STRUCTURE magazine | July 2023

EDITORIAL BOARD

Chair John A. Dal Pino, S.E. Claremont Engineers Inc., Oakland, CA chair@STRUCTUREmag.org

Jeremy L. Achter, S.E., LEED AP ARW Engineers, Ogden, UT

Erin Conaway, P.E. AISC, Littleton, CO

Linda M. Kaplan, P.E. Pennoni, Pittsburgh, PA

Charles “Chuck” F. King, P.E. Urban Engineers of New York, New York, NY

Nicholas Lang, P.E. Vice President Engineering & Advocacy, Masonry Concrete Masonry and Hardscapes Association (CMHA)

Jessica Mandrick, P.E., S.E., LEED AP

Gilsanz Murray Steficek, LLP, New York, NY

Jason McCool, P.E. Robbins Engineering Consultants, Little Rock, AR

Brian W. Miller

Cast Connex Corporation, Davis, CA

Evans Mountzouris, P.E. Retired, Milford, CT

Kenneth Ogorzalek, P.E., S.E. KPFF Consulting Engineers, San Francisco, CA (WI)

John “Buddy” Showalter, P.E. International Code Council, Washington, DC

Eytan Solomon, P.E., LEED AP Silman, New York, NY

Jeannette M. Torrents, P.E., S.E., LEED AP JVA, Inc., Boulder, CO

EDITORIAL STAFF

Executive Editor Alfred Spada aspada@ncsea.com

Production production@structuremag.org

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STRUCTURE ® magazine (ISSN 1536 4283) is published monthly by The National Council of Structural Engineers Associations (a nonprofit Association), 20 N. Wacker Drive, Suite 750, Chicago, IL 60606 312.649.4600. Periodical postage paid at Chicago, Il, and at additional mailing offices. STRUCTURE magazine, Volume 30, Number 6, © 2023 by The National Council of Structural Engineers Associations, all rights reserved. Subscription services, back issues and subscription information tel: 312-649-4600, or write to STRUCTURE magazine Circulation, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606.The publication is distributed to members of The National Council of Structural Engineers Associations through a resolution to its bylaws, and to members of CASE and SEI paid by each organization as nominal price subscription for its members as a benefit of their membership. Yearly Subscription in USA $75; $40 For Students; Canada $90; $60 for Canadian

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38 400 & 430 CALIFORNIA STREET

By Kirk Johnston, S.E. and Yvonne Tsui, S.E.

Built among the rubble of the 1906 San Francisco earthquake, The Bank of California Banking Hall, located at 400 California Street, was among the first commercial buildings to be completed following the 7.9 magnitude quake – the deadliest natural disaster in California’s history.

Cover Feature

10 WHEN THE GROUND SHOOK

By

Egemen Sönmez, Ph.D., Rebecca Collins, P.E., S.E. and Reid Zimmerman, P.E., S.E.

On the morning of February 6, 2023, Turkiye awoke to the news of a destructive earthquake in the southeast of the country. Although the epicenter of the earthquake was estimated to be near the city of Kahramanmaras, there was no proper news from many other cities in the region.

48 SAVING A HISTORIC VIRGINIA CITY ICON

By Paul A. Ferrari, P.E.

St. Mary’s in the Mountains Cathedral, built in 1876, in Virginia City, Nevada provided spiritual guidance to the rough and tumble miners of the Comstock Lode.

By Gail Kelley

By M. McElmury, D. Nickell, R. Reckers, S. Pujol

Remote Work

Three Years Later

By Bruce Burt, P.E., P.Eng.

Earlier this year I attended the ACEC Coalitions Winter Meeting in Tampa. One of my favorite parts of the two-day program is the Business Roundtable, where attendees suggest topics for discussion among the participants. Recent topics have included firms’ plans for pay increases, recruitment and retention strategies, and tactics for increasing diversity, equity, and inclusion. With the writing of this editorial in mind, prior to the session I suggested to the moderator the subject of remote work, and how the attendees view it three years since the pandemic’s onset. The moderator just rolled his eyes and said, “Not that topic again!” and laughingly agreed to include it in the list of potential discussion points . As it turned out, a lively conversation ensued concerning the participants’ experiences with remote work and the associated company policies they’ve implemented.

The time spent discussing remote work during the roundtable was informative, confirming some self-evident observations while offering insights into effective remote work practices. The following is a summary of comments from the roundtable participants.

Most of the firms in attendance have some form of remote work policy, ranging from “We hope to see you again sometime” to a restrictive one day a week or less. There was consensus that some office time is necessary to maintain morale and promote corporate culture. Some firms have designated in-office days, while others require in-person attendance at company meetings. Many firms offer incentives to attract staff to the office, such as free lunches, dedicated workstations, or frequent in-office get-togethers.

From a demographic perspective, younger staff tend to prefer being in the office, for the social interaction as well as for the on-the-job learning opportunities. Staff with younger families prefer the flexibility of remote work, and while many of the senior staff enjoy working remotely, they balance that with their in person mentoring and role modeling responsibilities.

Nearly all participants agreed there are significant benefits to entry-level and junior staff being in the office. Some participants felt that younger staff who have worked primarily in the office were performing significantly ahead of their mostly-remote peers. As to the readiness of entry level staff, there were numerous

observations that academic achievement of recent graduates had suffered due to a yearand-a-half of online coursework. I’ve spoken with numerous recent graduates, and all readily admit that remote learning was not nearly as effective as in-person learning, in particular regarding upper-level coursework.

One participant’s firm has increased their onboarding program to expedite cultural indoctrination. Another has beefed up its QA/QC program.

There was little concern about lost productivity. Virtually all firms require timekeeping, which the participants feel encourages accountability. Some firms have a cameras-on policy for internal meetings to discourage attending meetings while standing in line at the grocery store.

There was general agreement that interactions with clients have increased, but most of those connections were of the FaceTime (or Zoom, or MS Teams Meeting) variety. In-person client contact has significantly decreased.

Most firms felt quality has remained the same or improved. There were no reports of claims increases that could be attributed to remote work, although it was acknowledged that given the latency period it may be that claims have yet to surface. One ISO-certified firm did see an increase in “internal audit flags,” and although I don’t know precisely what that means it sounds bad. And many firms have postponed or abandoned physical office expansion plans.

The final question on the topic was whether remote work has had a positive, neutral or negative impact on participants’ firms. In a show of hands, the vast majority indicated remote work was considered a strong positive within their firms.

As a follow-up to the ACEC Coalitions roundtable discussion, ACEC Coalitions staff conducted a web-based poll of members on their views of remote work. Although a small sample size, the results are worth sharing.

Most poll respondents, nearly 80%, echoed the roundtable attendees in having either a formal or informal hybrid work policy, while a small percentage require daily attendance (allowing for exceptions). Most respondents felt that by many measures (quality, productivity, employee morale) it was business as usual, with little perceived change from pre-pandemic days.

However, there was some divergence from the

views of the roundtable participants. A not insignificant number of poll participants, around 20%, felt they are underperforming in many areas, and company profitability has suffered. One respondent, while acknowledging the benefits to the employee, listed many disadvantages experienced as the employer: a disconnect between firm leaders and the staff; a loss of ad-hoc problem-solving sessions; difficulty in scheduling impromptu meetings; and inefficiencies in the checking process, in particular drawing review. While many of these issues might be overcome with improved office policies, it’s hard to deny the loss of valuable spontaneity in the office dynamic.

So, are there lessons for your firm in this small sampling of responses? I think there are a few:

• There does not seem to be an optimal remote work approach. Remote work policies appear to follow a classic bell curve, with outliers who require full office attendance and those with no restrictions on remote work. Most firms have adopted a two or three-day-a-week in-office requirement.

• Many firms are employing strategies to increase staff engagement, maintain office culture, and promote firm loyalty. Those firms express the highest remote work policy satisfaction rates.

• Quality, productivity, and profitability do not seem to have slipped for most firms.

•The in-office experience has diminished. Remote work policies will continue to evolve and may at some point coalesce around a consensus in-office workweek. Humans are social beings (even engineers!) and desire some level of in-person interaction. Young staff will develop more quickly and more fully with in-person mentoring. The office work environment is not dead, but workspaces will be adapted to an appropriate density. As the demand for engineers continues to outpace the supply, the flexibility remote work affords will be an essential component to attracting and retaining the engineering firm’s most valuable resource – its staff.

Bruce Burt is Vice President of Engineering with Ruby+Associates, Inc., located in Bingham Farms, MI. He is Chair of CASE and a member of the CASE Contracts Committee. (bburt@rubyandassociates.com)

structural INFLUENCERS

Dr. Kit Miyamoto

Dr. H. Kit Miyamoto, Global CEO of Miyamoto International, built a fiveperson engineering practice into a worldwide engineering firm that has 30 locations on four continents with one humanitarian purpose: to make the world a better, safer place from a local, five-person engineering practice. Miyamoto International is known for its innovative engineering and disaster response and reconstruction. Its purpose-driven mission and strength-focused culture drive the company's growth and attract equally passionate and teamwork-focused team members. Dr. Miyamoto holds graduate degrees from the Tokyo Institute of Technology and California State University, where he has been recognized as a Distinguished Alumni. He has been a California Seismic Safety Commissioner for the past 12 years. He has won the Engineering News Record’s “Global Best Project” award an unprecedented five consecutive times. Major media, such as CNN, CBS, ABC, Los Angeles Times, New York Times, and Rolling Stone, have profiled him. He is featured in the “Designing for Disaster” exhibit at the National Building Museum.

You are well known for your post-disaster investigation and recovery efforts. But before we discuss that subject, would you tell our readers a little about your life and engineering career before the formation of Miyamoto International?

I was born and grew up in Japan. When I turned 18, I left for the United States... to play for the Dallas Cowboys. Well, as you can see, that plan didn’t quite work out. I played college football for a couple of years, but then blew out my knee. Plan B was to be a structural engineer! That worked out okay. I attended California State University Chico (Chico State as it is often called) and received a Bachelor of Science degree in civil engineering. It took me eight years since I was too busy playing football, being a fraternity brother, a resident advisor, a forest firefighter, a gold miner, and all those other opportunities that college gives you.

So then you started your engineering career. How did that turn out?

Yes, I joined a local Sacramento structural engineering company called Marr Shaffer. It was a small, half-dozen-person company but well respected. The firm had been around since 1947. When I arrived, they specialized in engineering school and small commercial projects like most small firms. Honestly, I was an underperforming young engineer. I remember that John Shaffer wanted to fire me during my first year. But I talked him out of it. It was a good thing for both of us.

How is that?

I got better, and eventually, I bought him out. It was my first acquisition. It was not cheap, but I was able to pay him out over several years. I am glad I could do that and pay back my first mentor, John. Nothing stopped him. Nothing was impossible for him. This attitude carries on today at Miyamoto.

What about the early days at Miyamoto?

After I took over Marr Shaffer in 1997, we grew quickly in Sacramento. We did a lot of interesting things. We used viscous dampers and performed performance-based engineering. It is common now, but it was unheard of then as applying to day-to-day practice. I knocked on the doors of many famous professors like Popov, Whittaker, Hanson, and Shole to learn from them. Roger Shole was truly my mentor for displacement-based engineering. He always encouraged me. I can still hear his voice saying that I was a shoulder above many others. I don’t know if that was true or not, but I believed it. So, Miyamoto established very distinct technical foundations. Wherever we go,

there are structures with dampers, from Kathmandu, Haiti, Istanbul, to Delhi. You obviously wanted to do more than run a local engineering practice. Would you tell us about that?

We opened offices in Los Angeles and Orange County in the early 2000s. We changed our name to Miyamoto International. Some thought I went crazy since we didn’t have any “international” anything then, not even a single project. But it was my aspiration. Around this time, Mark Zweig joined our board. He was the CEO of Zweig White, the AEC business consultant of the world. He taught us the foundations of entrepreneurship. His thoughts and philosophy of business live well here today.

Then I met Peter Yanev at an EERI meeting, fresh out of EQE. Peter founded EQE, served as CEO, and led it through tremendous growth. He grew a two-person firm to a 700-person firm globally. To me, he was a towering figure. I wanted to be just like that. When I met him, I asked if he could join us as a board member. I don’t know why he said yes, but this was a pivotal moment for us to be international.

What did you learn from Peter?

Again, he was not cheap, but I learned a lot from his genius. One week, he asked me to go to Washington, DC, to meet all of his World Bank contacts. We leveraged all of the contacts I met that week. Soon after, we won our first international project in Romania; a World Bank-financed earthquake disaster risk reduction program. It was in 2005. Peter always pushed me to go and meet these foreign clients. I had no experience and no idea what I should be doing, but it all worked out. Our performance-based engineering background really helped us to understand these countries’ engineering issues and codes. In 2008, we worked on the Istanbul 2000 School Retrofit Project. This project became very famous in the circle of international development influence. He also pushed me to personally go on reconnaissance trips whenever and wherever earthquakes happened. The 2008 Sichuan earthquake was a change maker. There was a Chinese government ban on foreign engineers visiting the area, but we purchased tourist visa to Beijing and went there anyway. We saw thousands of kids killed in the collapsed schools. At that time, the Chinese used unreinforced masonry piers to support concrete precast slabs.

Essentially a death trap. We quickly published a technical bulletin showing the failure mechanism. Someone translated it into Mandarin, and it was on the internet everywhere. After that, I understand that the government implemented seismic strengthening programs throughout the country. I am not saying that we influenced its decision, but at least our information came out quickly and widely.

It is a leap from traveling to see earthquake damage to becoming an internationally recognized expert in earthquake risk reduction. What happened next?

I met Professor Wada from the Tokyo Institute of Technology on this recon trip. He was one of the most celebrated earthquake structural engineers in Japan. Even today, he is. I asked him to take me as a Ph.D. student. Why not? He is the best. I may have been one of the first remote learning students in the world. I was very interested in probabilistic risk analysis. I was really influenced by what EQE was doing with all their fancy analyses for the insurance industry. So for the next four years, I diligently worked on ‘Probabilistic Risk Identification of Steel Structures with Viscous Dampers.’ It essentially created a database of collapse by simulation of damped structures with OPENSEAS. Dr. Amir Gilani in my office taught me a lot. Professor Kasai was on my thesis judging committee, and he was a tough one. I had to add an additional chapter just for him.

So, you can see a common theme in this story. Meet the best and ask them to be your teacher. Today, I have mentors. One is Jaimie Clare Curtis. Jaimie sits on our board, and she is much younger than I am. She was an executive manager of Zweig Group and now with Private Equity. Her business insight is so critical to us. It is nothing to do with age. One needs to learn from the smartest in the business.

Was there a seminal event that really triggered the “international” aspect of Miyamoto International?

In 2010, a great earthquake happened in Haiti. This 7.0M earthquake near Port au Prince’s capital killed more than 300,000 people. We were soon contacted by the World Bank and other international entities for whom we had worked since the fateful one-week DC trip. From all the things we have done and learned in the past, we were able to connect the dots in absolute chaos. It seemed that no one knew what needed to be done or how or when. We connected UN agencies, the US military, the World Bank, the Haitian government, and Haitian engineers to do damage assessment and the start of repairs to broken houses by building capacity in Haitian engineers and masons. We assessed roughly 420,000 structures, and in the end, more than 15,000 structures were repaired or reconstructed, often in extremely difficult conditions. It was the middle of a deep recession in California, so everyone in California’s offices had spare engineers that went to Haiti.

We learned a lot from Haiti, and Haiti really put us in the spotlight of the UN, World Bank, and the US government. They had never seen anything like us.

We have discussed the traditional structural engineering part of the business that you bought and grew. How much greater is the sum of the parts than 1 + 1 = 2? How do the two sides of the business support each other?

Our company is organized around these two expert areas, engineering and international development/humanitarian assistance.

In the US, we provide structural and mechanical, and electrical engineering. We have offices in all large California and Nevada cities; Sacramento, San Francisco, San Jose, Pleasanton, Ontario, San Diego, Reno, and Las Vegas, plus Los Angeles and Orange County, as I mentioned earlier. We are also in DC and Puerto Rico. We do great engineering there, from the Google Headquarters to the Hollywood Bowl to Amazon distribution centers.

The International Development/Humanitarian Assistance work is not only highly technical. But, also involves many more moving parts, such as politics,

culture, language, and legal than our US work. Unfortunately, only a few engineering firms are involved. This always bothers me. Even the great EQE really didn’t get into humanitarian work that much.

The structural engineers work alongside the International Development/ Humanitarian Assistance staff as one unified team, not as separate profit centers as some do. Both expertise work together well. US offices learn from what we see in disasters globally, and international development uses US expertise. Like a ping pong ball. But each shot gets bigger and bigger after each volley.

Your firm has roughly 30 offices worldwide. Many firms struggle with and are scared by having two or three branch offices. How much effort is involved in keeping that shipping headed in the right direction? What is your secret?

We have 30 locations with 400 staff members on four continents. I found it challenging to lead the firm, but not too complex since the same mission unites us; make the world a better, safer place. This is not a hollow marketing slogan. We really mean it. Just hire right. Then trust and serve them with all you got. Set one vision and goals. Set up the right strategies. Build cohesive teams. Decentralize decision-making trees. Make each team accountable for results by using transparent financials. Communicate with all channels, internally and externally. That’s about it.

Concerning mentoring and the opportunity to run one’s own office, your firm seems to offer tremendous opportunities for personal growth. What is your approach, and what have you found that works best?

We set our goal to double our size every five years. Why? Because it will provide more opportunities for our staff and it will have a bigger impact on the world. We want to merge with like-minded companies each year. I see a huge benefit to this. I also see that the previous owners get a second wind. They can focus on what they really like to do. They usually have an excellent younger second line. We give them a significant opportunity to grow together. The learning opportunity is so steep for both sides. This is so cool. Love it. One of our vision statements said, “Positively impact society as a 1,000person firm.” This has been around for a while, but we will get there in about 5 to 10 years.

Of all the projects and recovery efforts you have been involved in, would you describe the one or two that you are the most memorable about and why?

We are currently deeply involved in Ukraine’s reconstruction. We are working with the UN to repair war-damaged schools and high-rise apartments. We have 100 local staff and growing. I spend about half of my time there now. We see this as a part of the effort to reverse Putin’s destructive agenda. We should have no room for killers, rapists, and thieves. I saw firsthand what Putin’s army put people through there. It is much worse than what CNN shows you, by the way.

We need to be unified to confront this tragedy. Believe it or not, engineering is part of the solution. We fix faster than Putin breaks. We are under constant air raids, but we are committed. We will be there until all is fixed.

