test and may suggest that as to a strain compatibility between a grout-to-ground bond and an end long as there is soil, even at a weak density, it can provide sufficient bearing plate, the torsional resistance should be ignored and only the confinement of the piles. bond length of the friction portion of the pile should be considered.

In this case, helical piles offered some key advantages, The Pavilion’s final structural design required 48 helical piles with including low noise during installation, eliminated the an allowable compressive capacity of 30 kips each. The final pile need to export/import large quantities of fill, and did not design was performed by Helical Drilling and featured a galvanized require site dewatering.▪ 80 ksi steel pipe section manufactured by The Ideal Group. The Rimas Veitas, P.E., is the Founder and CEO of Helical Drilling. piles consisted of a 27⁄8-inch-diameter, 0.276-inch-thick shaft with Since 1989, he has provided leadership that has differentiated the quadruple-helix (8-inch/10-inch/12-inch/14-inch) lead sections. The company as a ground improvement leader and innovator in the piles were designed to derive end-bearing capacity in the glacial till Northeast. He can be reached at rimas@helicaldrilling.com. layer below the fill, organic silt, and marine layers.

Derek Simpson, P.E., currently serves as Project and Business Development Manager at Helical Drilling. He can be reached at dsimpson@helicaldrilling.com.

Pile Installation Michael Cronenberger, P.E., currently serves as the Manager of Before the installation of production piles, the General Contractor prethe Specialty Geotechnical Construction Department at Helical excavated pile locations to remove potential obstructions, including Drilling. He can be reached at mcronenberger@helicaldrilling.com. timbers and granite blocks. Pre-excavation proved to be worthwhile, STRUCTURE magazine

44

July 2017

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A

The new building at Common Ground High School connects to the original campus via a mass timber footbridge.

Mass Timber as Structure and Finish Common Ground High School in New Haven, Connecticut

By Alan Organschi

The school’s second-floor lobby highlights the use of cross-laminated timber as both a finish and structural material.

STRUCTURE magazine

46

t Common Ground High School, environmental management is both a subject for study and a daily activity. This small public charter school in New Haven, Connecticut, serves 225 high schoolers and, through its umbrella organization, The New Haven Ecology Project, engages an additional 12,000 children and adults from New Haven’s inner city and metropolitan communities in a range of ecologicallyfocused campus activities throughout the year.

A building committee comprised of Common Ground faculty, administrators, and students commissioned a design team from Gray Organschi Architecture to collaborate with their school to develop a new art studio, science classrooms, and a community meeting and recreation facility that would border the educational farm complex that lies beneath New Haven’s West Rock Park.

They challenged the designers to create a building that embodied the ideas of sustainability, self-sufficiency, and environmental health so critical to the school’s pedagogical mission. In response, a design that utilized renewable resources in the building’s structural assemblies and architectural surfaces was developed.

Among an array of sustainable technologies at work in the Common Ground building, the implementation of emerging “mass timber” structural components is perhaps the school’s most novel feature. Designed to the 2005 Connecticut Building Code as a Type V-B fully sprinklered building, the new 14,600-square-foot addition to Common Ground’s campus features an innovative mass timber structure with a prefabricated system of wall and roof components developed by Gray Organschi Architecture and structural engineer Chris Carbone of the Bensonwood fabrication team.

Glued-laminated timber beams and cross-laminated timber (CLT) panels form the building’s structure and, in large part, its interior finish. Five-ply CLT makes up the central bearing walls and elevator shaft, working in conjunction with lumber shear walls to resist lateral loading, and line the large open stairway at the center of the building. Three-ply black spruce CLT provides the tension surface in cellulose-insulated stressed-skin panels that span the school’s classrooms and circulation spaces.

Glued-laminated Southern Pine beams aggregate to form a 90-foot bridge connecting the new building to the existing campus. Sixty-six-foot-long glued-laminated black spruce timber ridge trusses – assembled with customfabricated steel connections – support glued-laminated roof rafters which in turn span the large multipurpose room that functions as a gymnasium, theater, and community meeting space.

The use of these mass timber assemblies required careful attention to the lateral loading requirements, with the new building designed to a seismic R factor of 6½. The glued-laminated bridge is a buckling restrained frame with non-moment resisting beam-column connections. For the building, hurricane wind loads governed over seismic forces and are resisted primarily with the lumber shear walls sheathed with wood structural panels and assisted by the CLT bearing walls.

Although the project is located in a Wind Exposure Category B zone inland from the Connecticut coastline

July 2017

and is sheltered by a mountain ridge, the relatively light weightto-area ratio of mass timber assemblies dictated that wind load – rather than seismic load – ultimately governed the design of the lateral system. To improve the CLT’s ductility for lateral loads, ½ lap joints and screws connect the CLT panels used in the wall and roof applications. Within the insulated stressed-skin panels spanning the classrooms, individual North-facing clerestory windows provide consistent box beam panels resist indirect daylight for the school’s classrooms and lateral loading through public spaces. connections between the ZIP panel top skins and intermediary I-joists, which are in turn fastened to a lower skin of 3-ply black spruce CLT which is exposed as the ceiling finish.

In addition to the CLT stressed-skin roof panels, the arts and sciences building at Common Ground High School is comprised of an array of commercially distributed structural biomass, harvested wood products manufactured from wood lamellae, fiber, flakes, pulp, and furnish, then industrially agglomerated, adhered, and compressed through lamination or extrusion into sheets, boards, batts and other structural shapes.

Laminated Southern Pine planks, pressure-treated for durability, form the primary path between the upper and lower campus. Strands, veneers, strips, and fibers from aspens, poplars, and longleaf pines comprise the engineered joists, studs, and panels that make up the hollow prefabricated wall and roof assemblies which are in turn densely-packed with recycled cellulose pulp insulation. In addition to the usual array of hardwoods, in this case, birch and maple, which serve as trim and casework in a building of this type, wood strandbased acoustical panels line the walls of classrooms and roof of the multipurpose room to mitigate sound reflection.

Of the various classes of construction materials distributed throughout the arts and sciences building at Common Ground High School, 279 metric tons (308 U.S. tons) is structural biomass. Its sequestration of atmospheric carbon – 447 metric tons of CO2 (493 U.S. tons) – is the equivalent of the annual emissions of about 100 cars, banked instead in the physical structure of the building for the duration of its lifetime.

Visible in the public spaces and classrooms, this catalog of mass timber elements and surfaces serves as a reminder to the faculty and students who use the spaces of the regional forest landscapes 600 miles to the north that generated that material. The raw material from which most of this structure was assembled was harvested in a 60-year rotation of patch harvesting and replanting within a fivemillion-acre area of black spruce forests in Quebec. Processed into structural elements at the nearby Nordic Structures production facility in Chibougamau, it was then shipped to Bensonwood’s Walpole, New Hampshire, fabrication plant, where it was assembled into a structural system of structural elements and insulated wall and roof components. Those components were trucked south, where they were installed as the shell and structure of the new school in New Haven.

This integrated use of renewable material and low-impact construction technique enhances the health and ecological function of the immediate site. It also protects more distant productive landscapes, optimizing their biological and hydrologic processes so that they may continue to provide valuable environmental services such as clean air and water (as well as a steady supply of renewable building material) to the inhabitants of our planet, and, more immediately in this case, to important emerging institutions like Common Ground High School and its forwardthinking students, teachers, and administrators.▪ Alan Organschi is a Partner at Gray Organschi Architecture.

