STRUCTURE magazine | June 2017 by structuremag

A Joint Publication of NCSEA | CASE | SEI

STRUCTURE

June 2017 Tall Buildings Inside: MoMA Tower, New York City

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CONTENTS Features

44 Salesforce Tower

33 A Unique Opportunity for a Conventional Tall Building Frame By Mark Sarkisian, S.E., Neville Mathias, S.E., and David Shook, P.E.

By Ron Klemencic, P.E., S.E., Hon. AIA, Michael T. Valley, P.E., S.E., and John D. Hooper, P.E., S.E.

50 Brock Commons

36 New Clearwater Beach Hotel By E. Michael McCarthy, P.E.

By Paul Fast, P.E., P.Eng., Struct.Eng. and Robert Jackson, P.Eng.

40 53W53 – MoMA Tower

54 Flat Plate Concrete Construction

By Silvian Marcus, P.E., Gustavo J. Oliveira, P.E., Fatih Yalniz, P.E, and Nicholas Chack, P.E.

By Mark Sarkisian, S.E., Eric Long, S.E., David Shook, P.E., and Eric Peterson

59 SPECIAL SECTION: Tall Building Construction

Columns and Departments EDITORIAL

7 SEI’s Vision for the Future of Structural Engineering By Andrew W. Herrmann, P.E., SECB STRUCTURAL DESIGN

8 Seismic Design of Nonbuilding Structures and Nonstructural Components – Part 2 By J. G. (Greg) Soules, P.E., S.E., P.Eng., SECB STRUCTURAL PRACTICES

12 Development Along Old Party Walls By Dan Eschenasy, P.E., SECB

SPOTLIGHT

STRUCTURAL SYSTEMS

20 Stiffness Versus Strength

By Joseph Savalli, P.E., Matthieu

Keith M. MacBain, Ph.D., P.E.

Peuler, P.E., and Leslie Morris, P.E.

PRACTICAL SOLUTIONS

24 Wood Shear Wall Design Examples for Wind By John “Buddy” Showalter, P.E. STRUCTURAL FAILURES

29 Failure of Imagination By Stan R. Caldwell, P.E., SECB. INSIGHTS

70 Punching of Slabs By Aurelio Muttoni, Ph.D., Miguel Fernández Ruiz, Ph.D.

Hussien Abdel-Baky, Ph.D., P.E., and James C. Hays, P.E.

BUSINESS PRACTICES

86 Techniques to Successfully Navigate Networking By Jennifer Anderson

IN EVERY ISSUE 6 Advertiser Index 67 Resource Guide – Tall Buildings 80 NCSEA News 82 SEI Structural Columns 84 CASE in Point

and João T. Simões HISTORIC STRUCTURES

72 Sciotoville Bridge By Frank Griggs, Jr., D.Eng., P.E.

By Giulio Leon Flores, P.E.,

79 Madison Square Park Tower

By Paul A. Gossen, P.E. and

STRUCTURAL TESTING

17 Temporary Testing Facility

By Larry Kahaner

LEGAL PERSPECTIVES

75 An Overview of Consent to Assignment Agreements

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, C 3 Ink, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

By Gail S. Kelley, P.E., Esq.

On the cover With a slenderness ration of 1:12, MoMa Tower will rise to a total height of 1,050 feet on the New York skyline. Read about the many challenges facing the project’s Structural Engineers on page 40.

STRUCTURE magazine

June 2017

The 1,070-foot-tall Salesforce Tower advances the state-ofthe-art of high-rise seismic design. See page 44.

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Erratum Historic Structures – Lachine Rapids Bridge (April 2017) In this Historic Structures article, the explanation of Long’s use of continuous bridge segments across supports (page 42) contained an error: “In other words, he made his truss continuous over the intermediate supports so that the maximum positive moment (should read negative moment) over the support is equal, or nearly so, to the maximum negative moment (should read positive moment) located between the supports.” STRUCTURE magazine and the author apologize for the error. The online version of this article has been corrected. Thank you to H. Starzer, P.E. for pointing out the error.