Successful firms plan for ownership transition and fir h to get by is a trend too. You are personally on the complete opposite end of the spectrum. What advice would you give to young engineers?

Well, I think it was the famous John Maxwell who once said, “If you love what you do, you never work again.” It is true for me. I don’t think I “worked” for at least 20 years now. You need to find your passion and stick to it.

In terms of your personal legacy, what do you think you will be remembered for, and what are you most proud of?

In a jungle of Haiti last year, I met a young mason using a 135-degree hook per ACI details for school reconstruction. I asked him how he knew that. He said, “what’s wrong with you? It is common sense.” That’s the legacy I want to create in many more places before I leave this place.■

When the Ground Shook

Post-Disaster Observations of the Kahramanmaras Turkiye Earthquake Sequence — Part 1

Turkiye

Part 1

By Egemen Sönmez, Ph.D., Rebecca Collins, P.E., S.E. and Reid Zimmerman, P.E., S.E.

On the morning of February 6, 2023, Turkiye awoke to the news of a destructive earthquake in the southeast of the country. Although the epicenter of the earthquake was estimated to be near the city of Kahramanmaraş, there was no proper news from many other cities in the region. While everyone was trying to grasp the extent of the event and its effects, a second devastating earthquake occurred in the nearby area. As the day progressed, the gravity of the situation began to emerge.

At 4:17 AM local time, a 7.8-magnitude earthquake occurred as a result of a complex fault rupture with a total length of about 250 miles. Nine hours after that, a second 7.5-magnitude earthquake was triggered, 60 miles north of the first one, with a total fault rupture of

about 120 miles. This sequence of shallow earthquakes together with their numerous aftershocks caused widespread devastation, with the total number of fatalities exceeding 50,000 and more than 200,000 buildings collapsed or were heavily damaged over a vast region in southern and central Turkiye. Eleven provinces (Adana, Adıyaman, Diyarbakır, Elazığ, Gaziantep, Hatay, Kahramanmaraş, Kilis, Malatya, Osmaniye, and Şanlıurfa) were affected by these earthquakes with a total population of 14 million people. This corresponds to approximately 15% of the country’s population. It is estimated that more than 1.5 million people were left homeless. The financial burden of the earthquakes on the Turkish economy is estimated at $100 billion (US) which may amount to nearly 10% of GDP in 2023. Just the

damage in housing units alone constitutes 55% of this burden.

Shortly after the earthquakes, the authors joined an extensive reconnaissance team organized by the American Concrete Institute (ACI) Disaster Reconnaissance Committee (Committee 133) and including representatives from American Society of Civil Engineers (ASCE), Structural Engineering Institute (SEI), the National Institute of Standards and Technology (NIST), and structural engineering faculty and students from numerous universities. This article is the first in a two-part series and presents a background on regional seismicity, seismic design and practice in Turkiye, and the authors’ observations from the field. The second article will discuss observations relevant to structural practice in the U.S.

Regional Seismicity and the Earthquake Sequence

Turkiye is located in one of the most seismically active regions in the world. The country is mostly located on the Anatolian block, which sits between three main tectonic plates, namely the Eurasian, African, and Arabian plates. Nearly 20 earthquakes of magnitude greater than 7.0 and 200 that are greater than 6.0 have struck Turkiye since 1900, meaning one major earthquake about every five years.

The Kahramanmaraş Earthquake Sequence occurred in the Eastern Anatolian Fault Zone. It is a 400-mile-long left-lateral strike-slip fault system where the Arabian plate is pushing the Anatolian block towards the northeast. This fault system is known to have produced damaging earthquakes over its various segments throughout history. However, a segment that is named the “Maraş seismic gap” was silent for at least 500 years. The first earthquake began on a smaller segment close to this seismic gap and, after “jumping” to the main fault, propagated bilaterally towards the northeast and southwest. Ten minutes later, a 6.7-magnitude aftershock occurred 6 miles west of the mainshock’s epicenter. This mainshock and the following aftershocks triggered the 7.5-magnitude earthquake on a splay fault of the East Anatolian Fault Zone nine hours later. All were shallow earthquakes having focal depths of 4-5 miles, which means that the massive amount of energy did not have time to dissipate before reaching the foundations of structures.

Fortunately for researchers, the Kahramanmaraş Earthquake Sequence was recorded by a dense network of ground motion stations located both along the East Anatolian Fault Zone and its vicinity. Very high horizontal and vertical peak ground acceleration (PGA) values were recorded during the 7.8-magnitude earthquake by the station closest to the epicenter (2.2g horizontal, 2.0g vertical). Several stations in Kahramanmaraş, Hatay, and Gaziantep recorded PGA values between 0.5g and 1.0g. The stations along the fault rupture also recorded extreme peak ground velocity (PGV) values (on the order of 40 in/s) which could be interpreted as an indicator of the damage potential. The recorded PGV values at these stations were 3 to 4 times greater than the predicted values given by the seismic hazard map of the Turkiye Building Earthquake Code. For the records close to the fault rupture, the design spectral accelerations are generally exceeded for a period range that corresponds to the typical residential buildings (0.5

to 1.5 seconds) in the region. These high demands posed a significant risk even for code-conforming residential buildings in some locations. Moreover, the cumulative effect of the successive aftershocks and the 7.5-magnitude earthquake increased the damage and led to more collapses. In various parts of the region, soil bearing failures and excessive settlements of the building foundations were also reported due to soil liquefaction. The ground motion records indicate soil amplification in cities located by rivers due to soft soil conditions. For instance, the observed spectral accelerations for Antakya, which is located by the Asi (Orontes) river, exceeded even the maximum considered earthquake level in the period range from 0.5 to 1.5 seconds (Figure 1).

Seismic Design and Construction Practices in Turkiye

The history of seismic design codes in Turkiye begins right after the 1939 7.8-magnitude Erzincan earthquake. This rudimentary code evolved into a then-contemporary seismic design code in 1975 after revisions made in 1944, 1949, 1953, 1962, and 1968. Partly influenced by the “Blue Book” of the Structural Engineers Association of California (SEAOC), the 1975 code mentioned ductility and base shear force explicitly for the first time. Calculation of the seismic forces based on an observed-intensity-based seismic zonation map, minimum cross-section dimensions, minimum reinforcement ratios, and the design of beam-column joints were among the improvements of the code. This code remained in effect for more than twenty years which corresponded to a time of rapid economic growth in Turkiye. Therefore, a considerable portion of the current building stock was constructed during that period. In 1998, the code was renewed following the developments in structural analysis and design procedures in the world. For the first time, it included nonlinear seismic analysis methods and a probability-based seismic zonation map. The aftermath of the devastating 1999 Gölcük and Düzce earthquakes (7.6-magnitude and 7.2-magnitude respectively) led to revisions in 2007 including procedures for the seismic assessment

Figure 1 Comparison of recorded spectra versus the 2018 Turkiye Building Earthquake Code design-basis earthquake (DBE) and Maximum Considered Earthquake (MCE) spectra for (a) an assumed Site Class D location in Nurdagı, Gaziantep, (b) a Site Class C location in Antakya, Hatay, (c) a Site Class D location in Antakya, Hatay (note that the y-axis scale is different than the other plots), and (d) a Site Class C location in Türkoglu, Kahramanmaras.

and retrofitting of existing buildings. A modern reinforced concrete design specification (TS 500) that mostly follows American Concrete Institute (ACI) 318 was also enacted in 2000, which requires the use of ready-mix concrete and deformed low-carbon steel rebars. Additionally, a building inspection law was enacted after observing a widespread code-conformance problem. The code took its current form in 2018. It is a state-of-the-art code that mainly resembles ASCE 7 for the design of new buildings. The code includes dedicated chapters for the seismic analysis and design of reinforced concrete, steel, prefabricated construction, masonry, and timber buildings as well as high-rise buildings, seismically isolated buildings, and foundations. The code is accompanied by a coordinate-based probabilistic seismic hazard map that gives the design spectra for the service level (43-year and 72-year return periods), design basis (475-year return period), and maximum considered (2,475-year return period) earthquakes. Based on the developments in the seismic design codes and construction inspection, the building stock in Turkiye can roughly be classified as pre-2000 and post-2000. The pre-2000 buildings may be assumed as more vulnerable to earthquakes than the post-2000 buildings due to the inadequacy of the older seismic design codes, the lack of construction inspection, and poor material quality (e.g., plain rebars with low strength, hand-mixed concrete having 1-1.5 ksi compressive strength, etc.). Approximately half of the buildings in the region are known to have been built prior to 2000. However, poor seismic design and construction practices and the lack of proper inspection manifested themselves as severe damage, if not as collapse, in many post-2000 buildings as well. A construction boom in the region within the last 15 years resulted in a considerable number of approximately 10-story residential buildings, many of which were severely damaged or totally collapsed due to poor design and/or construction.

The typical residential building practice in the region consists of reinforced concrete moment frames infilled with unreinforced masonry (URM). The extensive use of URM in-fill and partitions creates significant nonstructural damage risk since these walls fail and fall apart at low drift levels (0.5 to 1.0% interstory drift ratios). The first stories of the buildings are typically open store-fronts set back from the upper floors which have overhangs as balconies or extended rooms (Figure 2). The typically greater height of the first story together with the partial or complete absence of the infill URM walls creates a potential soft/weak story. This issue appears to be infrequently addressed in the design practice, partially because the URM walls are often neglected in the structural design. The use of shear walls

is mostly limited to the perimeter of the elevator shafts or adjacent to stairs if any shear walls are provided. Especially for the pre-2000 buildings, weak-column strong-beam cases could occur which may lead to “pancake-type” collapses during strong seismic events. Another common construction issue is the workmanship that results in poorly consolidated concrete, 90-degree (instead of 135-degree) hooks in stirrups, inadequate splice lengths, and weak confinement regions. These deficiencies are quite common in the pre-2000 buildings but the observations indicated that they are not limited to older buildings. Furthermore, the configuration of the structural system is often compromised to satisfy architectural demands. The most common configuration problems include irregular column lines, beams framing into other beams, absence of beams between the columns, and offsets between the centerlines of beams and columns.

In many recently constructed buildings, instead of regular slab-withbeam construction, the use of shallow beams between the columns with relatively thin slabs and one-way joists is common. The gap between the joists is typically filled with nonstructural elements such as bricks or styrofoam to obtain a flat ceiling. This construction type is preferred due to the reduced formwork cost and the architectural appeal of flat ceilings. However, it reduces the lateral stiffness of the building and increases diaphragm flexibility which are typically not accounted for during design. Although the 2018 Turkish Building Earthquake Code permits the use of this type of system if a minimum amount of shear wall is provided, observations from the field show that this requirement was violated in many new buildings.

One less common building type in the region is the “tunnel form” which is mostly used for the social housing built by the government. The lateral force-resisting system of tunnel-form buildings consists of only shear walls, which are cast simultaneously with slabs using box-shaped formwork. Due to the high shear wall amount, this type of construction could be deemed less vulnerable to earthquakes (Figure 3). The shear wall detailing for tunnel form construction is the same as for shear wall buildings constructed in a conventional manner.

Figure 2 Typical residential construction type in Turkiye (Photo from Nurdagı, Gaziantep)

Figure 3 Example tunnel form building plan

Other contributing factors to the poor seismic design are the lack of engineering licensure in Turkiye and the heavy reliance on all-in-one structural analysis and design software. Additionally, the enforcement of building regulations remains problematic in Turkiye. Although the building inspection law has been in effect for twenty years, strong relations between contractors and the government and possible corruption among the inspection firms and local authorities appear to still exist. In 2018, a construction amnesty was declared by the government to register illegal buildings, most of which disregarded the building regulations. In the hardest-hit cities, over 300,000 buildings benefited from this amnesty as of 2021.

Observations from the Field

Over a week-and-a-half period in late March and early April, the reconnaissance team observed buildings in the earthquake-affected

region. Locations included Adıyaman, Antakya, Gaziantep, Hassa, Islahiye, Kahramanmaraş, Kırıkhan, Malatya, Nurdağı, and Türkoğlu (Figure 4). Buildings were chosen based on ease of access and ranged from functional to severely damaged. The team did not study many collapses, especially as debris removal had already begun. Almost all the buildings were concreteframed, as described above, though occasionally with wood- or steel-framed roofs. The vast majority of buildings observed were built after 2008.

Types of structural damage observed included the following:

•Longitudinal bar buckling and/or fracture of moment frame column and shear wall vertical bars, especially at locations of poor ties (due to 90-degree hooks) or inadequate ties (large spacing, either due to design or construction practices) (Figure 5). In particular, a lack of ties extending into the foundations, leading to bar buckling at the foundation, was observed in many locations.

• Longitudinal bar buckling and/or fracture in moment frame beam horizontal bars, especially just beyond lap splices or where a portion of the longitudinal bars terminated (Figure 6). This damage was also observed where beams framed into the ends of shear walls in the strong direction.

•Shear failures in moment frame columns or in shear walls, especially due to inadequate ties or in shear walls with horizontal bars that were not restrained (Figure 7).

• Shear failure in moment frame beams and in moment frame beam/ column joints (Figure 8).

• Tie fracture in columns or shear wall boundary elements (Figure 9).

• Residual drift, typically occurring in buildings with soft/weak stories (Figure 10).

• Pounding, occasionally at adjacent buildings but more frequently at buildings with a minimal construction joint between building sections (Figure 11).

Tunnel-form construction performance following the Kahramanmaraş Earthquake Sequence varied. Tunnel-form buildings in parts of Karacasu (six stories) and Antakya (eight stories) were undamaged and occupied or the occupants were planning for reoccupancy. However, in Kahramanmaras, the authors observed one 15-story tunnel-form building that had collapsed with several

Figure 4 Cities within the earthquake-affected region visited by the authors Figure 6 Longitudinal bar buckling and fracture in moment frame beam

Figure 5 Longitudinal bar buckling and inadequate ties in wall

nearly identical tunnel-form buildings in the neighborhood severely damaged but standing. The shear wall damage and bar buckling observed in standard concrete buildings also occurred in the tunnel-form construction.

A few items of damage were directly attributable to poor construction practices. Poor concrete consolidation was more frequently observed than in U.S. buildings of similar vintage and led to shear failures or longitudinal bar buckling in shear walls (Figure 12). Column ties were observed bunched up in one area, leading to large gaps above/below the bunched area, which then led to longitudinal bar buckling. Ninety-degree hooks in lieu of the code-prescribed 135-degree hooks were also

observed. In several instances, mechanical ducts and conduit were routed through shear walls that were relatively short in plan, leading to shear failure through the hole.

Buildings with short, or captive, columns were rarely observed in post-2000 reinforced concrete buildings the authors generally surveyed.

Both successful and unsuccessful nonstructural performance was observed. The primary nonstructural damage of note was to unreinforced masonry (URM) infill walls (Figure 13). Not only were the failures a major safety hazard but their damage was also a major obstacle to reoccupancy. Stairs were the other major nonstructural component which were often damaged. The

Figure 7 Shear failure in moment frame column

Figure 8 Shear failure in moment frame beam

Figure 9 Tie fracture in shear wall boundary element

Figure 10 Residual drift

vast majority of stairs consist of cast-in-place concrete without gaps/drift accommodation at the attachment to the floors. Longitudinal bar buckling and/or fracture along with concrete spalling were observed at many stairs where they connected to the main floor. The damage observed to stairs is also concerning as they are the primary, if not sole, exit path for building occupants. Many were rendered unsafe. Nonstructural damage was also observed to water tanks, ceiling grids, overhangs and canopies, and clay roof tiles. Very little to no damage was observed to gas tanks, generators, and mechanical equipment located at grade.

Conclusion

While the Kahramanmaraş Earthquake Sequence could not have been predicted, the destruction and loss of life were, unfortunately, not entirely unpredictable. Seismic hazard research indicated a “seismic gap” in large magnitude earthquakes along the fault system which ultimately ruptured in the Kahramanmaraş Earthquake Sequence. Issues in design, construction, inspection, and enforcement in Turkiye had been previously identified. All of this combined to affect the seismic performance of the primarily reinforced concrete buildings in the region.

Acknowledgments

The authors would like to extend their thanks to the leadership of Santiago Pujol, Cemalettin Dönmez, and Ayhan Irfanoğlu of the ACI Committee 133 team with whom the authors traveled. The authors would also like to thank the extensive list of faculty and students from the Izmir Institute of Technology, Middle East Technical University, Hacettepe University, Harran University, Eskişehir Osmangazi University, University of Canterbury, University of Nebraska-Lincoln, University of Kansas, The National Autonomous University of Mexico, and University of Belgrade who provided essential and exceptional in-field support.■

Egemen Sönmez, Ph.D. (egemen.sonmez@ieu.edu.tr) is an Assistant Professor at Civil Engineering Department of Izmir University of Economics, Turkiye. His research interests include but not limited to seismic design, assessment, and numerical modeling of reinforced concrete structures.

Rebecca Hix Collins, P.E., S.E. (RebeccaC@cplinc.com) is a Senior Structural Project Manager at Coughlin Porter Lundeen in Seattle, WA with extensive experience in structural renovations spanning many material types.

Reid Zimmerman, P.E., S.E. (reid.zimmerman@kpff.com) is the Technical Director at KPFF in Portland, OR and has focused his career on design, evaluation and retrofit of buildings in regions of high seismicity.

Figure 11 Pounding between adjacent buildings

Figure 12 Poor concrete consolidation leading to longitudinal bar buckling

Figure 13 Nonstructural masonry infill damage

on the GROUND Anatomy of Failure

Turkey Earthquake Disaster

By Dr. Kit Miyamoto, Miyamoto International

As I sit on the 11-hour overnight local bus to Istanbul, my mind is still deep in the magnitude of the disaster I have just witnessed in the Syria-Turkey border area. The destruction is beyond anything I have seen in my 20 years of disaster response work.

Yesterday, I navigated the rubble of Hatay, known as Antioch in ancient times. I could barely recognize the oncebustling city. The narrow streets filled with shops, restaurants, and apartments are gone. The tremors from the earthquake last week destroyed 90 percent of the city's buildings, and the death toll is in the thousands.

Although the town is a couple of hundred kilometers from the ruptured fault line, it is situated on a soft riverbed that amplifies earthquake ground motions. Coupled with older building infrastructure, the result was death. Even many modern mid-rise reinforced concrete buildings were destroyed or tilted. The devastation reminded me of Hiroshima after the nuclear blast.