He also the founder of JIG Design Build, which engages in the research, prototyping, fabrication, and installation of building components and systems, and a faculty member at the Yale School of Architecture. With support from the Hines Fund for Advanced Sustainability Research in Architecture at Yale, he directs the interdisciplinary research initiative Timber City, which examines potentially regenerative supply chains linking sustainable forest management and dense urban development of housing and infrastructure in wood.

Project Team Owner: Common Ground High School Structural Engineer of Record: Bensonwood Structures and Christopher Carbone (Timber Superstructure), Edward Stanley Engineers (Foundations) Architect of Record: Gray Organschi Architecture Construction Manager: Newfield Construction Fabricators: Bensonwood, JIG Design Build, AGA Architectural Millwork Integral Team Members: Atelier 10 (Environmental Engineer), Mark Papa (Landscape Architect), Altieri Sebor Wieber (MEPFP Engineer), Godfrey Hoffman Associates (Civil Engineer), Gray Design (Interior Design) Structural Software Used: Cadwork (3D Model), Visual Analysis by IES (Structural Analysis)

Gray Organschi Architecture designed the building’s faceted stressed-skin timber roof assembly to maximize indirect daylight within the building and to provide optimized surfaces for photovoltaic panels.

STRUCTURE magazine

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July 2017

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core requirements and lifelong learning for structural engineers

Education issuEs

Higher Education That Includes Timber Engineering

By Uchenna Okoye, P.E., LEED AP, Michelle Kam-Biron, P.E., S.E., SECB, Brent Perkins, P.E., S.E., and Craig Barnes, P.E., SECB

T

imber is one of the most widely used construction materials in the U.S., especially for low- to mid-rise residential structures. It is also one of the most sustainable materials available. The life cycle of the product involves the sequestration (the physical storage) of carbon, the production of oxygen, and the reduction of energy consumed in creating manufactured products. Even considering these significant benefits and extensive use, surveys conducted by the National Council of Structural Engineers Associations (NCSEA) seem to indicate that many structural engineers graduating from accredited university programs have, in general, not been provided with even a basic understanding of timber engineering.

In a 2004 article, Wood Use in Nonresidential Buildings: Opportunities and Barriers, most designers surveyed had learned little or nothing about wood in school, and many felt the lack of post-professional education was also an issue. The most recent NCSEA Basic Education Committee national survey of engineering schools showed that only fifty-five percent (55%) of institutions offer courses that teach the fundamentals of timber engineering to either their undergraduate or graduate students. Importantly, most of these schools do not actually require their students to take a timber course to graduate.

In years past, this dearth of knowledge prompted some response from industry organizations. In 2007, the Wood Products Council launched the WoodWorks program which provides free project assistance, continuing education, and resources related to the design of non-residential and

multi-family wood buildings. WoodWorks partnered with the California State Polytechnic University Pomona in 2008 to create the Wood Education Institute (WEI) program, which was a virtual learning model intended to assist in offering wood education for undergraduate, graduate, and continuing education programs nationwide. However, due to a lack of financial support, WEI is no longer active.

Additionally, the American Wood Council (AWC) provides continuing education on building Southern California mid-rise light frame construction. codes and the organization’s standards to design professionals, required resources. This has led to workmostly through in-person training and free ing toward a split-semester course in both web-based learning. And, although AWC timber and masonry. The objective is to resources for educators include free AWC provide basic knowledge that will allow standards and students receive discounts on the student to either self-teach to a greater AWC standards used in timber engineering concentration in either topic or to recogcourses, there is still a void that needs to nize that greater academic concentration be filled for almost half of universities that is required. have an engineering program.

Currently, the NCSEA and AWC together Generally, educators recognize the need are devising new ways to address this issue. for timber programs but are under pressure By working with universities, structural to move students through the program as engineering organizations, and trade orgaquickly as possible and to do so in a way nizations, the goal is to demonstrate the that controls tuition costs and minimizes need for students to have quality timber

Percent of Engineering Schools that Offer the Indicated Recommended Course 100

100 100 90 80 70 60 50 40 30 20 10 0

100

98

84 63

60

Reasons Why Timber Design is Not Offered Reasons Why Timber Design is Not Offered

61

66

55

76

25%

19% 19%

25%

Lack of Student Demand

Lack of Student Demand

40

14%

14%

14%

17%

STRUCTURE magazine

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July 2017

17%

11%

11%

14%

Lack of School Support

Lack of School Support

Lack of Timber Research Funding

Lack of Timber Research Funding Lack of Timber Design Professors

LackImposed of Timber Design Professors Unit Restriction Imposed Other Unit Restriction Other

engineering education. Work is also underway to identify methodologies to effectively offer and teach timber engineering nationwide.

Some initiatives include: • Provide students a resource for locating schools and universities that can satisfy timber engineering requirements, • Work with the NCSEA Structural Engineers Certification Board (SECB) to encourage those engineers seeking certification to have timber engineering in their background and to require students seeking SECB Education Certification to have some timber education, • Provide further evidence of the growing need for timber education as has been revealed in a recent Structural Engineering Practitioner survey that showed that 95% of respondents felt their new hires should have timber engineering instruction, and • Work with practitioners and educators to determine the content of a rigorous timber engineering program, and expected outcomes, via a Timber Symposium held at the ASCE/SEI Structures Congress in April, 2017.

This last initiative is a unique partnership between ASCE/SEI Wood Education Committee, NCSEA Basic Education Committee, and AWC, with an ultimate goal to develop a strategic plan for getting timber engineering courses into universities.

Although these initiatives are ongoing, some wood industry educational efforts have continued to target only design professionals. However, the best opportunity to engage engineers is still at the collegiate level. Manufactured timber processes have

continued to get better every year and yield some amazing results. The advent and use of cross laminated timber and mass timber construction are just an example of how far timber construction has evolved. Unless a strong effort is put forth now to create comprehensive timber engineering programs at a majority of universities with engineering programs, the profession will continue to fall behind in having young engineers who can effectively design with timber. This would only hurt the entire AEC industry.▪

Uchenna Okoye is a Project Manager and Lean Integrator with Skanska USA Building, Oakland, CA. He serves as the education and training chair of the Lean Construction Institute Northern California community of practice. He has severed as Co-Chair and continues to be an active member of the NCSEA Basic Education Committee. Uchenna can be reached at uchenna.okoye@skanska.com.

Michelle Kam-Biron is Senior Director of Education for the American Wood Council (AWC).

She is a certified Earthquake Disaster Assessment volunteer and a member of the International Code Council. She also volunteers her time on the NCSEA Basic Education and CALBO Structural Safety Committees, is Chair of ASCE-SEI Wood Education Committee, SEAOC Director, and an SEAOSC Past-President. Michele can be reached at mkambiron@awc.org.

Brent Perkins is a Project Engineer with Dudley Williams and Associates, P.A. in Wichita, KS. He also volunteers his time on the NCSEA Basic Education Committee. Brent can be reached at bperkins@dwase.com.

Craig Barnes is the Founding Principal of CBI Consulting Inc. Craig can be reached at cbarnes@cbiconsultinginc.com.

DESIGNED not to be seen

Micropile

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LegaL PersPectives

discussion of legal issues of interest to structural engineers

Key Concerns in Consent to Assignments

By Gail S. Kelley, P.E., Esq.

C

onsent to assignments, often referred to in the banking industry as “will-serve letters,” come in many different forms.

However, at its most basic, an engineer’s Consent to Assignment is an agreement under which the engineer acknowledges that the design agreement has been assigned to the owner’s lender. The engineer also agrees that if the owner defaults on its construction loan, the lender can exercise its rights under the assignment and require the engineer to provide its services for the benefit of the lender.

The article An Overview of Consent to Assignments, (STRUCTURE, June 2017), discussed the general format of consent agreements. It discussed one of the key concerns with respect to these agreements – whether the lender is obligated to pay outstanding amounts due to the engineer.

This article discusses two other concerns – the lender’s right to use the plans and specifications and the engineer’s obligation to provide certifications or other information to the lender.

owner defaults on its construction loan, the engineer will have completed the plans and specifications and is only providing construction administration. The entity that takes over the project may prefer to have a different engineer perform these services. As a result, some version of the following clause is found in almost all consent agreements: Lender and Lender’s Successors shall be entitled to use the Plans and Specifications prepared by Engineer for the completion of the contemplated improvements without further cost to Lender or Lender’s Successors.

Under the AIA and EJCDC standard form agreements, such as AIA B101 and EJCDC E500, the owner only receives a license to use the Instruments of Service; the engineer retains the copyright. However, owners often edit this wording so that the owner receives all rights in the Instruments of Service, including the copyrights.

Regardless of whether the design agreement states that the owner has received a license for the Instruments of Service or ownership, the above clause in the Consent Agreement should be edited as follows: Lender’s Use of the ...without further cost to Lender or Lender’s Successors, other than payment of all Engineer’s Work Product sums owed to Engineer by Borrower.

Consent to assignments typically indicate Some design agreements stipulate that if the that the lender has the right to require owner terminates the design agreement for the engineer to continue providing ser- convenience or the engineer terminates the vices if requested to do so by the lender. agreement because of a prolonged suspenThey generally also give the lender the sion, the owner may use the Instruments right to use the plans and specifications of Services to complete the Project upon (the “Instruments of Service”) without payment of a licensing fee (see for example hiring the engineer. In most cases, if the § 11.9 of AIA B101). It is a good idea to specifically reference this fee in the Consent Agreement if the Demos at www.struware.com engineer wants the right to recover this Wind, Seismic, Snow, etc. Struware’s Code Search program calculates these and other loadings for all codes based on the IBC or ASCE7 in just minutes (see online fee from the lender, video). Also calculates wind loads on rooftop equipment, signs, walls, chimneys, An example of such trussed towers, tanks and more. ($195.00). wording might be: CMU or Tilt-up Concrete Walls Analyze solid walls for out of plane loading and ...other than payment panel legs next to or between openings by automatically calculating loads to the wall of all sums owed to leg from vertical and horizontal loads at the opening. ($75.00 ea) Engineer by Borrower Floor Vibration Program to analyze floors with steel beams and/or steel joist. and the licensing fee Compare up to 4 systems side by side ($75.00). stipulated in Article Concrete beam/slab Program to provide bending, shear and/or torsional reinforcing.

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Indemnification for the Use of the Engineer’s Work Product In addition to conditioning the lender’s use of the Instruments of Services on payment of all amounts owed to the engineer, the engineer should require the lender to indemnify the engineer from claims arising from their misuse.

Typical wording for this requirement is: Lender agrees to indemnify, defend, and hold the Engineer harmless from any claims arising from changes made to the Instruments of Service by others or use of the Instruments of Service for any purpose other than the purpose they were prepared for under the Design Agreement.

Lender’s Requirements for Certifications and Information Virtually all consent agreements include some requirement for certification of the engineer’s work. Many also require the engineer to provide information to the lender. These requirements range from the completely reasonable to the very unreasonable. An example of a reasonable requirement is: Engineer certifies that Engineer’s statements in this letter have been made, and Engineer’s services have been performed, in accordance with the standards of care of Engineer’s profession for building projects of the similar scope and quality.

The engineer is simply being required to certify that it has complied with the standard of care for its profession, which is the standard of care under the common law.

An example of an unreasonable requirement is: Engineer, at no cost to Lender, shall furnish to Lender upon written request any information Engineer may have regarding the Project, including: (a) information regarding defects in workmanship or materials provided for the Project; (b) Engineer’s estimate of the stage of completion of construction of the Project; (c) any known deviations or variations in construction of the Project from the Plans and Specifications; (d) information regarding any defaults by Borrower, Contractor or any Subcontractor under any of the Construction Contracts; (e) information regarding construction practices or conditions in effect at the Project which Engineer regards as unsafe or

should never sign a Consent Agreement unless it has read the agreement carefully. If there are any terms the engineer does not understand, it should ask its legal counsel to review the agreement.▪ Disclaimer: The information in this article is for educational purposes only and is not legal advice. Readers should not act or refrain from acting based on this article without seeking appropriate legal or other professional advice as to their particular circumstances.

Gail S. Kelley is a LEED AP as well as a professional engineer and licensed attorney in Maryland and the District of Columbia. Her practice focuses on reviewing and negotiating design agreements for architects and engineers.

She is the author of “Construction Law: An Introduction for Engineers, Architects, and Contractors,” published by Wiley & Sons. Ms. Kelley can be reached at Gail.Kelley.Esq@ gmail.com.

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Conclusion Owners seldom read consent agreements; they typically pass them on to the engineer with the explanation that it is a form required by the lender. Often, the owner will indicate that a delay in returning the agreement could hold up the closing on the construction loan. As evidenced by the clauses cited above, however, a Consent Agreement can impose significant obligations on the engineer.

It can also affect the engineer’s right to payment for its services. An engineer

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dangerous; (f ) claims of nonpayment by any person in connection with the construction of the Project; and (g) evidence of full payment to Engineer with respect to any work performed

by Engineer in connection with the Project.

As written, this provision is obligating the engineer to provide information to the lender even before any default on the construction loan. In other words, the engineer would be agreeing to act as the lender’s agent, without compensation. The engineer would also be agreeing to provide the lender with information that may have no relationship to its obligations under the design agreement. If the engineer, in good faith, provides the information required and the information subsequently turns out to be false, the engineer could be liable for claims of negligent misrepresentation or tortuous interference with contract. The engineer should generally only agree to provide information with respect to its own contract. Thus, the above clause should be edited to read: Engineer, at no cost to Lender, shall furnish to Lender upon written request evidence of full payment to Engineer with respect to any work performed

by Engineer in connection with the Project.

Design agreements often require the engineer to provide “such certifications as reasonably required by the Lender.”

To clarify this obligation, the engineer can add the following provision to the Consent Agreement: The Engineer shall not be required to execute certificates that would require knowledge, services or responsibilities beyond the scope of the Design Agreement. Any such certificate will state that it is based on the best of the Engineer’s knowledge, information, and belief.

Historic structures

significant structures of the past

Lake Champlain Bridge

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng., P.E., P.L.S.