Our team spent a week assessing the damage and devising recovery strategies based on the Turkish government's response and available international funding. We also assisted countless citizens in their broken homes. We estimated that 10,000 structures had collapsed,

and an additional 20,000 buildings might need to be demolished due to extensive damage. On top of that, 100,000 buildings have sustained damages that rendered them uninhabitable. As a result, 2.7 million people were instantly homeless.

Turkey implemented earthquake-resistant design requirements in the late 1990s. Still, anything built before the 2000s, especially older concrete structures on soft soil, were death boxes, as we saw in Hatay. Retrofitting these structures would have made them more resilient, but it takes political will and public support. Unfortunately, such efforts usually only occur after a major disaster.

The most tragic fact is that Turkey's construction industry has no enforced quality assurance. There is no licensing for contractors or engineers, allowing anyone to become either one. Additionally, there is no requirement for quality supervision of construction by engineers. While we inspected many collapsed modern structures, we found that many still collapsed. There were many critical elements that needed to be corrected. During earthquakes, small details matter. For example, smooth river pebbles were used as aggregate instead of small, rugged rocks, which are essential to binding the cement together. Critical tie reinforcement for columns lacked 135-degree hooks even though there were many in quantity. Engineers must be present to verify these small details. When back-to-back 7.8- and 7.5-magnitude earthquakes happen, more than 90% correct is needed. The result was collapse.

Contrary to media reports, the Turkish government's response was decisive, quick, and effective over a widespread area. The affected zone is a 500-km-long by 200-km-wide mountainous area. Not an easy place to access. The Turkish FEMA equivalent, AFAD, spearheaded the effort, providing 150,000 warm tents in the first few days of the disaster and coordinating a 10,000-member international search and rescue team that saved hundreds of lives. All primary and secondary roads were opened by the end of the first week, and the damaged Hatay airport was opened simultaneously. They even started demolishing dangerously damaged buildings

within a few days. They are a determined people. They have a long way for recovery. But if any country can build back stronger and better on this scale, it will be Turkey.■

Dr. Kit Miyamoto is a world-leading disaster resiliency, response, andreconstruction expert. He provides expert engineering and policy consultation to the World Bank, USAID, U.N. agencies, governments, and the private sector. He is a California Seismic Safety Commissioner and Global CEO of Miyamoto International.

Rebuilding America’s Infrastructure

Seismic strengthening of bridge piers.

By Mohammad Jalalpour, Ph.D., P.E., S.E. and Tarek Alkhrdaji, Ph.D., P.E., P.Eng., FACI

US-50 (Lincoln Highway) is one of the country’s last intact transcontinental highways, stretching 3,073 miles from West Sacramento, California to Ocean City, Maryland. Originally constructed in the 1960s, a section of the highway system in the Sacramento region needed extensive improvements to meet the current code. Utilizing an innovative approach, the Multimodal Corridor Enhancement and Rehabilitation (MCER) project expanded the needs of the Sacramento community by widening seven miles of carpool lanes and reconstructing 12 bridges in the process. The project also included replacing the existing concrete pavement, restoring roadway elements, and the seismic retrofit of bridge piers using steel jacketing.

Prior to 1972, concrete bridges were designed using older design codes and standards that did not include requirements for reinforcement detailing to ensure ductile behavior. The current seismic design practice, on the other hand, promotes the use of confinement reinforcement detailing to avoid brittle failure modes such as shear or plastic hinge failure when subjected to load cycles that push the element into the inelastic range of behavior. Bridge columns that were not designed to to avoid these failure modes would not be able to accommodate inelastic displacements when subjected to high seismic forces. In adequare reinfocement detailing of column regions where plastic hinges may form during the seismic event are particularly concerning. With this in mind, seismic retrofit of the bridge columns, such as those on the US-50 corridor, is essential to ensure life safety during an earthquake event.

The seismic retrofit scope of the Multimodal Corridor Enhancement and Rehabilitation (MCER) project included seismic enhancement of bridge columns in two locations on US-50: the Southside Park Viaduct bridge and the 15th–16th Street Viaduct bridge. Both bridges consisted of Cast-in-Place (CIP) box girder superstructures supported on 4-ft diameter columns with a hyperbolic flare at the upper 10 feet. Bridge column heights vary from 19 feet to 29 feet, with about 4 feet of the columns below grade, and supported on pile foundations. Both bridges were initially specified to receive a steel solution that consisted of half-inch thick steel jackets for the full height of the columns, including the top hyperbolic flare portion. The project scope for the Southside Park Viaduct included retrofitting 42 columns, and 24 columns for the 15th–16th Street Viaduct.

Steel Jacket versus FRP Jacketing

A common technique for improving the seismic behavior of columns when retrofitting is by using steel jacketing. Although there are many structural advantages to using steel jacketing, there are also a few disadvantages. The fabrication of special steel shapes to match existing

column geometry, the use of multiple steel pieces to create the jacket, full-depth Complete Joint Penetration (CJP) groove welding, jacket modification to accommodate existing conditions, and grouting of the annular space between the steel jacket and the column are some of the disadvantages of using steel jacketing. As a result, the total cost of materials, fabrication, installation, and quality control of these jackets is typically high. An effective alternative to steel jacketing that can eliminate schedule delays and higher costs, is jacketing with Fiber Reinforced Polymer (FRP).

The use of FRP is gaining popularity in the repair and retrofitting of concrete structures. FRP composites are high-performance materials that can be used for seismic rehabilitation of reinforced concrete bridges, and offer many benefits for seismic retrofit applications. Some of these benefits include:

•High strength-to-weight and stiffness-to-weight ratios

•Material properties that can be tailored for unique applications

•The ease of handling and installing FRP wraps that reduce the project schedule and minimize disruption to traffic

•Corrosion resistance with practically no maintenance costs

•Increased strength and ductility without altering the stiffness of the structure

Economical retrofitting techniques using FRP jackets can be used to improve the seismic performance of bridge columns by increasing plastic hinge confinement, to improve rotation capacity as well as avoiding lap-splice failures. It can also increase the shear resistance of columns.

Pullman, the specialty retrofit contractor, proposed to use FRP jacketing in lieu of the specified steel jacketing. FRP provided an ideal solution for seismic retrofitting of the existing columns considering their non-uniform geometries. Moreover, unlike steel jacketing, FRP can be cut and shaped on-site, and the design can be easily modified for the varying demands of different regions of the columns. Pullman engaged Structural Technologies, a specialty retrofit consultant, to develop a value-engineered option that utilized FRP systems with the main objective of reducing costs while maintaining the project schedule and eliminating challenges associated with steel jacket installation.

FRP Jacket Design Approach

The columns in question have a hyperbolic flare at the upper portion and a circular section at the bottom portion. Seismic guidelines of the California Department of Transportation (CALTRANS) did not address the seismic retrofit of this type of column using FRP jackets. Therefore, the FRP retrofit solution had to be reviewed and approved by CALTRANS’ technical team before being considered as a viable option.

A performance-based approach was used for the design of the FRP jacket. To this end, WSP, The Engineer of Record, provided Structural Technologies with the target drift demands in longitudinal and

transverse directions for the columns. Structural Technologies then developed analytical models to identify the location and length of plastic hinges in order to calculate the required plastic rotation capacity to withstand target drifts. The model consisted of frame elements with pinned support at the base and fixed support at the top. Figure 1a shows existing column geometry, and Figure 1b shows the analytical model D. The locations of the plastic hinge from the non-linear analysis occurred at mid-height of the column. The location of the plastic hinge occurred at the vertical reinforcing bar splice region, at the base of the flared top of the column. The location of the plastic hinge did not correspond to the location of the maximum moment. To address the plastic hinge limited rotation capacity, lap-splice slippage, and shear strength deficiencies, a Lumped Plasticity Model (LPM) was utilized (Priestley et al. /2007). According to this approach, the plastic hinge rotation capacity (��p) and ultimate displacement (∆u) at the tip of a cantilever concrete column can be calculated using an LPM in conjunction with bilinear moment-curvature curves, as described in equations 1a and 1b.

In which, ��y, and ��u, are the sectional curvature at the first effective yield and ultimate, and My and Mu are the moments at the first yield and ultimate condition, respectively. H represents the column height and Lsp and LP are the strain penetration and plastic hinge length. To ensure that the reinforcement at the lap splice region would not slip during the rotation of plastic hinge region, FRP jacket thickness (tFRP) was determined to provide the additional clamping pressure required at the splice considering equation 2.

Where D is the diameter, fl and fh are required and existing clamping pressure, ��FRP is the FRP strength reduction factor, and EFRP is the FRP modulus of elasticity. After calculating the minimum FRP thickness based on the required clamping pressure, the plastic hinge rotation capacity was calculated based on the idealized M-�� curve. The idealized curve was constructed based on an elastic perfectly plastic response and the idealized plastic moment capacity was obtained by balancing the areas between the actual and the idealized M-�� curves as specified in section 5.3 of CALTRANS’ Seismic Design Criteria Version 2.0 (2019). Figures 2a and 2b show the idealized moment curvature and moment rotation of the FRP confined plastic hinge region. The FRP jacket required for the plastic hinge rotation and splice clamping pressure was extended to a length not less than the diameter of the column. A secondary FRP jacket was specified below the plastic hinge jacket to prevent the potential for local failure outside the main plastic hinge region. The secondary FRP jacket was designed to produce half the stiffness of primary plastic hinge jacket and extended half the length of the primary jacket. In this project, FRP layers outside the plastic hinge region were gradually reduced to avoid a sudden change in the confinement pressure. Finally, to avoid brittle shear failure of the enhanced columns and ensure that ductile flexural yield will be achieved, an additional FRP jacket was designed for the lower portion of the column to increase the shear strength of the columns as stipulated in equation 3.

Where �� is the resistance factor for shear considered as 1.0, Vn is the nominal shear capacity of the column, Vc , Vs , and VFRP are shear

Figure 1a Column configuration

Figure 1b Non-linear analysis model

strength of concrete, steel and FRP jacket, and VP is plastic shear corresponding to the idealized plastic moment capacity of the designed M-�� curve. It is worth noting that due to the expected cracking in the plastic hinge region, during a seismic event using the calculated concrete shear contribution was reduced, and the FRP jacket was checked to ensure it can provide the additional shear strength to cover the concrete shear strength losses. Figures 3a and 3b show a typical FRP layout along the column height and installation of the FRP jacket on a bridge column.

The FRP jacket solution was presented to both WSP engineers and CALTRANS technical staff. After addressing technical comments related to FRP detailing, the proposed FRP jacket solution was approved to be utilized to improve the seismic performance of the non-ductile concrete bridge columns.

FRP Jacket Installation

The existing columns extend 4-feet below grade; therefore, excavation was required to access the below grade portion of the columns and the FRP installation process ended up being broken into two phases. In Phase 1, the FRP jacket was installed on the upper portion of columns ahead of the excavation, and in Phase 2, the FRP jacket was installed on the column segment below grade. The following steps were used by the FRP contractor to install the FRP jacket:

1. Concrete surface preparation was achieved using mechanical grinding to open up the concrete pores and grind down existing concrete form lines.. The target surface profile was CSP-3 as in ICRI Guideline 310.2R-2013.

2. Applying epoxy primer coat to the prepared substrate and then using epoxy putty to smooth the surface and fill bug holes.

3. Impregnating the dry FRP sheets with epoxy using mechanical saturators to control the amount of epoxy applied to the fiber sheets and avoid under or over-saturation of the sheets which can lead to dry fiber or sagging.

4. Installation of FRP sheets.

5. Using rib rollers to remove air bubbles after the installation of each FRP layer.

6. After cure of the FRP jacket, applying the topcoat for longterm protection of the jacket.

Post-Construction Inspection

As with every project, it was essential to develop a quality control program to verify the process of the FRP installation and ensure that it can achieve the design properties and performance intent.

Multiple FRP panels were constructed onsite during FRP installation

Figure 2a Idealized plastic hinge moment curvature curve

Figure 2b Idealized moment rotation curve

Figure 3a Designed FRP jacket layout along the column height

and cured on-site daily. A few of the panels were randomly selected and sent to a thirdparty laboratory for tensile testing per ASTM 3039 to evaluate the properties of the final FRP composite product installed on the columns. Visual inspections were used to verify the surface preparation prior to FRP installation. Acoustic sounding tests were performed on the installed FRP jackets to verify bond and check for any air pockets or gaps behind the cured FRP or between FRP layers. In addition, the bonding of the FRP jacket to the concrete substrate was verified by pull-off tests performed per ASTM D7522.

Performance Factors Steel Jacket FRP Jacket Cost

$23,900/Col

$20,800/Col

Durability Requires periodic maintenance Corrosion resistant Flexibility for Odd Shapes Moderate High Installation Schedule Slower compare to FRP jacketFaster compare to steel jacket Fabrication Lead Time 12 – 16 weeks 2 – 3 weeks

Innovative Improvements

The use of conventional construction materials or procedures does not always produce the most technically viable or cost-effective solutions. Utilizing FRP jackets on the US-50 Multimodal

Corridor Enhancement and Rehabilitation Project illustrates how advanced FRP composites can be utilized to improve the seismic performance of concrete structures and provided a costeffective alternative to conventional solutions. Table 1 provides a comparison of a few parameters for the FRP jackets versus steel jackets based on the current project that highlights the benefits of using FRP jacketing.

Seismic retrofit of older bridges and buildings is not the only area where advanced FRP composites can be utilized. FRP composites have been successfully used by the project team to address construction and/or design defects in newly constructed structures, as well as to increase their capacity for change in use, additional gravity loads, or to address conditions resulting from structural modifications such as the installation of openings and penetrations. Lastly, it is very important that FRP installers receive the proper training and undergo 2–5 years of experience before becoming qualified to work on complex projects.

By selecting a favorable installation sequence consisting of 90% of the upper portion of the columns before the excavation to expose the footing, Pullman was able to gain full access and availability of all columns at once by not having to rely on limitations and hazards related to excavation. Pullman, with the technical expertise of Structural Technologies, was able to work collaboratively with all stakeholders to explain the proof of concept and initial design of FRP for CALTRANS’ approval. The design and construction teams provided input for FRP optimization and specifications and supplied and installed a carbon fiber strengthening system that met the needs of the US-50 Multimodal Corridor’s structural needs. The project was completed on schedule and the final project cost was greatly improved through the value-engineering efforts of the project team.■

Full references are included in the online version of the article at STRUCTUREmag.org.

Mohammad Jalalpour, Ph.D., P.E., S.E. is a senior structural engineer with Structural Technologies. He is an accomplished structural engineer and researcher with extensive experience in seismic design and retrofit of structures using conventional techniques as well as fiberreinforced polymer (FRP) systems. He is actively involved in various ACI committees, such as ACI 369, ACI 374, ACI 440, and ACI 562. He can be reached at (mjalalpour@structuraltec.com).

Tarek Alkhrdaji, Ph.D., P.E., P.Eng., FACI is a senior vice president with Structural Technologies. Dr. Alkhrdaji has been involved in over 500 projects, involving structural repair and strengthening – many of which have received awards from ACI, PTI, and ICRI. Dr. Alkhrdaji has authored or co-authored over 40 papers and articles on repair and strengthening of structures. He can be reached at (talkhrdaji@structuraltec.com).

Figure 3b Applied FRP jacket on the site

Table 1 A comparison between FRP jackets and steel jackets

Emerging Risks and Claim Trends in the Design Profession

Survey findings offer insight to potentially mitigate claim risk. Part 2

By Dan Buelow and Mark Blankenship

In our previous article in the June 2023 issue of Structure, we discussed two emerging claims trends identified in the WTW A&E Professional Liability Carrier Survey Report, social inflation and cyber liability. In this issue, we will continue by addressing the impacts of climate change, Covid-19, and risk-shifting in contracts.

Climate Change

Climate change is the one emerging risk that nearly every one of our carriers has on their list. The insurance industry firmly believes in climate change, and there's a lot to think about regarding the exposures facing design professionals. Current codes are based on what we consider to be a faulty assumption: weather conditions will remain static. Recent changes in weather patterns have led to damage claims resulting from too much water (flooding) and insufficient water (wildfire). Drought conditions have also caused ground desiccation, destabilizing building foundations. Even when the design professional met all applicable codes, we have seen claims that the design professional should have anticipated the extreme event and planned for it.

This is especially true about flooding. Flood is not covered under most standard property policies. Flood insurance is primarily a stand-alone policy from the National Flood Insurance Program administered by The Federal Emergency Management Agency (FEMA). Many home and business owners do not buy flood insurance, so they must sue someone else on a liability theory if they suffer a flood loss. And seeing as most municipalities have sovereign immunity, the only parties left to sue are contractors and design professionals who worked on the grading and sewers.

This raises the question: what is the standard of care, and how will that adapt to the new reality of increased frequency and severity of storms? We recommend that clients be presented with options ranging from mere compliance with codes to an adaptive and resilient design. And we want to encourage the client to give more value to the adaptive and resilient design by asking them to look at the cost of it over the life cycle of the building instead of a three- or five-year window return on investment.

The desired result is to cause the owner to place more value on an adaptive and resilient design. But, if we make that offering and document that they declined that offering of a higher level of service to design a better, more resilient structure, then if it is damaged in a flood, a windstorm, or a wildfire, we have a defense. We explained the risks to you, and you decided to accept that risk in exchange for saving money. It is “I told you so.”

The legal concept that we want to obtain is the owner's informed consent. We want to show them the options and explain the risks of merely complying with current codes. And if the owner elects to do the cheaper thing, to save money by implementing the less adaptive and resilient design, then we have transferred the risk of failure to them under the concept of informed consent.

COVID 19

When COVID hit, we wondered what was going to be the impact on design professionals. Initially, we thought it might affect the standard of care by requiring the participation of doctors or other medical professionals in assisting us in the design to try and prevent the spread of viruses. With the development of vaccines, attention has shifted away from disease prevention to the economic impacts of supply chain interruption. Long-tail items are taking even longer to order and obtain, with the result that changes to a design that can lead to delays in completing a project. Some firms are adapting by ordering long-tail items earlier than we would have done historically. This suggests an increased value associated with pre-construction services. We predict that this will become even more valuable shortly, both as a proactive service to the client and as a defensive procedure for design professionals.

This again goes back to the concept of informed consent. Explain to the owner that long-lead systems can only change with a risk of

substantial delay. So, we need to think very carefully about the program and understand that it will delay the project and cause additional expenses to make changes. So we want to get owner approval of major purchase decisions in writing.

To make those informed decisions earlier, meeting with the owner, the contractor, and the major subcontractors to set goals and do the value engineering and the lifecycle cost analysis early on in the process would be beneficial. The goal is establishing procurement management procedures and getting our bid packages together early. But again, all with the understanding that it will be difficult and expensive to make changes once those orders have been placed.

Contracts

About contracts, we often see owners that draft their contracts to attempt to shift risk to the design professional. Sometimes this results in liability assumed in the contract that exceeds the legal liability of the design professional, which is for damages to the extent caused by failure to meet the professional standard of care. This creates a potential coverage issue because every professional liability policy has an exclusion for liability assumed under the contract that would not exist in the absence of the contract.

It is in nobody's best interest to have an uninsurable agreement –especially the owner. Owners often miss this point and lose sight that a design professional’s PL insurance is third-party coverage – meaning that this coverage is for the client. Couple this with the fact that the average design firm has few liquid assets other than their insurance policies, and one has to wonder why the owner would want an uninsurable agreement. Unfortunately, this doesn't stop the flow of onerous and uninsurable agreements that design firms must address. Contractual liability issues usually flow from an increase in the professional standard of care, an assumption of the liability of others, or derogation of due process rights. We want to avoid agreeing to a “highest” or “national” standard of care or giving any guarantees or warranties regarding our service. So, their first priority is to establish an acceptable standard of care.

Assumption of the liability of others usually occurs in the indemnification clause. The duty to defend is probably the leading cause of pain here. Owners' attorneys often need help understanding or appreciating the difference in responsibilities between design professionals and contractors. And they certainly need to understand the differences in insurance coverage. They are used to dealing with contractors whose obligations are related to general negligence and who can add additional insureds to their primary insurance policy, the general liability policy. Additional insureds will secure a defense for covered claims, so contractors agree to defend their clients from third-party claims. In jurisdictions subject to the economic loss rule, design professionals only owe a duty to their clients for purely economic loss, which is the primary type of claim made against design professionals.

Because we can't add additional insureds to professional liability policies, there's no defense obligation coverage relative to a professional liability claim. If the owner insists on a defense obligation, the best solution is a bifurcated indemnity. For claims-based professional liability, we will indemnify the owner, its officers, and employees for claims to the extent caused by our negligence. But we will not defend any of them.

For claims covered by general liability insurance, we can indemnify and defend the owner and other designated parties for damages to the extent caused by our negligence, under additional insured status. This bifurcated approach should satisfy the owner and preserve the insurance coverage available to the design professional.

Due process is implicated in any clause allowing the owner to withhold payment from the design professional without adjudicating that the design firm is liable.

Another contracting trend is indemnification for copyright infringement. This is troublesome because copyright infringement is typically alleged to be an intentional tort. Sometimes this is a bad act where a designer ripped off somebody else's plans, but more commonly, it's accidental. It occurs when a design professional has a dispute with the owner and the owner wants to take their plans and have somebody else finish them, or it's an extension to an existing project. It's not malicious. But we do not normally consider copyright infringement an error or omission. When we see these clauses, we try to revise them to reflect indemnification for negligent copyright infringement to bring that within the scope of professional liability coverage. This exposure can be managed by always creating original plans. The exposure increases if you are asked to finish a design that was started by another firm or when you are asked to match the appearance or style of existing buildings.

In the third and final article of this series, we will discuss the enhanced risks associated with certain project types, the most important thing you can do to preserve your professional liability coverage, and the state of the professional liability marketplace.■

Dan Buelow is Managing Director of Willis A&E. He leads a team of insurance and risk management experts that are exclusively dedicated to providing insurance and risk management solutions to Architects and Engineers. Dan can be reached at (Dan.Buelow@wtwco.com) or 312-288-7189.

Mark Blankenship is the Director of Risk Management for Willis A&E. He draws on his 25 years of experience in professional liability claims, underwriting and brokerage to provide risk management guidance, claims advocacy and contract review support. Mark can be reached at (Mark. Blankenship@wtwco.com) or 312-525-2281.

structural QUALITY

FEMA P-58

The preferred alternative for seismic due-diligence.

By R.O. Hamburger, S.E., C.B. Haselton, Ph.D., P.E., and D.J. DeBock Ph.D., P.E.

It is impossible to predict precisely the damage a building will experience in a future earthquake, because each earthquake is unique, and building simulations only model approximate responses to shaking. Losses due to earthquakes are even more unpredictable because they depend on the post-eartahquake actions of individual owners, insurers, design professionals, building officials, and contractors. Nevertheless, property investors need to understand the risk associated with property investments. Seismic risk assessment reports have become standard practice in financial due diligence.

Building owners, investors, lenders, and insurers use Probable Maximum Loss (PML), Scenario Expected Loss (SEL), and Scenario Upper Loss (SUL) as convenient means of characterizing earthquake risk for buildings. Financial decisions, including whether to purchase, finance, or insure buildings and under what terms, are often made based on the assessed PML/SEL/SUL values. However, only some decision-makers understand what these values represent, how they are derived, and the wide variation in validity of values obtained using different methods.

Historic Development

Today’s SEL/SUL parameters evolved from the Maximum Foreseeable Loss (MFL) parameter developed by the fire insurance industry. MFL represented the likely maximum repair cost for a building for a worst-case fire. Insurers reasoned that MFL was not useful for earthquakes, fearing that worst-case loss for a worst-case earthquake for any building would be 100%. Instead, the industry adopted PML, defined by K. V. Steinbrugge as “the expected maximum percentage monetary loss which will not be exceeded for 9 out of 10 buildings in a given earthquake building class, when subjected to vibratory shaking from a maximum probable earthquake.” PML is applied to buildings on firm soils, not subject to landslide or direct fault rupture, located within a 30-mile-wide strip centered on and along the length of the causative earthquake fault.

Steinbrugge suggested that in a great earthquake, a few risks will exceed the PML; a greater number will be at or near the PML, but the majority will be less than the PML. He attributed this variation to differences in shaking intensities within the zone, building configuration, construction quality, and condition. Steinbrugge presented suggested PML values for buildings of different classes, with modification factors associated with shaking intensity, soil type, location relative to the fault trace, and other factors based on his personal observations and insurance industry data.

In the 1980s, commercial property lenders concerned about earthquake risks in California began to require PML reports as a pre-condition for loan approval. The industry adopted a modified definition of PML as the 90th percentile building repair cost as a fraction of building replacement cost resulting from 475-year earthquake shaking. The PML did not include loss of use, soft costs, such as

design and permitting fees, damage associated with tenant furnishings and contents, or post-earthquake fire. The 475-year earthquake was selected for compatibility with 1980s-era building code requirements. Lenders would generally offer loans on buildings with PMLs of 20% or less. The 20% level was thought to represent a repair cost that most owners would undertake rather than abandoning the property. Soon lenders found these criteria too limiting, leaving few buildings eligible for loans. So, they revised their criteria by accepting mean, rather than 90th percentile estimates, and sometimes by specifying PML for more frequent, less intense earthquakes. Ultimately, the mean loss estimate became known as Scenario Expected Loss (SEL), while the 90th percentile loss became known as the Scenario Upper Loss (SUL).

In the late 1990s, the American Society for Testing and Materials (ASTM) published standards for producing PML reports. ASTM E2557 and ASTM E2026 set criteria for the level of evaluation performed, evaluator qualifications, and standard terminology but do not specify how to estimate the loss. Evaluators currently use many methods, including the public domain models shown in Figure 1 or proprietary models such as ST-Risk, SeismiCat, and the SP3-RiskModel to predict loss.

ATC-13 Method

In 1985, the Applied Technology Council, supported by the Federal Emergency Management Agency (FEMA), published ATC-13 Earthquake Damage Evaluation Data for California. This publication intended to provide government planners with a tool to assess potential regional losses following a major earthquake. However, it is used by some PML providers today for individual building evaluations and forms the basis of some proprietary methods.

Figure 1 Seismic Due-Diligence Methods

The ATC-13 report identifies building class archetypes, e.g. wood light-frame, lowrise masonry shear wall, etc. The report presents a damage loss matrix for each archetype indicating earthquake damage repair cost ratio as a function of ground shaking intensity using the Modified Mercalli Intensity (MMI) scale and probability of exceedance. This methodology has no direct way to account for individual building characteristics that could lead to increased or decreased vulnerability of a particular building relative to the archetype model. Engineers using this method must exercise individual judgment to account for building-specific vulnerabilities, such as soft stories or torsional configuration. Unlike Steinbrugge’s tables, the ATC-13 loss matrices represent the statistical distribution of expert opinion on probable losses rather than the actual losses experienced by such buildings. The report, published in 1985, does not account for building vulnerabilities discovered in later earthquakes (e.g., the vulnerability of welded steel moment frame connections to brittle fracture). The ATC-13 judgment-based tables also do not reflect building design improvements made in the last 40 years, such as enhanced building code provisions and new structural systems. Furthermore, the MMI scale used in the methodology to characterize the strength of earthquake shaking is a dated and imprecise measure of ground motion.

Thiel-Zsutty (TZ) Method

In 1987, Charles C. Thiel Jr. and Theodore Zsutty published the TZ Method. The TZ Method depends on parameters representing the building's structural system, similarity of the site and building periods, the site materials, and the site's acceleration. The TZ Method utilizes the ATC-13 building class archetypes. Engineers using the TZ methodology will typically use engineering judgment to adjust values obtained from the defined building system parameters to account for building-specific characteristics.

In 2016, Thiel and Zsutty reformulated the distribution function and extended the TZ Method to reconcile the TZ and ATC-13 methods. While the TZ Method is widely used in due diligence assessment, the methodology needs to reflect the many advancements made in our understanding of building seismic performance over the last 35 years.

HAZUS Method

FEMA sponsored the Hazus project to develop guidelines and procedures for estimating regional (not building-specific) earthquake loss. An Advanced Engineering Building Module (AEBM) is available for individual buildings based on simplified single-degree-of-freedom

analyses. Due to these limitations, Hazus is not typically used for seismic due diligence of individual buildings.

FEMA P-58 Methodology for Building-Specific Seismic Risk Assessment

FEMA P-58 was developed by the Applied Technology Council (ATC) under Federal Emergency Management Agency (FEMA) funding to address the insufficiencies of the existing seismic risk assessment methods. Development involved over a hundred researchers and industry professionals; this new method incorporates earthquake engineering research and computational advancements, applying statistically rigorous methods to characterize loss and has been calibrated against loss data from the 1994 Northridge earthquake. Development was completed in 2012, with some extensions and refinements published in 2018.

The FEMA P-58 Method can be used for the performancebased design of new buildings, retrofit of existing buildings, and for building-specific seismic risk assessments. The method starts with a building performance model (Figure 2), which defines all damageable structural and non-structural building components. A fragility function database of nearly 1,000 components quantifies the damageability of each component based on experimental data and research studies. The use of building-specific, component-level damage estimation enables better risk assessment of individual assets and more informed design decisions. The method uses Monte Carlo simulation and provides building-specific estimates of both mean losses (SEL) and 90th percentile losses (SUL). The method also predicts building recovery times including time to re-occupancy, time to regain building function, time to full repair, and red tagging probability.

Figure 3 provides some examples of FEMA P-58 output data for a well-performing 5-story concrete shear wall building in Los Angeles. For this site, the 10% in 50 year peak ground acceleration (PGA) is 0.52g and the Maximum Considered Earthquake (MCE) motion is 1.06g.

Figure 2 Building Performance Model

Selecting a Method to use for Seismic Due-Diligence Studies

PML providers considering selecting a methodology should understand how the results between methods compare. The authors performed a comparison study between FEMA P-58 and historical methods for approximately 2,000 sample buildings. Figure 4a shows some sample results of this study for Pre-Northridge steel frames, comparing FEMA P-58 and TZ results. This study found that, on average, FEMA P-58 repair costs are similar to those of historical methods for high-seismic sites and are consistently lower for lower-seismic sites. However, the results for individual buildings vary widely, because FEMA P-58’s performance model includes more building-specific and site-specific detail. Figure 4b shows how the more accurate FEMA P-58 building-specific losses are partly explained by the use of Sa(T1) as a better ground motion intensity measure, and use of a calculated site-specific base shear strength in the building performance model.

Summary and Recommendations

Historic loss estimation methods were all developed to assess the risk for classes of buildings instead of individual buildings. These class-based methods rely on judgment to reflect individual building characteristics. This approach creates two main problems – (a) PML loss ratios are based on “average” buildings and are not very building-specific (even after judgment is added to change the results), and (b) PML results are not consistent and repeatable across providers.

The FEMA P-58 analysis method was created to provide buildingspecific seismic risk analysis for repair cost and recovery times and eliminates the inaccuracy associated with building class-based models. Because it is grounded in engineering analysis rather than primarily being based on judgment, FEMA P-58 results are consistent and repeatable across providers.

The FEMA P-58 analysis method was introduced in 2012, and commercial use of the method started in 2014. The method now covers all common building types and heights. The results have been validated through comparisons to earthquake data and other historical methods and have been used and vetted within the structural engineering profession for the past decade. As of 2023, no better building-specific seismic risk assessment method is available. Moving from the classbased methods of the 1980s to the current building-specific FEMA P-58 Method will enable commercial lenders, investors, insurers, owners, and other stakeholders to make more informed decisions regarding resiliency and risk.■

Figure 3 FEMA P-58 Building Repair Cost Ratio Outputs and Building Recovery Times for Business Disruption

Ronald O. Hamburger, S.E., SECB is a Consulting Principal at Simpson Gumpertz & Heger (rohamburger@sgh.com).

Curt B. Haselton, Ph.D., P.E. is a Professor of Civil Engineering at Cal. State University Chico and is Co-Founder/CEO of Haselton Baker Risk Group (chaselton@csuchico.edu)

D. Jared DeBock, Ph.D., P.E. is the Department Chair of Civil Engineering at Cal. State University Chico and is a Senior Associate at Haselton Baker Risk Group (ddebock@csuchico.edu).

Figure 4 Comparison of FEMA P-58 versus TZ Method Results for Pre-Northridge Steel Frames

Leadership and Mentorship

A step toward saving the profession.

By John A. Dal Pino, Claremont Engineers, Inc.

Earlier this year, I wrote an article on the apparent shortage of structural engineers and what caused it to happen, if, in fact, it has. I identified many contributing factors that made the employment situation worse than it ought to be. But since the employment market is a dynamic eco-system, with many moving parts that change over time, for reasons we only see in hindsight, finding “the solution” or “a handful of possible solutions” is nearly impossible. I concluded, however, that if a shortage exists, then it was something that engineers and educators did to themselves, with more than a bit of help from government policies enacted by the officials we elected.

A survey by the SE3 Committee of the Structural Engineers Association of Northern California (SEAONC) generated data from working professional engineers on career satisfaction, career development, pay and benefits, worklife balance, etc. Key findings included the following:

• Eight in ten respondents reported being either “satisfied” or “very satisfied” with their career overall. •56% of respondents had considered leaving the profession, the top reasons for which were seeking higher pay, better work-life balance, and less stress.

• Respondents who had left the profession noted poor management/ leadership as one of the leading reasons they had left, suggesting that overall, poor management/leadership is a core cause for people leaving the structural engineering profession.

•Employees who worked more hours each week were more likely to have considered leaving the profession.

• Respondents overall indicated that pay/compensation was the top reason they had considered leaving the structural engineering profession and, for those who had left the profession, one of the leading reasons why they had left. Male respondents earned considerably more than female respondents, except for entry-level respondents.

Data can be skewed by who responded to the survey. Still, the overall picture for the future of engineering doesn’t look so good to me: 20% of employees are less than satisfied with their jobs, a little more than half have considered leaving the profession entirely, people believe they are over-worked and/or underpaid, and most importantly, they believe that management/leadership is poor. Ouch! It doesn’t matter if these impressions are true, since there is always the other side of the coin,

but if the engineering staff believes it, then it needs to be dealt with. But to be fair, employers need to earn a profit to survive and grow in a competitive marketplace where forces largely out of anyone’s control set fees and salaries. Not surprising, not everyone is cut out for an often challenging and pressure packed career in structural engineering. COVID and the lasting and destructive impacts on the workplace didn’t help either.

So what are we as a profession collectively going to do to improve employee retention and general happiness? I suggest stepping up leadership and mentorship at all levels would be a good start.

It should come as no surprise that people like working with people they enjoy being with, people whom they respect, people who are engrossed in their work, are having fun, are willing to teach and inspire, and by their own careers and professional involvement show a path that others can emulate. Part of being a great leader or mentor is working harder and smarter than everyone else and showing how it is done through action. It should be clear to leaders and mentors that the word is out on your firm if staff is leaving or it is hard to recruit. No one said it would be easy to save the profession.

So to start, the firm needs to focus on its culture. One could argue that this may be the most critical function of leadership. It can be a positive reinforcing process that motivates or a vicious downward circle that depresses and creates negativity, depending on how it is

The construction site -- A great place to learn.

done. If the results of the SE3 survey are a reflection on the entire industry, we aren’t doing a very good job on the people part.

Unfortunately, most leaders/mentors have yet to receive formal training. Therefore, they have established company cultures somewhat through trial and error, what they observed and perhaps personally experienced within their firm or elsewhere, not always for good either. One could also argue that great leadership and/ or mentorship skills can’t be taught, and the culture built will likely reflect their own personalities. You either are a great leader or mentor, or you are not. Starting your own firm or being the most aggressive person in the room isn’t a qualifier.

A trend that has emerged over the past several years is for “speed mentoring” and “arranged mentoring.” Speed mentoring is akin to online dating. Collect a group of people in a room, generate conversations, and hope that mentorship relationships develop. Arranged mentoring is facilitated by eliciting interests and experiences and setting up mentoring relationships, somewhat comparable to arranged marriages by the village elders in the old days. If this has worked for you, that is fantastic, but this is a very passive approach with a low probability of success.

I suggest there is a better, more active way. Be constantly observant, seek knowledge, and approach each interaction as a learning opportunity. Many people have influenced me over the years and I learned quite a bit. But for every good thing I learned, I learned some valuable lessons about what not to do or how not to act. And those lessons, good and bad, happen all the time, 24/7/365. I realized quickly that no one is perfect, and searching for a saintly mentor who will dispense from on high that knowledge you seek in terms of learning, growing, and being successful, may

be disappointing. I want to suggest that those people don’t exist, or if they do exist, there aren’t many of them. If you found one, that is fantastic.

If creating a positive leadership and mentorship culture wasn’t hard enough, the really hard part is making it a natural and enduring part of your organization. Patting yourself on the back at the annual meeting or the holiday party and then forgetting until the “next time” doesn’t make it so. The staff will see through this, and their silence will speak volumes.