L

ake Champlain is 125 miles long and separates New York and Vermont for much of its length. Discovered

by Samuel de Champlain in 1609, the lake at its widest is 14 miles. Between Chimney Point, Vermont, and Crown Point, New York, it narrows down to about 2,000 feet. For many years, the only way to cross the lake was by ferry. Vermont was the early initiator for a bridge starting in 1923, and New York joined up in 1925. In 1926, borings were made at six possible crossing sites, funded by the federal government. In that same year, the Lake Champlain Bridge Joint Legislative Committee (New York State) requested the State Engineer and Surveyor to prepare a preliminary plan for a bridge at the site. On December 15th, the State Engineer recommended a tied arch on two cantilever arms making a total span of 320 feet, with 200-foot flanking spans. On May 11, 1927, the Lake Champlain Bridge Commission was formed with both states appropriating $200,000 for studies and designs.

J. A. L. Waddell was called in to check three of the six possible bridge crossings and to make a recommendation to the Commission.

He selected the Crown Point site and estimated a bridge could be built for $920,000.

For some reason, Waddell (then Waddell & Hardesty) was not selected for the design. On August 2, 1927, the Commission selected the Boston firm of Fay, Spofford, and Thorndike (FST) as their engineers. All three were MIT graduates and formed the firm in 1914.

Spofford, then also a Professor at MIT, was the main designer. As was and still is common, the designer looks at various options as to the type of bridge. Since the channel could not be blocked during erection, the choice of bridge type would need to be built without falsework. In an article published in the Transactions ASCE, Spofford wrote: The Commission and its engineers were in agreement that the bridge should have as pleasing an appearance as possible consistent with the foregoing. The historic importance of the site and the fact that the bridge would be conspicuous for many miles, on account of its height, made its appearance of special importance. Borings and test piles disclosed that the soil overlying the bed-rock[sic] was extremely soft and that the bridge piers would have to be carried to bed-rock[sic] which over a part

of the site occurs at a depth of 100 feet below low water level. After consultation with Army officials, it was decided to provide a vertical clearance at the channel span of 90 feet above standard low water (Elevation 92.5) for a width of 186 feet, and a clearance of 73 feet above the same level for a width Lake Champlain Bridge 1929-2012. of 300 feet. The total width of the lake at type can be given a more pleasing appearance, this water level is approximately 1,500 feet. consistent with economy, than any of the other In selecting the type of bridge, consideration types of truss bridges. One objection sometimes was given to the relative advantages and raised to continuous bridges is that settlement disadvantages of the following types: a) Endof foundations causes serious changes in truss supported truss bridge; b) Cantilever truss stresses, but with piers supported on bedbridge; c) Continuous truss bridge; and d) rock[sic], as in the Lake Champlain Bridge, Suspension bridge. this objection does not exist.

Spofford rejected the simple span (end supThe central part of the bridge had flankported) truss stating, “The writer found it ing spans and a main span that was to be impossible to sketch any simple span design continuous over two piers. Even though that was at all satisfactory in appearance.” He structurally different, the span resembled rejected the cantilever as it would have needed one proposed by the State Engineer and a short suspended span, which was also unsat- Surveyor, even to the curved lower chords isfactory with respect to appearance. This of the theflanking flankingspans spans andand a suspended a suspended deck left the continuous truss span or suspension for decka portion for a portion of the central of thespan. central FSTspan. subbridge. Spofford then tried two suspension mitted FST submitted its plans and its plans estimate andonestimate November on bridge designs. One had a center span of 700 15, November 1927. 15, 1927. feet, with two flanking spans of 350 feet each Structurally, and visually, it was simiand two 185-foot approach spans. The other lar to Smith’s Lachine Rapids Bridge had a center span of 1,000 feet with two flank- (STRUCTURE, April 2017) in that the ing spans each with a length of 350 feet. He central span span could couldbebeerected erectedbybycantilever cantileand the Commission agreed neither would methods ver methods usingusing the flanking the flanking spans asspans anchor as look as well as the continuous span finally spans. anchorEach spans. cantilever Each cantilever would then would be then conadopted. Spofford wrote of his decision to nected be connected at mid-span at mid-span to create the to continuous create the use a continuous truss: continuous span. span.

The continuous truss type has all the advanThe 2,190-foot bridge, as designed and tages of the cantilever type except that of built, consisted of three plate girders of 50 statical determination, and is also economical feet each, and one simply supported 225of material, especially when the dead stresses foot Warren deck truss, two continuous are large compared with the live stresses as in deck trusses (225 feet and 290 feet), the the case of a highway bridge with a concrete three-span deck to through truss (290 feet, floor and with spans as long as those of this 434 feet, 290 feet), one span deck truss bridge. The lack of statical determination (290 feet), with five plate girder spans on requires additional mathematical investiga- the Vermont side. tions on the part of the designers, but involves As to the computational methods used, no special theoretical difficulties, merely Spofford wrote, “In designing the continuous increasing the labor of making the necessary trusses, preliminary stress computations were computations. The deflection of a continuous made using the values given in Griot’s tables span is less in amount and occurs with less of influence data for shear and moment for rapidity than that of a cantilever span, this continuous beams of constant moment of being an important element in favor of the inertia. In the main channel spans, the values continuous bridge. Moreover, the continuous of the dead reactions, as obtained at Piers 5

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and 8 [ends of the central continuous spans] from the preliminary design, were used as final values. A clause was introduced into the specifications stipulating that these predetermined reactions should be secured by the use of hydraulic jacks fitted with gauges, and with the ratio between inches raised and reactions applied determined by actually raising the ends of the trusses before the laying of the concrete floor.

Before the final establishment of the reactions by jacking, the Method of Least Work was used to determine the dead-load reactions due to the steel work and the temporary tracks and concrete forms which were on the span prior to the jacking. The Method of Least Work was also used for a final determination of the live stresses, the irregular depth of the trusses in these spans making it seem advisable to obtain these stresses by a method more accurate than that used in the preliminary design.”

FST addressed the three main problems that engineers had with continuous truss design. Since the piers would rest on rock, the settlement would not be a problem.

Using both the theorem of three moments with a constant E and I for preliminary design and later the Method of Least Work

for final live load stresses, they refined the computational methods. Spofford based the size of the members in his continuous channel spans on the dead loads of the members plus the travelers.

Merritt-Chapman & Scott was awarded the contract to build the bridge on May 15, 1928, and work began on June 14. The American Bridge Company fabricated and erected all of the steel work. The total cost of the bridge was $967,800, and total cost of the project was $1,149,000. The bridge opened to traffic on August 26, 1929, on time and under budget. FST would later build similar bridges across Little Bay (275 feet) in Maine (The General Sullivan Bridge) and two across the Cape Cod Canal, the Sagamore (616 feet) and Bourne (616 feet) bridges.

With the completion of the Lake Champlain Bridge and the Ross Island Bridge (Portland, Oregon) by Gustav Lindenthal (one of the most vocal supporters of continuous truss bridges), more engineers accepted the merits of this type of structure, especially for highway purposes.

Many engineers still believed, however, that long span simply-supported trusses were economically more efficient.