For those concerned about the future of engineering, look inward first and make your firm the kind of rewarding place that people will enjoy working at and that you would find attractive to work at if you were considering an engineering career. If you bring joy and excitement to your work, lead by example, treat people fairly, show interest in them, make every day an opportunity for teaching and learning, and help staff grow personally and professionally and succeed, there won’t be a shortage of engineers. If you treat your engineers as easily replaceable, abuse the employment relationship, put your success above that of others, focus on short-term results above long-term investment, or don’t think you are being watched closely, there will be a shortage of engineers. And the deficit will start at your firm first. Put yourself in the other person’s shoes. The firms that do it well will succeed. The others will fade away, as they should.■

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Have a direct impact on firm development (in more than just project work). We believe in the benefits of a broad employee-ownership base that crosses business units, tenures + experience levels – 60+% of our team members are stockholders. schaefer-inc.com/

John A. Dal Pino is a Principal with Claremont Engineers, Inc. in Oakland, California. He serves as the Chair of the STRUCTURE Editorial Board (jdalpino@claremontengineers.com).

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HIRING and RECRUITMENT

Inclusive Recruitment and Hiring Practices

Actionable steps to improving diversity and inclusion in the hiring process.

By Angie Sommer, S.E. and the NCSEA SE3 Committee

As conversations about diversity and inclusion occur in board rooms and break rooms of structural engineering firms nationwide, many professionals ask, what can we do to obtain and retain a more diverse workforce?

The NCSEA Structural Engineering Engagement and Equity (SE3) Committee has been working to understand and improve the dayto-day experiences and career trajectories for structural engineers via comprehensive nationwide surveys, data analysis, and a variety of other initiatives since 2015. Addressing recruitment and hiring practices has been one recent focus of the committee, which was explored in a 2022 SE3 publication entitled “Inclusive Recruitment & Hiring Practices” that can be found on the SE3 Committee website: https://www.se3committee.com/publications. This article explores some of the ideas in this SE3 publication, including actionable steps to improving inclusion in recruitment and hiring. The 2020 SE3 survey included responses from over 5,000 structural engineering professionals from all 50 states. Of this group:

• 24% of total respondents (but only 10% of principals) were women

• 23% of total respondents (but only 14% of principals) were nonwhite or mixed race

•65% of women report having considered leaving the profession, compared to only 51% of men

• Women and minority racial groups were 1.5 to 19 times more likely than white men to experience discrimination and/or harassment (Figure 1), and respondents who had experienced discrimination or harassment were 20% more likely to have considered leaving the profession.

Topic briefs that include this data can be found at the SE3 committee website noted previously. While this data isn’t necessarily comprehensive, these findings point to an overall lack of diversity (and likely a lack of inclusion) in the structural engineering profession, especially at higher levels of leadership. This, coupled with both the “business case” (the argument that diversity benefits a company’s bottom line) and the “fairness case” (the argument that justifies diversity on moral grounds), have prompted many firms to explore how they can improve recruitment and hiring practices to be more inclusive.

The Business Case for Diversity

In recent years, there have been many studies and subsequent articles that cite statistics about how companies with staff — and especially leadership — from diverse backgrounds benefit financially over those with less diversity. This is often cited to be due to the increased innovation and creativity infused by people who think in different ways from an otherwise homogenous group. Clients may also relate better to staff at a company who are like them, so having staff from different backgrounds could appeal to a larger pool of clients. In fact, some public jurisdictions are starting to have requirements (or at least bonus points) for diversity on teams that submit for RFPs.

One robust study of the business case for diversity is a McKinsey & Company report entitled “Diversity Wins: How Inclusion Matters.” This report is broad, studying the financial performance of over 1,000 large companies in 15 countries. While not specific

Figure 1 Odds of Discrimination and Harassment by Race and Gender Demographics from the 2020 NCSEA SE3 survey data.

to structural engineering, the data is compelling. The 2019 analysis found that:

1)Companies in the top quartile for gender diversity on executive teams were 25 percent more likely to have above-average profitability than companies in the fourth quartile, and

2)Companies in the top quartile for ethnic and cultural diversity on executive teams outperformed those in the fourth quartile by 36 percent in profitability.

While the “business case for diversity” remains an important rhetoric to initiate diversity discussions in corporate settings, recent research indicates that specifically citing this reasoning within job ads or on a company’s website can alienate potential candidates from diverse backgrounds, as noted in the Harvard Business Review (HBR) article entitled “Stop Making the Business Case for Diversity.” This article suggests instead to either cite the fairness case for diversity on outward-facing materials, or, better yet, state diversity as a value without justifying why you believe it to be important, in the same way you would not justify other values such as innovation, resilience, and integrity.

Assessing Goals

Improving inclusion in recruitment and hiring practices begins with assessing goals. What is the purpose of this effort for your company? What do you value about diversity — the business case, the fairness case, or something else? What types of people and perspectives do you have in abundance, and which are you lacking? Consider that diversity is not only about race or ethnicity; there are many identities that can provide different points of view, such as sexual identity, religious affiliation, country of origin, educational background, age, and ability/disability status.

Rethink Your Job Postings

After creating goals, review how your job postings are written. Certain wording can evoke gendered or other slanted imagery that can prevent certain candidates from feeling a connection with the position. For example, a few potential phrases that slant masculine are noted below, with suggestions for more neutral wording.

More Masculine

Strong communication skills

Ability to perform in a competitive environment

More Neutral

Proficient oral and written skills

Collaborates well in a team environment

Superior ability to satisfy clientsSensitive to client needs, can develop positive relationships

Additionally, easing some of the minimum requirements to apply for the job (e.g., a master’s degree is preferred but not required) or allowing remote work can appeal to different types of qualified people. Consider specifically, as originally found in a Hewlett Packard report and noted in the book Lean In among a variety of other articles, that women are significantly less likely than men to apply for a job for which they do not meet all the stated requirements. In the HBR article entitled “Why Women Don’t Apply for Jobs Unless They’re 100% Qualified,” they note that this isn’t due to a lack of confidence exactly, but rather that women and girls are rigorously socialized to follow rules, while boys and men are often taught to view rules more as guidelines in which

other factors also play a role. These differing approaches can have a significant effect on one’s entry into the job market.

Thus, while it is important to find qualified candidates and pose your job ads to do so, it is also important to think about how certain concepts are phrased and which skills are required versus merely preferred. This way, good candidates don’t opt out of the process before you have a chance to consider them.

Review Your Recruitment Practices

It is important to understand that where and how you recruit has as much of an impact as who you recruit. If you send the same staff to the same colleges year after year, it seems reasonable that you recruit similar people. Evaluate where your company attends recruiting fairs and find opportunities to branch out to different locations. Virtual recruiting fairs can help you reach a wider variety of locations without investing significant additional resources.

Sending staff from diverse backgrounds (of different ethnicities, genders, ages, positions, etc.) to recruitment fairs is another way to attract candidates from diverse backgrounds. People are naturally attracted to those who are similar to them; having a diverse group of people meaningfully involved in your recruitment and interviewing efforts can help yield more diverse hires.

By including a diverse group in your hiring process, you also showcase who is successful and valued at your firm, no matter their age or position. Note that buy-in is key; the staff involved in this process must be valued decision-makers, otherwise you risk tokenizing those who are involved, which can undermine your efforts. Additionally, be aware of overextending certain people who may be one of only a few members of a particular group at your company; their participation in the recruitment and hiring process must be valued and rewarded as much as other tasks that promote the company’s well-being.

Minimize Unconscious Bias

Interactions with candidates affect which candidates are offered, and accept, positions at your company. Understanding and minimizing bias can help reduce unintended actions and consequences of natural biases. Three types of bias that can affect the recruitment and hiring processes are confirmation bias, affinity bias, and gender and race biases.

Confirmation bias Affinity bias Gender and race biases

The tendency to interpret new evidence as confirmation of one's existing beliefs or theories.

A preference for people who are similar to or have the same qualities as ourselves.

When a person faces unfair disadvantages (or benefits from unearned advantages) because of their gender or race.

Basic bias awareness can be achieved with minimal effort through simple online research. Searching the phrase “bias awareness” on the internet, for example, will yield a host of articles describing what bias is and how everyone has biases based on the way the human brain works and how we are socialized. Acknowledging that we all have biases that may cause us to act in certain ways, even when our intentions are good, is the first step to consciously minimizing the effects of our biases.

Doing the work of addressing bias in a professional setting requires discussion, reflection, and revision. As Professor Iris Bohnet from

Harvard Kennedy School puts it, “While it’s exceedingly difficult to remove bias from an individual, it’s possible to design organizations in ways that make it harder for biased minds to skew judgment.” Professional coaches and classes are available to train staff about how to limit the effect of bias in the workplace.

Establish a Formalized Interview Process

Establishing a consistent interview process is a way to get as close to fairness as possible. An example of a robust and consistent interview process might include:

1)A pre-interview internal kick-off meeting in which roles, timelines, and process are discussed;

2)Candidate interviews, which may involve a shorter, more casual interview, followed by a technical quiz or test, then a more formal interview with a different person or people; and

3)A post-interview recap with the internal recruitment team to discuss a candidate’s ability to perform the required job functions.

In addition to establishing a consistent and clear internal process, interviews with candidates should be structured to create consistency. Research shows that unstructured interviews that lack defined questions and instead allow a candidate’s skills to unfold organically through conversation do not reliably predict job success. Use the same set of questions in the same order for each candidate, allowing each question to be well-thought-out and vetted instead of presenting topics on the fly.

Conclusions and Actions

There are many steps a company can take to improve inclusion in recruitment and hiring practices. This process requires time, reflection, revision, and dedication. Do not be discouraged if the first effort does not yield the intended results; take the opportunity to learn something from the experience and adjust to do better next time. Organizational changes do not happen overnight. Be patient, be persistent, and track your progress over time to ensure you are making progress toward your goals.■

Full references are included in the online version of the article at STRUCTUREmag.org.

Angie Sommer, S.E., is a principal at ZFA Structural Engineers in San Francisco, California, who has been a member of both the SEAONC and NCSEA SE3 Committees since their inception. She has co-chaired both committees and led the efforts on multiple SE3 initiatives and publications. (angies@zfa.com)

This article was written in collaboration with, and supported by data from, the NCSEA SE3 Committee. The mission of the SE3 Committee is to raise awareness and promote dialogue on professional practice issues to improve engagement and equity in the structural engineering profession. (www.se3committee.com)

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Flexible Diaphragm Deflection

Wood diaphragm deflection compatibility.

By George Ghusn, Jr., P.E., S.E.

The conventional approach to diaphragm deflection has always been puzzling. Like many other practicing engineers, the author put his reservations aside and calculated as the code demanded.

In a recent webinar on large diaphragm design, the lecturer casually remarked that something is wrong with the diaphragm deflection equation. It was like a side quest hint in an online game. Quest accepted! Thus began a dive into the calculation of diaphragm deflection to understand the equations mandated by the building code.

In all of structural engineering, there are only two principles: Equilibrium of forces (or energy for methods like virtual work) and Compatibility -that is, a single point on a structure has one deflection under load, and all the members that connect to that point must have that same deflection at that point. The principles are simple, but the devil is in the details, and they can be very complex.

The fundamental problem with diaphragm deflection as we calculate it per code is a violation of compatibility. The diaphragm has a single deflection, but we do not force the individual components to have the same deflection.

The calculation of flexible diaphragm deflection is specified in the American Wood Council (AWC) 2015 Special Design Provisions for Wind and Seismic (SDPWS) as:

consequences if you “close the loop” on the calculations.

To illustrate, take the simple example of a flexible diaphragm with dimensions and loading as shown. The diaphragm will be the same thickness and nailing throughout to make the example close to the assumptions in the equations. The loads are based on 15 pounds per square foot of roof dead load, a total panel height of 30 feet, and a panel weight of 100 pounds per square foot. Using a Cs value of 0.2 for the 100 × 200-foot diaphragm shown in Figure 1, the calculated diaphragm load v is 900 pounds per linear foot.

This equation and approach have been around since ATC 7 was published in 1981. The first term is bending deflection based on a simple beam model using isolated top and bottom chords. The second term combines the apparent shear stiffness of the diaphragm and the nail slip combined. The “split” version of this term is still contained in the 2018 and 2021 International Building Code (IBC), where diaphragm material shear deflection and staple slip deflection are separate terms. The split version for nailed diaphragms appears in the SDPWS Commentary. The third term is a deflection contribution from the slip at chord splices, typically taken as zero for welded steel chords. From a practical view, the summation of deflections doesn’t make sense. If one considers the diaphragm system as two springs (neglecting the slip term), then this summation would only make sense if the two springs can deflect independently, allowing each spring to only experience its own calculated deflection. This is a “springs in series” approach to calculating overall deflection. However, the diaphragm only has a single deflection, and both components must have the same deflection. This creates interesting

To determine an area for the chords, use the typical approach of taking the maximum moment and resisting it as a couple in the chords. In this case, we will assume that it is an angle or channel ledger using ASTM A36 steel:

= 45,000 lb.Area of chord required: =1.38 sq inches,use 1.5 sq inches.

The bending portion of the deflection is calculated using the first term:

= 1.03ʺ

The design shear is divided by a φ factor of 0.8 to get a table value of 1125 p ounds per foot (SDPWS 4.2.3). Using SDPWS Table 4.2A, using blocked 15/32ʺ ordinary sheathing on 1-1/2inch framing and 10d nails at 4 inches on center (Case 1, 2, 3 & 4) gives a shear capacity of 1150 pounds per foot and a Ga value of 21 kips per inch. The shear portion of the deflection is calculated using the second term:

= 2.14ʺ

Figure 1 An Example Diaphragm

Thus, the total deflection, per code, is 1.03 + 2.14 = 3.17 inches. Typically, the design stops here. The engineer compares the calculated deflection to specified limits and moves on. However, what happens when one calculates the required force to make the chords experience this additional deflection?

Re-arranging the first term of the deflection equation and solving for v using the calculated total deflection gives:

= 2757 plf

Case Webs Chords

1 – soft web, soft chord

2 – soft web, stiff chord

3 – stiff web, soft chord

4 – stiff web, stiff chord

This gives a chord force of:

= 137,837 lb. > 45,000 lb from statics.

There is a disconnect (literally) between the deflection calculated in the diaphragm and the forces required to generate that deflection. The shear deflection will also require additional force to match the total deflection. This example has a very rigid diaphragm because of the uniform nailing. In a more realistic setting, where the diaphragm capacity and rigidity are reduced across the length of the diaphragm, the apparent chord force would be even larger as the calculated total deflection increases.

One could consider this a relative rigidity problem – that the force to the two springs would be proportioned based on their relative rigidity and sum up the total deflection. This approach suffers because each of the springs must, by statics, carry the forces assigned. Thus, making a few trial runs made it apparent that this was not the correct approach. Examination of beam shear deflection quickly led into a rabbit hole of theories and differential equations that have been developed or reported. The short answer here is that no one understands beam shear deflection completely and that the equation used for diaphragm shear is likely a decent approximation. This result is left as an exercise for the reader. Good luck.

After the unhelpful review of beam shear equations, the author decided to look back at the nominal approach for diaphragms: trusses. The diaphragm idealization is often referred to as similar to a truss. Indeed, the chord force is based on a truss (more or less) with a non-existent web. So, let’s look at an admittedly odd truss to see how well this assumption works out.

The model truss, illustrated in Figure 2, is intended to mimic a diaphragm ratio of 2:1, similar to the example diaphragm. In this case, the truss has a 40-foot span and a 20-foot depth. The truss is composed of 40 panels (one per foot of span) to minimize the edge effects of the modeling. The truss webs are composed of all pin-connected members. There are four cases for this geometry:

HSS2x2x1/8, E = 1800 ksi

HSS2x2x1/8, E = 1800 ksi

HSS2x2x1/8, E = 29000 ksi

HSS2x2x1/8, E = 29000 ksi

HSS2x2x1/8, E = 29000ksi

HSS6x6x1/2, E = 29000 ksi

HSS2x2x1/8, E = 29000ksi

HSS6x6x1/2, E = 29000 ksi

Why use wood modulus values for steel sections? To easily see the effect of the web and chord stiffness on total deflection. This truss example is just to examine the stiffness parameters’ effects on deflections. Yes, these webs would buckle, and there would be some serious geometric nonlinearities if they were real. This is a mathematical model for testing an approach. Always keep in mind the Box Admonition:

“All models are wrong; some models are useful.” – Statistician George Box.

Figure 2 Truss Model for Parametric Study of Chord and Web Stiffness Effects

The theoretical axial chord force for this truss geometry is:

= 10,000 lb or 10 kips

The theoretical chord force value is a little larger than the truss models show. The theoretical chord force can be interpreted as the upper limit of a truss with an infinite number of panels or, say, a diaphragm. There is an edge effect as some of the uniform load “bypasses” the truss in the end diagonals. If this were a twopanel truss, the results would be much different and would not approximate a diaphragm. Not only a bad model for our purposes but a useless one.

Note that the top and bottom chords carry very similar force for all cases regardless of their stiffness or the stiffness of the webs. This is consistent with the statics of the geometry. The results also show that changing chord stiffness has no significant deflection component for the overall diaphragm deflection.

The chord force is based on geometry alone. Thus, the maximum chord force is constant for any given load and truss or diaphragm geometry. Since this force is constant across combinations of stiffness, the total axial strain in the chord is also constant. This strain is horizontal, perpendicular to the applied load, as illustrated in Figure 3. No vertical deflection is developed in the chord from the applied load based on statics alone. The vertical deflection comes from the equilibrium of energy by setting internal strain energy equal to the work performed by the external forces.

The chord vertical deflection is calculated by equating the external work to the strain energy in just the chords. Then one calculates

the external work energy to match the internal strain energy. This “virtual work” allows one to then determine a shear deflection. Add the two components together and one has the total deflection. This is the rationale behind the current code equation. This also aligns with the results of Timoshenko’s Beam Theory, where shear deformation adds to the overall deflection due to an increase in internal strain energy.

The truss analyses presented here indicate that the original approach does not apply. The two deflection components are interdependent, and the chords have a limited force and strain based on geometry alone.

Figure 4 shows that the top and bottom chord paths must be different because their lengths are different, yet they have the same maximum vertical deflection. This creates a complication for the external work calculation because the distances “traveled” by the external forces are not the same top and bottom.

The results for the truss cases suggest that there is no single equilibrium vertical deflection state for the chord. As long as the curved path of the chord can accommodate the total axial length of the chord, any path will be acceptable, provided there is something to carry the perpendicular component of the axial force that allows the chord to conform to the path as shown in Figure 5

Essentially, the shear deflection of the webs pushes or pulls the chord out of a horizontal orientation. This movement results in a perpendicular component of the chord force that the webs must resist. This does NOT increase the force or strain in the chord because the ends of chord are not sufficiently restrained (this is NOT a catenary). This component force acts opposite to the shear deflection and slightly reduces the shear strain energy. The deflection of the chord and web is the same. Equilibrium and compatibility are satisfied.

This result also implies that chord slip will have little effect on the ultimate deflection of the diaphragm – it simply changes the final length of the chord. The chord-produced reduction in shear strain will be less. As the chord slip becomes great or the chord becomes very small, the maximum deflection will approach the web-portion (shear) deflection alone.

Great, let’s eliminate chords! Wait, not so fast. Although they contribute little to deflections, they still resist the edge force that would otherwise need to be resisted at the edge of the diaphragm by statics.

What about Timoshenko? A truss or diaphragm is not a beam and is not modeled well by Timoshenko’s equations or approach. It doesn’t work so well with Euler’s model, either – which is the basis of the chord deflection equation.

Interestingly, the metal deck industry already incorporates this approach into their design deflection equation which only has a shear component. Their rationale is that the chord effect is small because metal deck diaphragms are much stiffer than wood diaphragms. Based on this little example, that’s not true – the chord deflection contribution is small, no matter how stiff the diaphragm. Right answer, poor justification. So, we could use just one equation to represent the diaphragm deflection for both wood and steel diaphragms. Wouldn’t that be nice?■

Figure 3 Axial Strains in Chords

Figure 5 The Perpendicular Force Component for a Curved Chord Path

Figure 4 Chord Path Illustration

George is a principal and president of BJG Architecture & Engineering in Reno Nevada. He has been a practicing structural engineer for 30 years. He can be reached at: (gghusn@bjginc.com)

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Built among the rubble of the 1906 San Francisco earthquake, The Bank of California Banking Hall, located at 400 California Street, was among the first commercial buildings to be completed following the 7.9 magnitude quake – the deadliest natural disaster in California’s history. Constructed of steel, concrete, and granite, plans for the Banking Hall at 400 California Street had been completed before the quake, allowing it to be a keystone of the rebirth of the financial district. Almost 60 years later, the new Bank of California headquarters was built directly adjacent to the Banking Hall. Located at 430 California Street, the new headquarters building (the Tower) is a modern, 21-story steel-and-concrete high-rise tower, structurally separated but functionally connected to the Banking Hall.

On the afternoon of October 17, 1989, California shook once again. Originating from the San Andreas fault with an epicenter approximately 70 miles south of San Francisco, the 6.9 magnitude Loma Prieta earthquake caused substantial damage throughout the greater San Francisco Bay Area. Among the buildings damaged,

400 & 430 California Street Making a Historic Connection

By Kirk Johnston, S.E. and Yvonne Tsui, S.E.

large cracks emerged in one of the corner piers of the Banking Hall. While the Tower did not show any signs of damage, it was believed that damage to the Banking Hall resulted from the two buildings colliding during the quake.

The sale of the two buildings in 2016 set in motion a series of renovations and improvements for new tenant occupancy. With full-building tenants moving in, the non-structural alteration provisions in the San Francisco Existing Building Code triggered a seismic evaluation of the Tower. While not required by code, the team performed a seismic evaluation of the Banking Hall – building number three on the city’s list of designated landmarks - due to its functional connection and close adjacency to the Tower. The evaluation also achieved the owner’s interest in understanding the seismic performance of the historic Banking Hall.

San Francisco Existing Building Code provisions allowed the use of ASCE 41-17 Seismic Evaluation and Retrofit of Existing Buildings as the basis of the seismic evaluation. The design team worked

Figure 1 400 & 430 California Street

with the Owner and the San Francisco Department of Building Inspection (SFDBI) to establish the ASCE 41-17 seismic performance objective for the two buildings at Life-Safety under the BSE-1E seismic hazard.

400 California Street –Historic Banking Hall

The Banking Hall has plan dimensions of 124 feet by 84 feet and has a single-story roof height of 62 feet and a basement below. At the roof level is a terrace, constructed in 1965, consisting of concrete fill on metal deck over steel framing and is accessed from the 5th floor of the Tower. The 1908-era concrete roof slab, supported by steel plate girders spanning the width of the building, sits below the terrace. Below hangs an ornate plaster ceiling, high above the interior space. The multiple layers of structural framing, interior finishes, and granite cornices create a large seismic mass at the roof level, equivalent to approximately nine typical steel-framed floors. Spectacular granite columns, five feet in diameter, rise along two sides of the building from the sidewalk to the roof cornice. Timber piles support the building at the foundation. Tall windows on three sides of the building limit the lateral forceresisting system to concrete shear walls along the Tower interface, where a series of concrete cells extend from the basement to the roof level. These cells were originally used as ventilation shafts and a longabandoned stair shaft.

430 California Street – High-rise Tower

The Tower is a 21-story building with plan dimensions of approximately 128 feet by 105 feet, a roof height of 315 feet above street level, and three basement levels below grade. The building’s primary lateral system is a dual system of reinforced lightweight concrete shear wall cores and structural steel moment-resisting frames. Upper floors of the Tower – from six and above – cantilever 30 feet over the roof of the Banking Hall, and a portion of the 5th floor is supported directly on the roof of the Banking Hall. Separating the Tower and the Banking Hall was a 3-inch seismic joint.

Seismic Evaluation

An initial linear analysis study determined that the Tower possessed detailing and capacity characteristics that would benefit from a nonlinear response history (NLRH) analysis. Therefore, the next analysis phase for the Tower used the NLRH procedure per ASCE 41-17. The PERFORM 3D analysis model contained eleven pairs of scaled earthquake ground motion records applied to the model. Component

acceptance criteria checks occurred for each ground motion. Fiber and shear elements for the concrete walls and coupling beams, and plastic hinges for the steel moment frames captured the nonlinear behavior.

Given the historical construction of the Banking Hall, its analysis used the linear static procedure per ASCE 41-17. The linear analyses performed in ETABS checked demand capacity ratios against the appropriate m-factors.

A testing lab conducted tests of select materials throughout the Banking Hall and the Tower to determine the material properties of the existing buildings for use in the modeling. The testing found that the existing materials in the Tower were consistent with the original design document requirements. However, testing revealed that portions of the existing concrete walls in the Banking Hall had a compressive strength of approximately 940psi, much lower than that required by code to be considered structural concrete.

The seismic analysis found that the Banking Hall had numerous seismic deficiencies: Inadequate shear wall capacity in both principal directions; an insufficient roof diaphragm; extreme torsional irregularity; and excessive roof drift demands. In contrast, the Tower met the Life-Safety performance objective under the BSE-1E earthquake hazard. However, the 3-inch seismic joint between the Tower and the Banking Hall presented a significant pounding hazard considering the estimated 20-inch building drift for the Banking Hall. The design team’s challenge was to find a retrofit solution for the Banking Hall that would maintain its historic character and, at the same time, address the seismic joint deficiency and pounding hazard on the Tower.

Seismic Retrofit

Initially, the retrofit design focused on the Banking Hall and solutions that would mitigate its seismic deficiencies. However,

Figure 2 Banking Hall Interior

it was impossible to develop a retrofit scheme that would reduce the building’s seismic drift to manageable levels while maintaining the historic character of the building. Therefore, the team had to look beyond the Banking Hall and consider the bigger picture of both buildings working together. The NLRH analysis determined there might be just enough residual capacity in the Tower to help prop up the Banking Hall if they were tied

together. This idea struck a chord and eventually led to the final retrofit approach.

While the two buildings are separate, they share many of the same building systems: HVAC, water supply, fire-water supply, and egress paths. Since they work together on so many other levels, having them help each other seismically made sense. The tied building approach was presented to the owner and SFDBI to obtain a general agreement to proceed with the analysis and design.

The retrofit solution focused on three main elements to address the identified seismic deficiencies: tying the Banking Hall to the Tower, the Banking Hall shear walls, and the Banking Hall roof diaphragm (Figure 3).

Three options were considered for tying the roof of the Banking Hall to the Tower to reduce the relative movement between buildings: solid strut connectors, viscous fluid damper connectors, and Damptech rotational friction damper connectors. The connector ties needed to limit the forces transferred to the Tower to avoid overloading it. Selected because it offered more control over the forces transferred to the Tower, its diaphragm and connections, the rotational friction damper also provided an energy dissipation mechanism that further reduced the Banking Hall seismic demands.

The rotational friction damper is comprised of steel plates sandwiching composite material friction pads clamped together by steel bolts. A total of eight rotational friction dampers with 337-kip capacity and ± 10-inch stroke tie the Banking Hall roof to the Tower. The dampers align with the Banking Hall roof steel plate girders and corresponding concrete shear walls. They hang from the Tower 4th floor framing with tube steel struts that extend across the seismic joint to connect to the Banking Hall shear walls. On the Tower side, new steel beams and struts provide a path to drag the transferred lateral loads into the main elevator and stair concrete shear wall cores.

New concealed concrete walls located in the existing void/cell spaces on the Banking Hall's west side provide the building's primary lateral system. The new walls extend from the roof level to the basement and are highlighted in orange in Figure 4.

At the roof level, a new steel-braced diaphragm was designed to transfer the seismic forces to the new concrete shear walls on the west end of the Banking Hall (Figure 4). Buckling restrained braces were utilized and tuned to provide a ductile diaphragm while limiting deformation in the existing roof slab. The new braced diaphragm framing is located above the 1908 roof slab but concealed below the roof terrace level.

The updated NLRH analysis included the damper ties, modeled simplistically as bar elements with elastic and perfectly plastic behavior. In addition, including the strengthened Banking Hall structure in the model captured the interaction between the two structures and the damper response. The analytical models for the retrofit solution showed that approximately two-thirds of the Banking Hall roof seismic load was transferred to the Tower, and the torsional behavior was improved. Tying the two buildings together altered the global behavior of the Tower structure, with increased story forces and inter-story drifts at the floor levels near the damper connections. However, the strength and deformation capacity of the existing

Figure 3 Seismic Retrofit Elevation

Figure 4 Seismic Retrofit Plan

Tower components continued to meet the Life-Safety performance objective under the BSE-1E earthquake without requiring additional strengthening. Even with the new dampers tying the two buildings together, an estimated ±5 inches of relative movement between the buildings is anticipated under the BSE-1E event. To avoid pounding under this earthquake and to allow the dampers to move freely as designed, the existing 3-inch seismic joint was widened to 9 1/2 inches with partial demolition of an existing lowstrength concrete wall along the building joint.

Retrofit Construction

The seismic retrofit began in June 2019 with substantial completion in December 2020. While the Banking Hall and Tower maintained occupancy, construction commenced with the roof diaphragm bracing at the Banking hall and the collector ties at the 4th-floor framing of the Tower (Figures 5 & 6). Once the buildings were fully vacated, the remainder of the work commenced, including the new shear walls in the Banking Hall cells and the final tie-ins between the Banking Hall and the friction dampers in the Tower. Since this project is the first Damptech rotational fiction damper application in the United States, the design team implemented a substantial testing and validation process. Independent testing labs performed the ASCE 41 required prototype and production tests of the damper for: seismic load testing, temperature variation testing, and loading frequency variation testing. Working closely with the damper manufacturer, a laboratory performed additional tests

to validate the modeling and capacity of the dampers used for the analysis and design.

As with any existing building retrofit project, unforeseen conditions during construction created challenges. The complexity of this retrofit project added an extra level of challenges that had to be overcome. The owner, contractor, and design team worked closely to address construction access and existing condition challenges. Each required unique and innovative solutions, from processes for shotcrete application in the 11-foot by 4-foot by 65-foot tall vertical cells, to the Covid-19 outbreak and shutdown.

The seismic retrofit of the historic 400 California Street Banking Hall is both a tremendous technical achievement as well as an achievement for historic preservation. The Banking Hall has been a fixture on the corner of California and Sansome Streets for 115 years. Enduring many years of smaller earthquakes, it was uncertain if it could survive the next big seismic event. By tying the Banking Hall to the Tower and using its available strength to provide lateral resistance, the two buildings will work together to resist the next big earthquake.■

Project Team

Structural Engineer: Degenkolb Engineers, San Francisco

Owner: 400 California, LLC (a partnership between Takenaka Corporation and Kennedy Wilson International)

Architect: Gensler, San Francisco

Friction Damper Mfr: Damptech, Denmark

General Contractor: Plant Construction Company, L.P. Steel Fabricator & Erector: Olsen Steel

Concrete Contractor: Bradley Concrete

Kirk Johnston, S.E., is a Principal and Group Director in Degenkolb Engineers San Francisco office. kjohnston@degenkolb.com

Yvonne Tsui, S.E., is an Associate with Degenkolb Engineers in San Francisco, California. ytsui@degenkolb.com

Figure 5 Friction Dampers Installed

Figure 6 Roof Diaphragm Bracing Construction

Climate Change and the Structural Engineer

How it may affect the Profession. Part 2

By Gail Kelley

Part one of this series, published in the May 2023 edition of STRUCTURE, discussed climate change in general. Part two, presented here will look at how climate change could impact an engineer’s standard of care. Most engineers are familiar with the “standard of care” – the standard that their services will be held to; failure to comply with the standard of care is negligence. As the effects of climate change intensify and our understanding of these effects evolves, it seems inevitable that there will be impacts on the standard of care.

While the standard of care is expressed slightly differently in the different design and construction standard form documents, the general concept is the same. The AIA B101, Standard Form of Agreement between Owner and Architect, provides a good example, and engineers often use similar wording when drafting their agreements:

§ 2.2 The Architect shall perform its services consistent with the professional skill and care ordinarily provided by architects practicing in the same or similar locality under the same or similar circumstances.

Designing for Location

Common to the formulations for the standard of care in the various industry standard form documents (AIA, EJCDC, ConsensusDocs, DBIA) is the concept that the standard of care is tied to location and circumstances. The importance of location to the standard of care is not new. For example, seismic concerns are central to many parts of California; designing for expansive clays is required in many parts of Texas. However, as the effects of climate change continue and become more frequent however, the engineer’s standard of care could include consideration of the potential future conditions at the project location. While the wording of the standard of care may not change, the environmental factors that must be considered will likely become more important.

Foreseeability and the Standard of Care

The basic question is whether the engineer has a duty to consider the future risks that could arise from climate change and adapt its design accordingly. Professional negligence (negligence in the provision of

professional services) is simply a type of “ordinary negligence.” Under ordinary negligence principles, if it is foreseeable that an individual’s action or lack of action could cause damage, the individual has a duty to prevent the damage. An engineer could be found negligent and liable to third parties as well as its client if its design fails to consider and protect against foreseeable damage. The challenge with designing for potential future conditions is that foreseeability can be subjective. Factors that could affect foreseeability with respect to climate change include news reports; vendor literature; scientific, government, and industry reports and guidelines; availability of technology and design strategies; what other engineers are doing, and code requirements in other jurisdictions.

Example of the Applicability of Guidelines

Engineers working on private projects should be aware of laws and guidelines governing public development. Future plaintiffs could argue that requirements and guidelines for public projects are admissible to attack or to buttress expert opinions on the prevailing standard of care for private development. An example would be New York City’s Climate Resiliency Design Guidelines, which identify climate change

risks and appropriate resiliency interventions for City projects— such as raising equipment when building in a potential floodplain. Various other states and cities have developed or are developing similar guidelines.

Warranties with Respect to Climate Change

Nevertheless, engineers should be careful about warranting that the project will comply with climate adaptation guidelines. In most instances, compliance with such guidelines will be a function of the owner’s willingness to pay for increased studies and investigations (and likely increased project costs), as well as the Contractor’s implementation of the design - none of which the engineer has control over.

Engineers should also be careful about agreeing to obligations that they cannot fulfill without cutting-edge technology. New technologies that purport to provide great benefits can carry more risk as their long-term performance may not be well understood. When an engineer is required to employ a novel strategy to address an issue because there is no proven solution, the engineer’s contract should explicitly acknowledge that the solution is untried and the engineer cannot guarantee the outcome.

Communication with the Owner

Implicit in the standard of care is the concept that engineers do not warrant that their services will be either perfect or defect free; they only warrant that the services will comply with the standard of care. The standard of care requires compliance with applicable codes and standards. However, current codes and standards incorporate historical weather data to determine how to design for today’s conditions; they may not accurately address projected future storms, sea level rise, or heat waves. While engineers may eventually get assistance in the form of updated code requirements that explicitly address climate change, it may be incumbent upon engineers to be proactive about considering how to incorporate climate resilience into their designs.

Depending on the project, the engineer and owner may need to discuss the due diligence that is required to determine the appropriate design standards for climate resilience. Appropriate due diligence might include conducting research on publicly available weather data and projections, as well as climate impact maps and models, to determine if the building should be designed for known or highly likely, future climate-related risks.

The engineer might need to consult with appropriate professionals to determine the accuracy of climate data, including flood or storm surge maps, and regional climate vulnerability studies. While

engineers are increasingly engaging climate experts in the early stages of the design process, it should be emphasized that the engineer is responsible for the work of its consultants. It may be advisable for the Owner to take responsibility for these investigations and determinations. Part three of this Series will look at contractual provisions that can help protect an engineer from potential claims that their failure to adapt their designs for climate change is negligence.■

Gail S. Kelley is a professional engineer as well as a LEED AP and a licensed attorney in Massachusetts, Maryland, and DC. Ms. Kelley's work is focused on representing design professionals in both contract negotiations and claims management.

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Designing for the Unforeseeable Simplicity in the face of uncertainty.

By M. McElmury, D. Nickell, R. Reckers, S. Pujol

To design a building to survive an unspecified threat, as it is done in design against progressive collapse, is outside the realm of conventional structural engineering. Without knowing anything about the event, other than it may occur, the objective of the design must be to produce as redundant a structure as possible and as practical within given economic restraints. This exercise amounts to asking the engineer to design for the unknown. There is no time or money to design a building to withstand everything. Given that level of uncertainty, the old engineering adage that computational effort must be proportional to the quality of results rules out the use of detailed, time-consuming calculations.

In agreement with these ideas, the apparent goal of current Codes addressing progressive collapse and blast resistance is not to ensure complete building survival in the mentioned unknown event but to provide a level of structural redundancy deemed sufficient. Nevertheless, the specific analyses and procedures required by the Codes to meet this goal often involve extensive computational time and effort.