The Lake Champlain Bridge underwent repairs in 1991. In 2009, underwater investigations indicated the pier foundations were defective, and it was decided the bridge be replaced. As the DOT Regional Engineer said, “under certain conditions, we were afraid the bridge could fail abruptly.” A new design was prepared that was similar to that proposed by the New York State Engineer and Surveyor in 1927, in that the channel span consisted of two cantilever arms on which sat a tied arch. In 2011, the tied arch span, fabricated off-site, was floated into place and jacked up to sit on the cantilever arms. The old bridge was dynamited, dropped into the Lake, and removed. The new bridge opened on November 7, 2011.▪ Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He was formerly Director of Historic Bridge Programs for Clough, Harbour & Associates LLP in Albany, NY, and is now an Independent Consulting Engineer. Dr. Griggs can be reached at fgriggsjr@verizon.net.

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Spotlight

21st Century Seismic Design Saves an Architectural Landmark Performance-Based Retrofit of a Mid-Century Masterpiece

By Bryan Seamer, S.E.

LPA, Inc. was an Award Winner for its Christ Cathedral Tower of Hope Seismic Retrofit project in the 2016 NCSEA Annual Excellence in Structural Engineering Awards Program in the Category – Renovation/Rehabilitation.

T

his is the story of a prominent piece of architectural history saved from the wrecking ball and the fusion of art and science that made it possible.

Built in 1968, the Christ Cathedral Tower of Hope has been an important Southern California landmark for almost fifty years.

Designed by celebrated architect Richard Neutra, one of the titans of the mid-century modernist movement, the 13-story cast-inplace concrete tower has been called “an overlooked masterwork in Neutra’s oeuvre” by architectural historian Barbara Lamprecht.

Once the tallest building in Orange County, the structure is notable for its small but daylight-rich floor plates and vertigo-inducing cantilevered, post-tensioned exterior concrete stairs. The primary structural frame of the tower consists of concrete slabs and beams, supported by rectangular and trapezoidal board-formed concrete columns.

In 2012, the Roman Catholic Diocese of Orange purchased the landmark Crystal Cathedral campus in Garden Grove, California, to serve as its long-planned diocesan cathedral and promptly renamed it Christ Cathedral. From the beginning, the fate of the Tower of Hope was in doubt. A seismic assessment suggested that it was the most vulnerable building on the Cathedral campus because of the limited ductility of the concrete moment frames that form the tower’s primary seismic force resisting system.

Like many concrete buildings built prior to the mid-1970s, the Tower of Hope is susceptible to severe damage or even collapse during a large earthquake due to the lack of ductile steel reinforcing in the moment frame beams and columns. While the Diocese recognized the Tower’s architectural and cultural significance, the safety of its large parish population had to take precedence. It was at this point that contingency plans were made to demolish the iconic building.

The Diocese’s desire to save the building was made even more challenging by the very small floor plates and the need to respect the signature inside-outside connectivity of Neutra’s original design. The renovation design team was challenged to significantly improve seismic

performance while not altering the historic lobby, or the famous glass-enclosed “Chapel in the Sky.” A traditional seismic retrofit would necessarily impact the historic fabric throughout all levels of the structure and necessitate significant new construction on the ground floor, extinguishing the fluid, airy character of the glass-walled lobby.

In the Spring of 2013, after several schemes involving new shear walls, braced frames, and foundations proved infeasible, a non-traditional retrofit approach was considered: performancebased design. Instead of disregarding the strength of the existing concrete frames because they did not satisfy the prescriptive detailing requirements of modern building codes, seismic forces would be dissipated passively, allowing the existing concrete frames to resist dramatically reduced demands safely. This approach would use a non-linear time-history analysis to demonstrate that the concrete frames would remain nearly elastic during a 475-year return period earthquake.

Two seismic retrofit strategies were combined to provide life-safety performance without impacting the historic architecture of the building or reducing the usable floor space. First, fluid viscous dampers in a two-story X configuration were installed on floors two through five to reduce the demands on the existing concrete frames. Where the damped demand remained higher than the capacity of the original columns, the columns were strengthened with carbonfiber-reinforced polymer. Because of its thin profile and high strength, the fiber-wrap allowed for the strengthened columns to be clad with finishes appropriate to the building’s period of architectural significance.

One of the inherent challenges of this retrofit strategy was developing high damper forces into the existing concrete frames. Extensive groundpenetrating radar scanning was performed to locate the existing rebar accurately. Each new connection was individually designed with customized bolt patterns so that through-bolts and expansion anchors could be placed without damaging the existing rebar. In cases where damage to column ties could not be avoided, confining fiber-wrap was added around the perimeter of the column beneath the damper connection

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plates. The performance-based design strategy limited major construction activities to four of the tower’s thirteen floors, which proved to be far more efficient and less disruptive than a traditional retrofit strategy.

In tandem with architectural and mechanical modernization and the restoration of the iconic lobby and 13th-floor chapel in the sky, the total project cost came to just over $6 million. The success of the seismic retrofit portion of the project led Diocesan representative Rob Neal to comment “LPA took the time to better understand our requirements and needs, resulting in an innovative and elegant solution that preserved the historic architecture of the tower and saved us almost $3 million in construction cost.”

Over the fifty-year lifespan of the Tower of Hope, the practices of architectural design and structural engineering have drifted from their historical unity leaving each profession worse for the parting. The salvation of the Tower of Hope is the result of a 21st-century design team recreating the deliberate synthesis of art and science that is at the heart of Richard Neutra’s mid-20th century modernist design.▪ Bryan Seamer is the Managing Director of Structural Engineering at LPA, Inc., an integrated design firm based in Irvine, CA. He can be reached at bseamer@lpainc.com.

2017 STRUCTURAL ENGINEERING SUMMIT

NCSEA News

News form the National Council of Structural Engineers Associations

October 11-14, 2017 | Washington Hilton | Washington, D.C.

Terrence Paret to Keynote Summit: Shaking Up D.C. – The Insider’s Story In 2011, the Washington Monument and the Washington National Cathedral were subjected to ground shaking from the Magnitude 5.8 Mineral, Virginia earthquake whose epicenter was roughly 80 miles from the National Mall in Washington, D.C. Both experienced significant damage, some of which was quite spectacular, which highlights the particular seismic vulnerabilities of monumental masonry structures subjected to modest ground shaking. The presentation will cover a broad range of topics related to these iconic damaged structures including some fascinating historical aspects, seismological background, post-emergency response, earthquake damage survey methods and survey results, seismic vulnerability analysis, repair designs, construction administration and lessons to be learned from the Mineral earthquake about low probability events in the Eastern U.S.

Terrence Paret

Since joining Wiss, Janney, Elstner Associates in 1986, Senior Principal, Terrence Paret has performed hundreds of engineering investigations focusing on the evaluation of structural performance after damaging events. Paret has authored or coauthored over 80 technical papers and has received a variety of awards for his research and practice, including the 2001 Moisseiff Award from ASCE, the 2008 AISC Presidential Award of Excellence in Structural Engineering, the 2012 Oliver Torrey Fuller Award from the Association of Preservation Technology International, the 2016 ICRI Project of the Year Award, and the 2016 ASCE Region 9 Seismic Retrofit of the Year.

The 2017 Structural Engineering Summit Keynote Session will be held on Thursday, October 12 at 9 AM.

Ending Soon: Register for the Summit Now to Save $100 with the NCSEA Early Bird Rate!

This year’s NCSEA Structural Engineering Summit celebrates 25 years of bringing together the best in structural engineering.