To look at the problem from a different perspective, consider this question: If a designer is tasked with assessing an existing building for progressive collapse (assuming, for instance, that a column may be lost) or tasked during conceptual design to proportion a structure to withstand a given blast, but is told there is no time available for a full structural analysis, how should they proceed? Spending engineering power on the details of the dynamics of a singular immaculate column removal that will never occur as assumed is a waste of engineering power.

This article shows that the stability of a flooring system after the abrupt removal of one support:

1. can be judged using the simplest ideas in statics and the conservation of energy,

2. can be addressed with a carefully chosen load or safety factor, and

3. can be evaluated without the use of software.

This simpler approach, based on principles of dynamics formulated by J.M. Biggs and the fundamental concepts of Virtual Work and Conservation of Energy, provides quick and reasonable results. These results can help the initial proportioning of structural members that can then be checked by more elaborate means if required by regulations.

By way of a design example, we show that this simpler approach produces results comparable to results produced by detailed analyses that require more time. More demonstrations are given by Smith (2009).

Design time can be spent more productively on critical detailing, which is often the main factor determining structural redundancy and tenacity. Detailing is the key to reduce susceptibility to progressive or disproportionate collapse. More time spent on detailing will likely lead to a better structure than computations about a notional event.

The Basis of The Proposed Procedure

In 1964, Biggs introduced the idea that the main aspects of the dynamic response of a continuous system can be understood using a single degree of freedom system chosen so that the kinetic and potential energies in both systems are equal to each other. He arrived at this equivalent system by

assigning to it:

• a fraction (0.5 to 0.75) of the mass of the continuous system,

• strength equal to the total strength of the continuous system for static conditions, and

• an applied force equal to the total force acting on the continuous system (which can vary with time).

The equivalent system, so tuned, produces displacements equal to the maximum (reference) displacement in the continuous system.

It follows that, ignoring strain-rate effects, the response of a floor system to the abrupt removal of a support can be understood in terms of the static strength or resistance of the system missing the removed support (�� y) and the total weight acting on it ��. Ignoring the effects of damping, the work done by that weight through the peak floor displacement ∆ is ��×∆. That work must be equal to the area under the resistance-displacement curve representing the static response of the system (if strain-rate effects on strength are ignored): ��× ∆ =�� y×(∆-∆y/2)

Here ∆ y is yield displacement. From this, it follows

where ��D�� = �� y /�� is the ratio of static capacity and demand (in terms of total forces acting on a flooring system in the area affected by the support removal), and µ is ductility demand (the ratio of peak floor displacement to yield displacement associated with resistance �� y). For the unlikely case that CDR>2, estimated ductility is smaller than 1. In that case, the equation of conservation of energy becomes R 2 1 2 W y

y 2 ## D D D = or

CDR CDR 2 if 2 () 2 n =

Figure 1 shows a plot of ductility demand v. capacity demand ratio CDR. Figure 1 suggests that a capacity-demand ratio (or safety factor) of 1.5 results in a small ductility demand of 1.5. A defensible capacitydemand ratio of 1.2 yields a modest ductility demand of 3. Strain-rate effects and strain hardening are bound to increase resistance, and the detailing required to achieve the mentioned ductility values (1.5 to 3) is unlikely to be taxing. This simplified method produces results comparable with those obtained from detailed and time-consuming nonlinear dynamic analyses (Smith, 2009). The engineering solution is simple: almost any acceptable safety factor suffices. The critical decisions affecting the survival of a structure subject to an attack that can render one of its columns useless should not depend on the type of material used in construction. Conservation of energy applies to steel, timber, concrete, masonry, and FRP plastics without discrimination. The critical decisions have to do with the following:

• whether to consider the removal of more than one column at a time

(as observed in Oklahoma City, The Pentagon, and WTC), and •whether to consider damage to the flooring system itself. Given these rather obvious sources of uncertainties, spending time analyzing instead of detailing seems futile. Instead, simple static analysis and a reasonable factor of safety (exceeding 1.2 to 1.5) should suffice and yield time for careful consideration of details (welds, splices, anchorage, confinement).

The Current Building Code

The U.S. Department of Defense’s UFC 4-023-03, Design of Buildings to Resist Progressive Collapse, offers three approaches for design: ALP – Alternative Load Paths, ELR – Enhanced Local Resistance, and the Tie Method. ELR and the Tie Method mostly relate to critical detailing. Therefore the focus becomes ALP and, in particular, the use of equivalent static analysis to determine initial member proportions.

In the UFC approach, as in conventional force-based design, capacity is required to exceed demand. Capacity refers to estimates for static conditions modified with factors, some of which increase and decrease it:

•an m factor (>1) related to the ductility of the structure,

• a strength-reduction factor (<1) – as used in conventional design for static forces – that presumably has to do with uncertainties in strength estimation and material properties as well as consequences of failure, and

• an overstrength factor (>1), reportedly related to the idea that actual material strengths tend to be larger than nominal strengths.

No explicit connection is made to the

plausible effects of loading rates. The idea to increase the capacity in apparently direct proportion to ductility capacity (through factor m) is not explained in better terms other than by saying that it is done to imitate the methods in ASCE-41, Seismic Rehabilitation of Existing Buildings (which has to do with earthquakes instead of impulsive loads).

Demand is modified in two ways (in the same direction this time):

•By using load factors (>1 at least for dead loads) to admit that loads may be larger than expected

•By using a load increase factor (or LIF) that, on average, and according to results from numerical simulations done to select the values recommended in the guidelines (McKay et al., 2012), approaches LIF=m+1 (depending, reportedly, on the properties of the materials in the structure). Note that the guidelines recommend more conservative values.

The main apparent justification for using the LIF seems to be that it was chosen to match dynamic analyses done in SAP2000 (McKay et al., 2012).

Lumping all strength-reduction, strength-increase, and load factors into a single ratio of capacity to demand, the current approach in UFC translates into:

Increased Capacity ≥ Increased Factored Demand: m CDR ≥1+m or, at the limit, m CDR

If one assumes m to be a measure of ductility demand µ, then this expression can be compared with the expression obtained before from conservation of energy (Figure 2):

The two curves in Figure 2 are remarkably similar. But why should they not be so? After all, well-written code for numerical simulation ought to produce results consistent with basic physics. Rather, the question is: why can’t the engineer rely on simple physics instead of numerical results?

Both curves tend to infinity at CDR=1. That stands to reason. Both

Figure 1 Ductility Demand v. Capacity to Demand Ratio (for static conditions)

Figure 2 Ductility Demand v. Capacity to Demand Ratio Comparison

curves produce a ductility demand of 1.0 for a capacity-demand ratio of 2. That is consistent with the proverbial example in which a linear spring supporting a mass deforms twice as much if the mass is let go suddenly instead of slowly. And both curves suggest again that the problem of the dynamics of a ductile floor responding to abrupt support removal can be reduced to choosing a safety factor and decent detailing.

Design Example

The following procedure illustrates how this approach can be used in designing a structure subject to the UFC progressive collapse requirements. It includes four steps:

1.Design for gravity + wind/lateral loads

2.Perform analyses to estimate static capacity (yield-line analysis should suffice)

a.If calculated CDR > 1.2, continue to Step 4.

b.If calculated CDR < 1.2, follow Step 3

3.Add reinforcing needed to reach the desired ductility

a.Revise structural system to provide higher capacity

b.Increase member sizes and/or revise detailing

4.Perform Code required analyses (as a check)

The following example illustrates that simple physics and consideration of detailing can be used to identify quickly where structures lack redundancy. Nevertheless, the example shows that a detailed analysis would still provide similar results.

Example 1 – UFC Model

Consider the five-story reinforced concrete building described in UFC (2005) and illustrated in Figures 3a and 3b. Column B3 (marked in dark shading) is assumed to be removed in the ground story.

The reinforcement schedule is given in Table 1. Effective depths are taken as total depths minus 2.5 inches.

Static Demand

In this example, it shall be assumed that girders resist the gravity floor loads as uniformly distributed forces corresponding to a uniform tributary width of 25 feet.

Specified loads are 50 psf live load, 90 psf dead load, and selfweight. The simplified analysis presented here focuses on the 25-foot- wide floor area shaded in Figure 3 on both sides of girder B, between axes 2 and 4. For this area, uniformly distributed values of live and dead load are 1.25 kip/ft and 2.25 kip/ft. Girder self-weight is 0.73 kip/ft. The total factored load is w u =4.2 kip/ft using load factors of 0.5 for live and 1.2 for dead loads.

Static Capacity

Table 1 lists flexural strengths estimated for a) internal arm lengths approximated as 0.9x effective depth, and b) reinforcement yield stress fy =60ksi. In the absence of column B3, static resistance is estimated using the principle of virtual work for a flexural mechanism with plastic hinges in beams and girders at nodes A3, B2, B3, B4, and C3. Virtual rotations are taken as 1/25ft at A3, B2, B4, and C3 and 2/25ft at B3. Total internal work associated with a unit vertical displacement at B3 is, therefore:

Figure 3a Building Plan.

Figure 3b Building Elevation.

1 Cross-Sectional Properties.

External work is taken simply as the total force acting on girder B2-B4 times average virtual displacement (½) consistent with the assumed mechanism and loading:

w 50 2 1 EW ft y # =

Equating EW and IW leads to w y =11 kip/ft=2.6 w u. The ratio of capacity to demand CDR=w y /w u is 2.6>2, indicating the floor system would remain linear. A linear analysis of the structure suggests the deflection associated with the yield force is 1.4 inches, indicating a vertical stiffness of K=7.8 kip/ft/in.

Dynamic Response

The maximum deflection for a sudden application of the load would be twice the static deflection: 2 w u /K =1.1in. A dynamic analysis in SAP2000, including slow application of the loads followed by the sudden removal of column B3, produces a peak displacement of 1.3in (smaller than the yield deflection).

Variations in Reinforcement

A reduction in the longitudinal reinforcement resisting tension in beams and girders to a uniform 4 in2 results in an estimated moment strength of 5900 kip in, instead of the values in Table 1. The associated yield load becomes w y =6.3 kip/ft=1.5 w u. For a ratio of yield capacity to static demand CDR=1.5, the expected dynamic ductility demand becomes

CDR 21 1.5 # n = = -

the design engineer performing analysis for progressive collapse is the importance of continuity of reinforcement. Independent of Code-required analysis or the simpler approach presented above, the designer can go a long way towards producing a structure that can perform better under an unspecified threat:

• By considering the following for reinforced concrete structures:

ˏMaximize the continuity of reinforcement

ˏProvide continuous reinforcement and check lap splices

ˏCheck the shear and punching shear of the frame and slab elements

ˏCheck shop drawings against the design intent of structural drawings.

•Or consider the following for structural steel frames:

ˏ Consider using prequalified Special Moment Frame (SMF) connections.

ˏ Optimize column spacing to maximize floor framing efficiency.

ˏProvide similar or same beam depths framing into columns w/stiffeners for ease of connection detailing.

ˏOrient floor framing to provide adequate LTB bracing of primary Alternate Path structural elements.

ˏCheck shop drawings against the design intent of structural drawings.

Conclusion

CDR

^h

The associated peak displacement is 2.1 in. A nonlinear dynamic analysis in SAP2000 produces a peak deflection of 2 in. The deflected shape and the history of peak displacement (at B3) obtained in that analysis are illustrated in Figures 4a and 4b.

In this example, the results from a simple analysis are compatible with results obtained with elaborate analysis. A close examination of the expression

µ=CDR/(2×(CDR-1)), as illustrated in Figure 2, will show that for values of CDR<1.25, small changes in CDR result in large changes in ductility demand. The sensitivity of this relationship may produce deviations among results obtained from different forms of analysis.

Additional Considerations

In line with detailing requirements in the ELR and Tie Method in UFC and outside the comparative example outlined above, a key takeaway for

To design a floor to survive abrupt column removal choose a ratio of static nominal capacity to static demand exceeding 1.2 to 1.5, and spend more time on detailing instead of the numerical analysis of the dynamics of an event that may never occur .

Full references are included in the online version of the article at STRUCTUREmag.org.

M. McElmury, S.E. is an Associate at TGRWA in Chicago, IL. (mmcelmury@tgrwa.com)

D. Nickell, S.E. is a Principal at TGRWA in Chicago, IL. (dnickell@tgrwa.com)

S. Pujol is a Professor of Civil Engineering at the University of Cantebury in New Zealand. (santiago.pujol@canterbury.ac.nz)

R. Reckers, S.E. is a Principal at TGRWA in Chicago, IL. (rreckers@tgrwa.com)

Figures 4a Deflected Shape in SAP 2000

Figures 4b Deflection over time in SAP 2000

St. Mary’s in the Mountains Cathedral, built in 1876, in Virginia City, Nevada provided spiritual guidance to the rough and tumble miners of the Comstock Lode. St. Mary’s, known as the “Jewel of the Comstock”, is 170 feet to the top of the steeple and is the most prominent feature in the city. For almost a century and a half, the church has celebrated mass and is one of Virginia City’s most visited tourist attractions.

Rock and Roll is a Problem

Virginia City is located in a high seismic zone, and has the potential for large earthquakes. In a previous retrofit effort in 1970, a 3½" layer of reinforced concrete gunite was sprayed on the face of the interior walls. However, this concept did nothing to establish a competent lateral load system, nor provide

SAVING A HISTORIC VIRGINIA CITY ICON

The seismic restoration/ rehabilitation of the historic St. Mary’s in the Mountains Catholic Church, Virginia City, Nevada

By Paul A. Ferrari, P.E.

proper strengthening of the wall. In the late 1990s, a structural engineering review of the church revealed it was subject to significant distress and failure from seismic activity, which threatened the life/safety occupancy of the building. The Catholic Diocese of Reno, the owners of St. Mary’s, found itself with a problem. Could the building be seismically retrofitted to provide life/safety while also preserving the architectural integrity of the historic building? St. Mary’s has a very prominent role in Virginia City’s early history, while continuing to be a visible and pronounced icon of both the city’s religious and secular history. Therefore, the Diocese decided it made sense to do a structural seismic retrofit with the caveat that the historic appearance of the building not be altered. The decision was warmly received by both the historic community and Virginia City’s general population. The Diocese then turned its attention to obtaining an engineer with the appropriate knowledge and skill in restoring historic buildings. The chosen

St. Mary’s west elevation

project engineer, Paul Ferrari, P.E., has performed seismic retrofits for many of the important historic structures in Virginia City, including Piper’s Opera House, the Fourth Ward School, and the Virginia and Truckee Railroad passenger depot.

Now the hard part. One of the biggest challenges of the St. Mary’s project that Ferrari faced was how to add seismic competence to a brick building that is 146 years old without changing the historic interior and exterior look of the building. The main seismic structural concern in the building is the sanctuary’s thirty-foot high brick walls. The walls are composed of five wythes (layers) of partially fired bricks bedded in a “mortar” of mine tailings sand and slaked lime, which, without Portland cement, has no structural integrity. Wood in the Comstock to fully fire the brick kiln was expensive; as a result, the bricks are only partially fired. This creates a brick with a hard outside and a soft clay inside – think of a stale marshmallow.

Earthquake engineering was unknown when the building was built in the 1870s. Typically, brick walls in Virginia City were constructed with a mass capable of resisting the high Zephyr winds of the Comstock. This was a lesson well learned, since the original wood church, built in 1860, was demolished in a wind storm. This resulted in brick bearing walls that have high mass with no strength, and rigidity with no ductility; not a good situation in an earthquake. Once the walls reach a certain seismic force level, they will rupture suddenly and catastrophically. The earthquake challenge is to strengthen the walls for both in-plane seismic forces and out-of-plane “pull away” seismic forces, while providing internal structural continuity between the wythes of the brick wall.

Thinking Out of the Box and Into the Walls

How can seismic competence be added to St. Mary’s brick walls that essentially have no strength without changing the building’s historic interior and exterior look? Engineering ingenuity was required. It was obvious that structural strength had to be integrated directly into the existing brick masonry walls. This would provide a structure capable of resisting both in-plane and out-of-plane seismic forces while also structurally strengthening the wall’s actual construction. Most importantly, the integration of structure directly into the wall would not change the historic appearance of either the interior or exterior of the building. To solve this problem, Ferrari employed a system he called “emplacement” to provide seismic strength, while retaining the essential historic character of the building. Emplacement involves integrating reinforced concrete columns and beams into the thickness of the brick bearing

walls. Temporary structural shoring was installed to relieve the bearing walls of their roof and floor loading and allow the emplacement of the column strips. Since the north and south bearing walls have regular window openings along the length of the sanctuary, the most obvious location is to emplace the column strips into the three-and-a-half foot wide wall area between the windows, which extend from the roof to the basement at the north and south walls (Figure 1). The emplaced column strip is then anchored to diaphragms at the roof and floor level (Figure 2). To emplace the concrete columns, strips were saw cut into the interior face of the walls and two of the interior brick layers were removed (leaving the exterior brick layers). This creates a column strip void extending from the roof to the foundation at the inside face of the wall (Figure 3). Reinforced concrete columns, designed for in-plane shear, and out-of-plane bending are then emplaced into the void strips in the walls (Figures 4, 5). The result was a three-foot wide x twelve inches deep reinforced concrete column at a spacing of six feet on-center. The column is designed for in-plane seismic shear as a shear wall; and for out-ofplane seismic forces as a beam spanning from the floor to the roof. The net result is that the emplaced column did not alter the exterior elevation, while the interior elevation could be plastered over, identical to the original, so that no change would be visible inside the church (Figure 9).

Special care had to be taken during the concrete placement in the

Figure 1 Graphic isometric of seismic forces on St. Mary’s and proposed retrofit concept.

Figure 2 Cross-section of the exterior wall showing the load path for perpendicular to wall seismic forces.

Figure 3 Column strip void in wall showing removal of brick at interior face of wall.

new column voids. Thirty feet of wet concrete hydrostatic pressure could “blow out” the exterior face brick of the wall. To avoid this, a plywood and stud panel form was constructed; one of the panels was used as the interior form of the column, the other panel was installed opposite, on the outside of the brick wall. The two forms were joined by threaded rods through the wall secured into each panel to create a “sandwich” which countered the hydrostatic concrete pressure and saved the wall (Figure 4). Also, each of the twelve column voids was only filled to one-third of the height in a roundrobin sequence, so that when the next third was placed, the previous third had already set, reducing hydrostatic pressure further (Figure 6).