Designed by structural engineers for practicing structural engineers, the Summit hosts all you need to advance your career and the profession. An array of educational sessions are available each day, featuring a special third track on Thursday dedicated to young engineers. The Trade Show is jam-packed with companies providing products, software and tools. The NCSEA Awards Banquet on Friday night honors structural engineering ingenuity as well as individuals committed to advancing the field and the association. Visit www.ncsea.com for the complete schedule and descriptions.

Hotel reservations for the 2017 Summit are going faster than ever. The Washington Hilton is located in the epicenter of vibrant neighborhoods and located only blocks from the Dupont Circle Metro. Convenient for those who wish to explore, the hotel is a pickup and drop-off stop for the Big Bus open-top sightseeing tour. Guests can enjoy a guided tour of Washington D.C.’s best landmarks and attractions.

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2017 Summit Sponsors

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Decadent bites, finely crafted cocktails, and live music are just a few highlights of this spectacular event!

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NCSEA News

NCSEA YMGSC Hosts Third Annual Online Trivia Event

SE Review & Refresher Course Next Live Course: Vertical—August 5-6, 2017 Lateral—August 26-27, 2017 visit www.ncsea.com for more information

Register now for instant access to the spring course recordings, the NCSEA guide to publications, seminars led by notable and trusted speakers, and the recordings from this course, available 24/7 until the exam!

Upcoming NCSEA Webinars July 11, 2017 Repair of Construction Defects David Flax There are construction defects on almost every job, unless it is a very small job or unless the owner is extremely fortunate. On a typical job, repairs may have to be done to slabs, or vertically, or overhead. This presentation will discuss identifying the defects, repair material selection, repair methods, surface preparation, bonding, curing, and more.

July 20, 2017 Assessment of First Generation PerformanceBased Seismic Design Methods for New Steel Buildings Jay Harris, P.E., S.E. The creation of NIST’s research program to determine if standards for designing new code buildings and assessing existing buildings provided consistent levels of performance lead to the evaluation of a group of buildings designed using ELF and RSA. This webinar will show how they performed during the linear and nonlinear assessments.

August 3, 2017 Geometry and the Design of Truss Structures William Baker, P.E., S.E.

Geometry is arguably the single most important parameter in the design of efficient truss structures. This presentation will explore how the theoretical innovations of Maxwell and other structural pioneers can be combined with stateof-the-art optimization tools to create highly efficient truss structures and will discuss how an understanding of the geometry of forces can be used to design the geometry of trusses.

July 25, 2017 Nonstructural Components Chris Kimball, S.E., P.E. Nonstructural damage has historically accounted for 25-50% of the damage observed in recent earthquakes in the United States. This webinar will assist the structural engineer in knowing when seismic restraint is required, and if it is required when seismic certification and/or special inspections should be provided.

Two exclusive annual plans are available to NCSEA corporate members & SEA members only.

The Live & Recorded Webinar Subscription Plan with access to all live webinars and the entire recorded webinar library, hosting over 180 webinars, or the Live Webinar Subscription Plan.

Plan Visit www.ncsea.com to purchase your subscription today!

Visit www.ncsea.com to register and read the full description of each webinar.

1.5 hours of continuing education. Approved for CE credit in all 50 states.

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News from the National Council of Structural Engineers Associations

On Thursday, May 11th, twelve teams participated in the third annual NCSEA Young Member Group Support Committee (YMGSC) Trivia Night.

This web-based video event brought together Young Member Groups from Virginia, Georgia, Minnesota, Massachusetts, New Hampshire, District of Columbia, Florida, Michigan, Arizona and California for an hour of trivia and social networking. The trivia questions were based on structural engineering-related topics including History of Architecture and Tools That Changed the World.

The event was moderated by YMGSC member Samantha Fox of BCE Structural and was designed to provide a fun activity for Young Member Groups (YMGs) as well as an opportunity for groups from across the nation to connect.

Congratulations to the winning teams MNSEA Young Member Group (Minnesota) and SEAOA Young Member Group (Arizona). These two teams will each be awarded a cash prize to go toward their attendance at the NCSEA Structural Engineering Summit, October 11th-14th, in Washington, D.C.

The Newsletter of the Structural Engineering Institute of ASCE

Structural Columns

Save the Date ASCE 2017 Annual Convention October 8 – 11, 2017

Structures Congress 2018 April 19 – 21, 2018

New Orleans, Louisiana www.asceconvention.org

Fort Worth, Texas www.structurescongress.org

Errata SEI posts up-to-date errata information for our publications at www.asce.org/SEI. Click on “Publications” on our menu, and select “Errata.” If you have any errata that you would like to submit, please email it to Jon Esslinger at jesslinger@asce.org.

SEI Election July 31, 2017, Deadline

The current Board of Governors positions on the Structural Engineering Institute Board of Governors are: representatives from each of the five Activities Divisions (Business and Professional, Codes and Standards, Global, Local, and Technical), one appointed by the ASCE Board of Direction, the current SEI President, the most immediate and available Past President of the SEI Board, and the SEI Director as a nonvoting member. The representatives from the Divisions each serve a four-year term. In accordance with the Structural Engineering Institute Bylaws, this year SEI is conducting an election for a Technical Activities Division (TAD) representative on the Board of Governors. The TAD Executive Committee has nominated Robert Nickerson as their candidate. If you are a member of SEI, please complete and mail your ballot to the address provided. Because we must confirm SEI membership, only signed ballots are accepted.

Robert E. Nickerson, P.E., F.SEI, M.ASCE, is an independent consulting engineer with 39 years of professional experience in the electrical transmission industry. He is a licensed engineer in the state of Texas and a registered professional engineer in seven other states. Mr. Nickerson has extensive experience in the analysis, design, failure investigation, and full-scale testing of electrical transmission structures.

His work includes structural assessment, analysis, and design of lattice transmission towers and tubular poles and substation structures.

Mr. Nickerson has long been active in SEI and has served on committees in the Technical Activities Division and the Codes and Standards Activities Division. He was a member of the SEI Technical Activities Division Executive Committee, Chair of the Special Design Issues Technical Administrative Committee, and Chair of the Electrical Transmission Structures Committee. He has also served as a member of Guidelines for Electrical Transmission Line Structural Loading Committee (MOP 74, 3rd edition), and was the ViceChairman of the Practice Guide to Design of Guyed Transmission Structures Committee (MOP 91). He was a member of the Blue Ribbon Panel for the ASCE Manual of Practice No. 74, 4th edition. Mr. Nickerson is the Chair of the Design of Lattice Steel Transmission Structures Committee and is a member of the Design of Steel Transmission Pole Structures Standards Committee.

Mr. Nickerson has been an active steering committee member for the Electrical Transmission and Substation Conferences held in 2002, 2006, 2009, 2012, 2015, and 2018, and was chair of the 2006 conference. He has received the Gene Wilhoite Innovations in Transmission Line Engineering Award and the SEI President’s Award. He has authored and co-authored papers presented at the Structures Congress and the Electrical Transmission Specialty conferences.

Full Name: _____________________________________Member’s ASCE/SEI ID No:________________ (Please print) Date:______________ Signature: _______________________________________________________________

Return postmarked no later than July 31, 2017 to: SEI Board Election, 1801 Alexander Bell Dr., Reston VA 20191.

SEI 2017 Board of Governors Election – Official Ballot STRUCTURE magazine

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ELECTRICAL TRANSMISSION & SUBSTATION STRUCTURES CONFERENCE 2018 Atlanta, Georgia November 4–8 Dedicated to Strengthening our Critical Infrastructure

Electrical Transmission & Substation Structures Conference 2018 Call for Abstracts and Sessions

The State-of-the-Industry Forum for Transmission and Substation Engineers

1) Expand your knowledge at technical sessions on transmission line and substation structures and foundations.

2) Earn professional development hours (PDH’s) by attending technical sessions and workshops.

3) Network with global leaders and colleagues working with high-voltage transmission structures around the world.

4) Connect with exhibitors showcasing state-of-the-art products, services, and solutions for your transmission line and substation projects.

5) Discover Southern hospitality and enjoy over 100+ live entertainment venues.

Exhibits & Sponsorships Increase your company’s visibility and reach hundreds of industry professionals at this important specialty conference. Contact Bob Nickerson at renicker@flash.net or 817-319-8779, or Sean Scully at sscully@asce.org or 703-295-6154, for exhibiting and sponsorship opportunities.

Questions? Contact Debbie Smith dsmith@asce.org or 703-295-6095. Submit your sessions at www.etsconference.org.

Dedicated to Strengthening our Critical Infrastructure Abstracts & Session Proposals due September 12, 2017

SEI Local Activities Get Involved in Local SEI Activities Join your local SEI Chapter, Graduate Student Chapter (GSC), or Structural Technical Groups (STG) to connect with colleagues, take advantage of local opportunities for lifelong learning, and advance structural engineering in your area. If there is not an SEI Chapter, GSC, or STG in your area, review the simple steps to form an SEI Chapter at www.asce.org/structural-engineering/sei-local-groups.

Local Chapters serve member technical and professional needs. SEI GSCs prepare students for a successful career transition.

SEI supports Chapters with opportunities to learn about new initiatives and best practices, and network with other leaders – including annual funded SEI Local Leader Conference, technical tour, and training. SEI Chapters receive Chapter logo/branding, complimentary webinar, and more.

CONGRATULATIONS TO THE 2017 SEI FELLOWS SEI welcomed its new class of Fellows at the Structures Congress Closing Plenary Lunch on April 8, 2017. Visit the SEI Fellows webpage to learn more.

Fabio Biondini, Ph.D., C.Eng., F.SEI, F.ASCE Larry A. Fahnestock, Ph.D., P.E., F.SEI, M.ASCE John E. Finke, D.Eng., P.E., S.E., F.SEI, M.ASCE Maria M. Garlock, Ph.D., P.E., F.SEI, M.ASCE Eric M. Hines, Ph.D., P.E., F.SEI, M.ASCE Ron Klemencic, S.E., P.E., F.SEI, M.ASCE Mary Kay Knight, P.E., F.SEI, M.ASCE

Lance Manuel, Ph.D., P.E., F.SEI, F.ASCE Cheryl L. Rishcoff, P.E., F.SEI, M.ASCE Bradford O Russell, P.E., F.SEI, M.ASCE Halil Sezen, Ph.D., P.E., F.SEI, F.ASCE Peggy Van Eepoel, P.E., F.SEI, M.ASCE Scott Wallace, P.E., F.SEI, M.ASCE Eric B. Williamson, Ph.D., P.E., F.SEI, M.ASCE

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The Newsletter of the Structural Engineering Institute of ASCE

• Discover Technical Knowledge • Hear Project Case Studies • Find Real-World Solutions • Visit Vendors and Learn about their Products and Services The SEI/ASCE Electrical Transmission & Substation Structures Conference is recognized as the must-attend conference that focuses specifically on transmission line and substation structure and foundation construction issues. This event – for utilities, suppliers, contractors, and consultants – offers an ideal setting for learning and networking.

Structural Columns

Top 5 Reasons to Attend

InSIghtS

new trends, new techniques and current industry issues

Designing for Tornados

By Roy Denoon, Ph.D., M.ASCE

T

raditionally, tornados have been overlooked in structural design in favor of more predictable straightline winds. Indeed, all of the design wind speeds presented in ASCE 7 are based on either surface data analysis of thunderstorms and synoptic storms, or Monte Carlo simulations of hurricanes. There is no direct allowance for tornados. This is because the probability of any individual building or structure being impacted by a tornado is small given their infrequency of occurrence and very limited spatial extent.

In recent years, however, a number of very damaging tornados resulting in extensive life and property loss have caused a re-examination of approaches to structural design for tornados. Some more substantial structures are now specifying additional robustness to protect building contents and operations, while an increasing number of more modestly constructed buildings are incorporating places of safety to provide refuge to occupants.

The most recent uptick in awareness of tornados probably began around 2007 with the adoption of the Enhanced Fujita (EF) Scale for classification of tornado intensities. Like the earlier Fujita scale, this uses post-event damage surveys to estimate the wind speeds that occurred and retains a six-point classification scale from 0 to 5. The major change in the updated scale was revised wind speeds that better correlated with observed damage.

The Enhanced Fujita Scale provides values as 3-second gust wind speeds, which are directly comparable with the design wind speeds in ASCE 7. They are not, though, directly comparable with the wind speeds in the SaffirSimpson hurricane scale, which is based on sustained wind speeds with a duration of around one minute.

An EF5 tornado has an estimated wind speed of greater than 200 mph, and this is surprisingly close to the Saffir-Simpson Category 5 hurricane (when converted to a 3-second gust).

Both the EF5 and Saffir-Simpson Category 5 are expected to result in catastrophic damage with a high percentage of homes destroyed.

Based on ASCE 7 wind speed maps, this speed would only be expected to occur around once every 1,700 years right on the southern tip of Florida, the most hurricane-prone area in the continental United States. In the Midwest U.S., where tornados are most likely to occur,

the ASCE 7 1,700year design wind speed is around 120 mph, although most buildings (in Risk Category II and III) would be designed for the 700year wind speed of 115 mph. This lower value equates to an EF2 tornado, which has an intensity that would be expected to result in severe damage, with roofs torn from well-constructed houses and foundations of frame homes shifted. As such, there is some degree of consistency in the reliability of design.

In April and May of 2011, there were a series of severe tornados that tore through Alabama, Missouri, and Oklahoma, the best known being the Joplin tornado. This tornado was extensively studied by academic and professional response teams from around the country.

The Joplin tornado was notable for the large loss of life (158 people) and the economic cost of damage as reported by the insurance industries (greater than $2 billion). While it is not economically viable to design typical woodframed residential properties to resist severe tornados, incorporation of reinforced tornado shelters (whether in basements or the interiors of homes) is feasible, and these are being increasingly adopted in newer construction.

The Alabama tornados occurred in an area with a large concentration of manufacturing facilities, particularly for the automobile industry. This was shortly after the Fukushima nuclear accident in Japan, which severely affected automobile production. These combined events led to risk studies by the industry to assess the potential costs of severe damage to component production facilities, especially when a number of those are located in a limited geographical area. Other facilities, such as data centers, are also increasingly specifying tornado resistance in their design specifications. In these cases, analyses of the economic benefits of additional robustness protecting building contents and operations are shown to outweigh the higher costs of construction.