Whiskers

Installing the emplaced concrete columns provides the walls with overall structural seismic capability. However, the five wythes of brick in the wall needed to be better integrated between the wythes; there are few perpendicular header bricks tying the wythes together. It is possible that, during an earthquake, the exterior face brick could peel away from the wall, causing a safety problem for pedestrians below, and ruining the historic appearance of the church. To help prevent this, Simpson stainless steel quarter-inch

diameter helical anchors “whiskers” were helically screwed into the brick layers from the new columns voids every thirty inches vertically. The five anchors at each level were installed in a “splayed” configuration. The anchors were screwed through the brick layers and penetrated into the exterior wall’s back face brick (Figure 4). The butt end of the steel whiskers is captured by the structural concrete in the new column, anchoring the whiskers, and therefore, the brick in place to prevent brick spalling. The splayed arrangement of the whiskers from adjacent columns strips also helped to secure the brick areas above and below the windows, stabilizing the face brick in the unreinforced wall areas.

Lateral Load Structural Concept

Although the new concrete column emplacement in the walls provides a structural system to resist seismic loadings on the walls, it is only part of the needed total seismic lateral load system. The in-plane and out-of-plane seismic loads in the emplaced columns had to be integrated into a coherent lateral load concept for the overall building. To accomplish this, the top of the new emplaced column were anchored into a new horizontal sixteen-inch high x twelve-inch wide concrete “ring beam”. The ring beam extends around the entire building along the top of the existing walls in the void between the top of the wall and the bottom of the roof. The ring beam, in turn, transfers both out-of-plane and in-plane seismic forces from the emplaced columns to a new plywood roof diaphragm that was glued and nailed into the existing plank roof (Figure 7). The diaphragm force is then transferred to the parallel east and west walls that have been reinforced and tied into the ring beam and roof diaphragm. The concrete ring beam is hidden from interior view by the original ornate wood crown molding that was salvaged and reused. The emplaced columns are also tied to a new horizontal plywood diaphragm at the bottom of the sanctuary floor joists. Placing the new plywood diaphragm at the bottom of the

Figure 5 Emplaced column reinforcing in place in wall void strip.

Figure 4 Cross-section of the wall showing newly emplaced reinforced concrete column, with stainless steel helical whickers and wood shoring at interior and exterior wall faces to prevent hydrostatic blowout.

Figure 6 Reinforcing and partial concrete placement in emplaced column.

floor joists (ceiling of the basement) preserves the historic floor above, while transferring out-of-plane seismic forces to ground via new basement shearwalls (Figures 2, 8).

Conclusions

The emplacement concept was a success for St. Mary’s. It met all of the original structural design constraints: a competent structural seismic system was integrated into the building, and the historic interior and exterior appearance was unchanged (Figure 9). In 2009, the cost of the structural portion of the retrofit was $1,500,000.

The emplacement concept is driven by a specific set of constraints that may be unique to St. Mary’s in the Mountains. The Church is located in a high seismic zone. The owner required that the historic appearance of the building be maintained. And most importantly, the bearing walls were constructed with “mortar” that had no Portland cement and no structural value; essentially, the walls have no modulus of elasticity. This last consideration was the most difficult and generated the need for the emplacement concept. This is probably not a widespread condition in most historic buildings. St. Mary’s wall construction is essentially a product of Virginia City’s mining boom where the builders made do with locally available materials. Unless similar conditions exist in a historic structure retrofit, particularly in the wall strength, it is unlikely that the emplacement method would be appropriate. However, if these conditions are present, emplacement may be the only concept that addresses all of the building’s constraints.■

Project Team

Engineer of Record: Paul A. Ferrari, P.E.

Owner’s Representative: Mike Quilici

Project Manager: Robert Wright

Contractor: Reyman Brothers Construction

Paul A. Ferrari, P.E., was the founder and president of a structural engineering consulting firm in Reno, Nevada for over 37 years, and is continuing in practice. He has structurally stabilized/retrofitted over 30 historic structures, and is currently chairman of the University of Nevada Reno Civil and Environmental Engineering Advisory Board.

Figure 7 Cross-section of the wall at roof ring beam/roof diaphragm to top of emplaced column.

Figure 9 Completion of the emplaced column at the interior of the wall. Emplaced column construction’s final configuration is identical to the original historic interior of the building, achieving the primary goal of the seismic retrofit.

Figure 8 Cross-section of wall at emplaced column transfer out of plane seismic force to new horizontal floor diaphragm.

2023 Diversity in Structural Engineering Scholarship Award Winners

In 2021, the NCSEA Foundation established the NCSEA Diversity in Structural Engineering Scholarship Program to award funds to students traditionally underrepresented in structural engineering (including but not limited to Black/African Americans, Native/Indigenous Americans, Hispanics/Latinos or Spanish, Asian, Native Hawaiian or Pacific Islander, and other people of color, those with disabilities, veterans and LGBTQIA+).

For this year's awards, three structural engineering firms, one structural engineering family, and two structural engineers' associations have joined the NCSEA Foundation to create opportunities for aspiring structural engineers in the form of endowed and named scholarships. More than 100 scholarship applications were received this year from across the country. After a thorough review of all applicants, the NCSEA Foundation awarded 10 scholarships worth $34,500.

More information about the awards can be found on www.ncsea.com/about/foundation/diversityscholarship/

Daniel Saulter NCSEA Diversity in Structural Engineering Scholarship Texas A&M University

Emily Dahbura NCSEA Diversity in Structural Engineering Scholarship Oregon State University

Valeria Martinez

SEAONC Diversity in Structural Engineering Scholarship University of the Pacific

Isabella Bernard NCSEA Diversity in Structural Engineering Scholarship The Catholic University of America

Wayne Carter

The Martin/Martin Skyrise Scholarship University of Missouri

Erick Hernandez-Centeno SEA-MW Diversity in Structural Engineering Scholarship Morgan State University

Haben Legesse NCSEA Diversity in Structural Engineering Scholarship *Graduating from Syracuse University

Ivan Alvarez

The Steven B. Tipping Memorial Scholarship for Innovation and Excellence in Structural Engineering The University of California, Berkeley

Jerry Tran The Degenkolb Engineers Diversity in Structural Engineering Scholarship University of California Irvine

Kelly Espino The Applied Technology Council Diversity in Structural Engineering Scholarship *Graduating from UCLA

News from the National Council of Structural Engineers Associations

Leadership Week: Energy, Excitement Propel NCSEA, SEAs

The second annual NCSEA Leadership Week was held June 5-8, 2023, to provide a valuable opportunity for collaboration and information exchange between the NCSEA Board of Directors, NCSEA Committee Chairs and SEA Leadership. More than 90 people gathered in the Chicago area representing 38 Structural Engineers Associations (SEAs) for four days of education, networking, leadership training and collaboration focused on propelling the SEAs and NCSEA to the future. Highlights of the week included:

• Strategic Planning, IT Innovation and Experience Creation for SEAs

o Several thought leaders, including Dean West, Association Laboratory, Kevin Ordonez, .orgSource, and NCSEA's Kat Ort Ely and Leo Baran, provided expert insights on strategic planning, functional websites, and creating impactful educational and networking experiences, both virtual and in-person.

• We SEE Above & Beyond Branding Campaign Update

o Elizabeth McKenna from AGMK reminded all attendees that the work of structural engineers IS truly remarkable while sharing new components of the We SEE Above & Beyond campaign. Visit www.weseeaboveandbeyond. com to explore the campaign.

• Collaborative Breakout Sessions

o Breakout sessions on both days were an incredible opportunity to connect with fellow SEA leaders and learn about NCSEA Committees. Conversations focused on common struggles, unique opportunities, and collaborative ways to find solutions.

NCSEA Webinars

• July 18 Structural Building Condition Reviews: Beyond Distress

• August 1 Optimizing Steel Design: A Guide for Collaboration between Engineers and Steel Fabricators/Erectors

• August 10 New Bridges or Transportation Structures: I-74 Bridge over the Mississippi River

Purchase an NCSEA webinar subscription and get access to all the educational content you'll ever need! Subscribers receive access to a full year's worth of live NCSEA education webinars (25+) and a recorded library of past webinars (170+) – all developed by leading experts; available whenever, wherever you need them!

Courses award 1.0 -1.5 hours of Diamond Review-approved continuing education after completing a quiz.

Recommendations for Performing Structural Engineering Quality Assurance Reviews

New addition to the ASCE/SEI 7-22 standard protects buildings from a 500-year flood event

ASCE 7-22 Supplement #2 Flood Chapter 5 now available as free download

Call for Public Proposals for ASCE 7-28

SEI is accepting proposals to modify the 2022 edition of ASCE/SEI 7 Minimum Design Loads and Associated Criteria for Buildings and Other Structures, as the committee prepares for the 2028 revision cycle.

Call for Members - Join an SEI Committee

The Seismic Effects for Non-Buildings Structures and Foundations Committee is seeking new members. The committee fosters communication between researchers and practitioners and improve understanding of the behavior, analysis, design, construction, inspection and retrofit of non-buildings structures and foundations subject to earthquakes and/or other dynamic loads. Apply at www.asce.org/SEICommittees

Volunteers Needed for NCEES PE Structural Exam Cut Score Panel

NCEES is assembling panels of licensed structural engineers and licensed professional engineers to participate in a series of two-day meetings (Sept 18-19 and 20-21) to determine the cut score, or establish the pass point for the 2024 Principles and Practice of Structural Engineering exam. This process requires a cross-section of engineers from various employment positions and technical specialties. Learn more at https://ncees.org/volunteer

SEI Members: Vote in online election for the SEI Board ofGovernors

Current SEI members (dues fully paid) above the grade of Student will receive a notice via ASCE Collaborate on how to verify and submit your secure ballot online in July. Ballots are due no later than 11:59pm US ET, July 31.

Learn and Grow with ASCE Continuing Education

Some highlights July – August:

Seismic Analysis of Structures and Equipment (8016CEN2023)

Introduction to ASCE 7-22 Requirement for Tornado (8053IW2023)

Earthquake Engineering for Structures (GOCEEFS23)

Seismic Evaluation and Retrofit of Existing Buildings (GOCSERB23) go.asce.org/StructuralContinuingEducation

Leverage your SEI Futures Fund

www.asce.org/SEIFuturesFund

Errata

SEI Standards Supplements and Errata including ASCE 7.

See www.asce.org/SEI. To submit errata, contact sei@asce.org.

donation with the CSI 3 to 1 match, and help realize our Vision for the Future of Structural Engineering

Congratulations to SEI and ASCE Structural Award Recipients

SEI and ASCE honor and thank those who have made significant contributions to the structural engineering profession through publishing outstanding papers, dedication to technical excellence, and by providing exemplary service and leadership:

SEI President’s Award

Dan Linzell, Ph.D., P.E., F.SEI, F.ASCE

Walter P. Moore, Jr. Award

J. Greg Soules, Ph.D., P.E., P.Eng, S.E., F.SEI, F.ASCE

W. Gene Corley Award

John Grieshaber, P.E., S.E., F.SEI, F.ASCE

SEI Chapter of the Year Award SEI Pittsburgh Chapter

SEI Graduate Student Chapter of the Year Award

SEI Graduate Student Chapter at the University of Oklahoma

2022 Awards also recognized at Structures Congress in New Orleans

Alfredo Ang Award on Risk Analysis and Management of Civil Infrastructure

Michael Beer, Dr.-Ing., M.ASCE

George Winter Award

Edmond Saliklis, Ph.D., P.E., M.ASCE

George Winter Award

Sankaran Mahadevan, Ph.D., F.EMI, M.ASCE

Moisseiff Award

Jurgen Becque, M.Eng., M.Sc., Ph.D.

Raymond C. Reese Research Prize

Ran Cao, Ph.D. A.M.ASCE; Anil Agrawal, M.ASCE; Sherif El-Tawil,Ph.D., P.E., F.SEI, F.ASCE; and Waider Wong

Shortridge Hardesty Award

Robert G. Driver, Ph.D., P.Eng., F.SEI, M.ASCE

Moisseiff Award

Liang Liu, Ph.D., Aff.M.ASCE; David Y. Yang, Ph.D., A.M.ASCE; and Dan Frangopol, Sc.D., P.E., F.SEI, F.EMI, Dist.M.ASCE

Raymond C. Reese Research Prize

Andronikos Skiadopoulos, Ph.D., Aff.M.ASCE; Ahmed Elkady; Dimitrios G. Lignos, Ph.D., M.ASCE

Shortridge Hardesty Award

Sherif El-Tawil, Ph.D., P.E., F.SEI, F.ASCE

Congrats to Recipients of the 2023 O.H. Ammann Research Fellowship for creation of new knowledge in structural design and construction

• Srinivasan Arunachalam, Aff.M.ASCE, University of Michigan, Ann Arbor

• Angshuman Baruah, S.M.ASCE, University of Illinois Urbana-Champaign

• Edvard Bruun, P.Eng, S.M.ASCE, Princeton University

• Konstantinos Kalfas, Ph.D., S.M.ASCE, Southern Methodist University

• Hannah Power, S.M.ASCE, University of Delaware

CASE in Point

Tools To Help Your Business Grow...

If you are a member of CASE this tool and all publications are free to you. NCSEA and SEI members receive a discount on publications. Use discount code - NCSEASEI2022 when you check out.

Check out some of the CASE Contract Documents developed by the Contracts CommitteeÉ

• CASE #1 – An Agreement for the Provision of Limited Professional Services. This is a sample agreement for small projects or investigations of limited scope and time duration. It contains the essentials of a good agreement including scope of services, fee arrangement and terms and conditions.

• CASE #2 – An Agreement Between Client and Structural Engineer of Record for Prof. Svs. This agreement form may be used when the client, e.g. owner, contractor developer, etc., wishes to retain the Structural Engineer of Record directly. The contract contains an easy to understand matrix of services that will simplify the “what’s included and what’s not” questions in negotiations with a prospective client. This agreement may also be used with a client who is an architect when the architect-owner agreement is not an AIA agreement.

• CASE #9 – An Agreement Between Structural Engineer of Record and Consulting Design Professional for Service. The Structural Engineer of Record, when serving in the role of Prime Design Professional or as a Consultant, may find it necessary to retain the services of a sub-consultant or architect. This agreement provides a form that outlines the services and requirements in a matrix so that the services of the sub-consultant may be readily defined and understood.

You can purchase these and other Risk Management Tools at www.acec.org/bookstore

You can also browse all of the CASE publications at www.acec.org/coalitions/coalition-publications/

Is there something missing for your business practice? CASE is committed to publishing the right tools for you. Have an idea? We’d love to hear from you!

Upcoming Events

Managing Small Projects Successfully: How to Prevent Small Projects from Becoming Big Problems

October 31 – November 9, 2023

Online

For engineering firm project managers and firm principals, smaller projects can be a core revenue driver. But, smaller projects still have the potential to carry big risk that can be a drag on resources, profitability, and client satisfaction. The good news is that, with the right set of skills in your toolbox, you can ensure that even the smallest projects deliver maximum profits. Register now for Managing Small Projects Successfully: How to Prevent Small Projects from Becoming Big Problems and learn the skills, hacks, secrets, formulas, trouble-shooters and problem-solvers that make engineering firm executives and clients delighted with small project progress and outcomes.

From planning, scheduling and budgeting to risk control and crisis management, this live online program packs everything you need into just 8 hours of instruction, broken into twohour sessions to work with your busy schedule. Even better, it is packed with proven insight from the engineering project management experts at PSMJ Resources, Inc. Earn up to 8 PDHs!

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The Structural Coalition at ACEC has several committees that meet regularly to develop documents that help guide engineers in their business practice. Committee member commitments include a monthly virtual meeting, a few hours a month working on relevant documents, and travel to the Coalitions winter and summer meetings!

If you are interested in joining a committee or have any questions, contact coalitions@acec.org.

Now more than ever we need to support the upcoming generation of the workforce. Give to the CASE Scholarship today!

CODES and STANDARDS

FAQ on SEI Standards

FAQ on SEI Standards

What you always wanted to ask.

By Jennifer Goupil, P.E., F.SEI, F.ASCE

his article addresses some of the questions received about structural standards the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE) In addition, questions from engineers, officials, and other are often considered in the of future editions of the Standards are some questions received with SEI’s

This quarterly article addresses some of the questions received about structural standards developed by the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE). In addition, questions from engineers, building officials, and other design professionals are often considered in the development of future editions of the Standards. Following are some questions received along with SEI’s responses to clarify the provisions.

Important Errata

What is the correct lower bound equation for the seismic acceleration coefficient Cpi?

In ASCE/SEI 7-16 and ASCE/SEI 7-22, 12 Seismic Design Structures Section 12.10, the value is calculated using the greater value Equations (12.10-8 and 12.10-9). The states the indicated in the Equation (12.10-8). Is the 0.8 to be

In ASCE/SEI 7-16 and ASCE/SEI 7-22, Chapter 12 Seismic Design Requirements for Building Structures Section 12.10, the value for design accelerations coefficients Cpi is calculated using the greater value from Equations (12.10-8 and 12.10-9). The Commentary states the reasons for providing a lower bound equal to Cpo and NOT 0.8Cp0 as indicated in the Equation (12.10-8). Is the coefficient 0.8 to be applied in determining the lower bound of coefficient Cpi?

Design acceleration coefficients Cpo, Cpi, and Cpn values are used to calculate seismic design forces for diaphragms, chords, and collections. Equations (12.10-8) and (12.10-9) are used to determine the greater value between these two equations for the value of Cpn used in (Equation 12.10-7). Currently, there is a discrepancy between the Equation and the Commentary. Per the Commentary, the lower bound value of Cpi is Cpo. Reviewing balloting for this value confirms the lower bound value of Cpi = Cpo. This lower bound value agrees with the Federal Emergency Management Agency’s FEMA P-1051, Chapter 6: Horizontal Diaphragm Analysis (2015 National Earthquake Hazard Reduction Program’s,

NEHRP, Recommended Seismic Provisions: Design Examples), which provided recommendations to ASCE/SEI 7-16. Figure 12.10-2 in ASCE/ SEI 7-16 and ASCE/SEI 7-22 are correct. The upper bound value for Cpo remains unchanged, as indicated in Equation (12.10-9). This correction will be issued as an errata for ASCE/SEI 7-16 and ASCE 7-22 editions.

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ASCE/SEI 7: Minimum Design Loads and Associated Criteria for Buildings and Other Structures

Figure 12.10-2 Calculating the Design Acceleration Coefficient Cpx in Buildings with N ≤ 2 and in Buildings with N ≥ 3

Jennifer Goupil, P.E., F.SEI, F.ASCE, is the Managing Director of the Structural Engineering Institute and the Chief Resilience Officer for the American Society of Civil Engineers.

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