In May 2013, a severe tornado hit two elementary schools in Moore, Oklahoma, resulting in 7 deaths in one school out of a total of 24 deaths and 212 injuries across the town.

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An ASCE/SEI-commissioned report in the wake of this tornado on Performance of Schools and Critical Facilities highlighted the roles of weak links in the load paths propagating more extensive failures. The reliance of the schools on plans based on “Best Available Places of Refuge” rather than dedicated tornado shelters was highlighted. Recommendations were given regarding the strengthening of existing buildings and incorporation of tornado shelters (for which ICC 500 provides design guidelines) in buildings that expect to shelter a large number of people in the event of a tornado. Further recommendations included revisions to ASCE 7 to incorporate additional guidance for practitioners, and examination of existing schools and critical facilities in tornado-prone regions for vulnerabilities.

Revisions to ASCE 7, at the current state of knowledge, would be expected to cover likely tornado wind speeds and methods of increasing provisions for tornado shelters and their robustness, rather than changing the pressure coefficients in the standard. However, recent research is beginning to suggest that pressure coefficients in a tornado wind field may differ measurably from those traditionally used for straight-line winds. Ongoing research is likely to result in modifications to the standard.▪ Dr. Roy Denoon is Vice President and Principal of CPP Wind Engineering. He has published numerous magazine and journal articles in the field, as well as co-authoring the CTBUH “Guide to Wind Tunnel Testing of High-Rise Buildings” and editing the Australasian Wind Engineering Society’s “Quality Assurance Manual for Wind Tunnel Testing of Buildings and Structures.” He can be reached at rdenoon@cppwind.com.

2017 CASE RISK MANAGEMENT SEMINAR Time-Tested Techniques for Managing Your Firm’s Risk Scheduled for August 3-4, 2017 in Chicago, Time-Tested Techniques for Managing Your Firm’s Risk will help your firm reduce its rate of claims against structural engineering projects, as well as raise the level of quality services provided by all project participants.

Who should attend? Principals, Owners, Project Managers, Risk Managers!

Session Topics • Strategies Forensic Engineers Use to Unravel Construction Disputes • Information Security in Contracting • Case Studies for Professional Liability A special dinner presentation on “The Future of Structural Engineering” by Ashraf Habibullah, President and CEO of Computers & Structures Inc. and structural engineering roundtable!

Register at: www.acec.org/coalitions/upcoming-coalition-events/ >> $399/Coalition Members >> $499/ACEC Members >> $599/Non Members LOCATION All sessions will take place right off Chicago’s infamous Magnificent Mile at the DoubleTree Magnificent Mile, located at 300 East Ohio Street Chicago, Illinois.

Phone: 1-800-222-8733 and refer to group code: ACE Special Rate - $194/night until July 11, 2017 or until block sells out!

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opinions on topics of current importance to structural engineers

Structural Forum

4 Ways to Empower Your Team

By Solomon Ives, P.E.

W

hat is your firm’s greatest asset? Your client’s trust?

Your firm’s branding? Or could it be the sum abilities of your staff? If staff ability even ranks in your top three, then developing staff and utilizing their talent effectively is crucial for success.

A key way to do this is to empower them.

Empowerment gives colleagues the courage and strength to take risks and try new roles that will stretch their abilities. As a young engineer, being empowered has released untapped talents within me and inspired my best effort. Whether you lead an entire firm or only a small project, everyone stands to benefit when those around you perform at their best.

Empowerment Latitude Give those around you latitude to take risks and try new things. An empowering moment in my career happened during my first yearend review. My supervisor asked me a series of questions to uncover my unused abilities and invite me to suggest ways to improve my experience at the firm. Questions such as “Are we fully utilizing your talents?” and “What is inhibiting your success?” allowed me to explore new ways to add value to the company. Far from the expected “summary of strengths and weaknesses” or “improvement areas,” the questions were aimed at maximizing my satisfaction and contribution at the firm.

These questions gave me the freedom to expand my job description. As time went on, my supervisor and I discussed design processes improvements, staff development, and client relationships. I began to manage projects, have lunches with architects, and write proposals.

As engineers, we tend to keep a narrow focus to maximize efficiency. An empowered engineer knows they have the freedom to zoom out and engage all of their unique talents and interests to work smarter, not just harder.

My career shifted for the better at that first year-end review because my supervisor gave me the latitude to think outside the box.

Empowerment gives colleagues the courage and strength to take risks...

Support When an engineer takes a risk and shares a new idea, they need support to work it out. Help them set SMART goals (Specific, Measurable, Action-Oriented, Realistic, Timely). If appropriate, set expectations for the budget of time they can spend on the initiative. As long as their ideas align with the company’s direction and values, support them to do it. If they’ve earned a measure of your trust, take a proportionate level of risk on them.

Ask yourself “How can I help them to get their idea off the ground?” or “What resources or public endorsement do they need from me?” This focus contrasts to the traditional leadership style that maintains top-down control to direct the team towards the leader’s goals. To empower is to get underneath a coworker’s idea and supply power to help them succeed. An action as simple as attending a meeting they lead or using a document they created goes a long way to communicate that you believe in their ability to create value.

In my experience, receiving that trust has energized my creativity and motivated me to return the goodwill by focusing on the company’s objectives.

Accountability Follow-up with someone and you will prove to them that you take their development and ideas seriously. Ask questions on the status of their initiative. The point is not to grade their performance, but to develop new strategies and new ways to provide support. Are any adjustments needed to ensure goals are met? Would it be helpful to set up regular meetings?

Celebrate moments of success, as well. It will build confidence and invite further creativity and risk-taking. When we see the connection between our efforts and tangible results, it also builds a sense of power and ownership.

It’s important to stay honest here. If it becomes apparent that a goal is unrealistic or that the budget needs to change, that should be clearly acknowledged. Pretending

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that goals have been met when they haven’t is demoralizing – remember, the initiative is fueled by a desire for genuine success.

Patience Empowerment is a messy business. Some ideas will turn out to be mistakes. Time can be wasted, and it can be awkward or disappointing. But those moments are the very best opportunities to learn and grow.

If a problem arises, discuss honestly what went wrong and what can be improved the next time. If possible, give them a second opportunity, soon, to help their confidence rebound. If you immediately take control when a problem arises, it steals the opportunity for both of you to learn. Additionally, when a company leader spends time fighting another engineer’s fires, they have less time to do their job: grow the company.

The Value Created Empowering people costs time and resources, but the profit gained outweighs the cost.

Imagine a staff team that takes the initiative to improve themselves and make suggestions.

Imagine a team that has earned your trust and is internally motivated to see you and the company-at-large succeed. I have experienced the shift in myself from being a “worker” who trades my labor for wages, to being a “partner” who seeks the mutual success of my company and myself. The result is more value created for the firm, its employees, and our clients.

Who could benefit from your empowerment? Who do you have influence with? There may be situations where more authoritative leadership styles are prudent. But a leader who never takes the time to empower will become the limiting factor in their team’s success. Let’s not fear mistakes, conflict, or “wasted” time when the potential talent inside our colleagues is waiting to be unleashed.▪ Solomon Ives is a Project Manager at Kordt Engineering Group in Las Vegas, Nevada.

He can be reached at sives@kordteg.com.