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STRUCTURE magazine

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EDITORIAL BOARD Chair Barry K. Arnold, P.E., S.E., SECB ARW Engineers, Ogden, UT chair@structuremag.org Jeremy L. Achter, S.E., LEED AP ARW Engineers, Ogden, UT Erin Conaway, P.E. SidePlate Systems, Phoenix, AZ John A. Dal Pino, S.E. FTF Engineering, Inc., San Francisco, CA Linda M. Kaplan, P.E. TRC, Pittsburgh, PA Dilip Khatri, Ph.D., S.E. Khatri International Inc., Pasadena, CA

Brian W. Miller Davis, CA Mike Mota, Ph.D., P.E. CRSI, Williamstown, NJ Evans Mountzouris, P.E. The DiSalvo Engineering Group, Ridgefield, CT

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June 2017

Greg Schindler, P.E., S.E. KPFF Consulting Engineers, Seattle, WA Stephen P. Schneider, Ph.D., P.E., S.E. BergerABAM, Vancouver, WA John “Buddy” Showalter, P.E. American Wood Council, Leesburg, VA C3 Ink, Publishers A Division of Copper Creek Companies, Inc. 148 Vine St., Reedsburg WI 53959 Phone 608-524-1397 Fax 608-524-4432 publisher@structuremag.org June 2017, Volume 24, Number 6 ISSN 1536-4283. Publications Agreement No. 40675118. Owned by the National Council of Structural Engineers Associations and published in cooperation with CASE and SEI monthly by C3 Ink. The publication is distributed free of charge to members of NCSEA, CASE and SEI; the non-member subscription rate is $75/yr domestic; $40/ yr student; $90/yr Canada; $60/yr Canadian student; $135/yr foreign; $90/yr foreign student. For change of address or duplicate copies, contact your member organization(s) or email subscriptions@ STRUCTUREmag.org. Note that if you do not notify your member organization, your address will revert back with their next database submittal. Any opinions expressed in STRUCTURE magazine are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, C3 Ink, or the STRUCTURE Editorial Board. STRUCTURE® is a registered trademark of National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

Editorial

SEI’s Vision for the Future of new trends, new techniques and current industry issues Structural Engineering

Where Do We Stand Author ? by-line

By Andrew W. Herrmann, P.E., SECB, F.SEI, Pres.12.ASCE

Promote Performance-Based Codes and Standards

EI just finished its successful 2017 Structures Congress in Denver where we celebrated collaboration, innovation, and the Vision for the Future of Structural Engineering: A Case for Change. Reflecting on the program, sessions covered areas such as communications, mentoring, ethics, codes and standards, history, performance, multi-hazards, training, professional practice, health monitoring, sustainability, resiliency, construction, loads, rehabilitation, and solutions. All areas are meant to contribute towards the future of structural engineering. SEI’s A Vision for the Future of Structural Engineering and Structural Engineers: A Case for Change was released on October 16, 2013, and serves as SEI’s blueprint for the future. For the full version of the report, refer to www.asce.org/SEI. The executive summary, dated September 2015, is also available there. As part of the Congress, the SEI Board of Governors met to assess our progress and identify areas where we need to increase attention. This primarily includes progress on the seven key initiatives that will enable SEI to achieve the Vision for the Future of Structural Engineering. The following is our status towards fulfilling our key initiatives to advance our vision:

A Task Committee on Performance-Based Design is reviewing the final drafts of white papers that will define the path required to give structural engineers new tools to liberate them from the limitations of prescriptive code-checking, encourage innovation in their designs, and increase the value of their services.

Lead Multi-Disciplinary Summits on Technical Matters of Broad Interest The Committee on Biennial Interdisciplinary Technology Summits has been formed to think outside of the traditional boundaries of structural engineering to identify and apply the most advanced new technologies and science to the practice.

Promote the Structural Engineer as a Leader and Innovator Each of the preceding responses to key initiatives contributes to promoting the structural engineer as a leader and innovator by supporting and encouraging the expansion of members’ roles to recognized positions of leadership in society by equipping them with the tools they need to succeed and be recognized by the public. SEI is also initiating new digital branding to communicate our vision.

Reform Structural Engineering Education The Committee on the Reform of Structural Engineering Education (CROSEE) has been formed to study new educational models to equip students with the broad technical, communications, and critical thinking skills necessary to compete in the global economy.

Advocate for Structural Engineering Licensure

Improve Mentoring and Continuing Education The Committee on Professional Mentoring has been rebranded as the Leadership Development Committee within the SEI Business Practice Activities Division (BPAD). The committee is developing a national, standardized framework to launch the careers of young professionals and create a meaningful platform for lifelong learning and constant professional growth. See the March 2017 STRUCTURE magazine Editorial for more information highlighting this committee. The Continuing Education Task Committee received a grant this year from the SEI’s Futures Fund and is scheduling a workshop in June to explore the direction of continuing education for Structural Engineers.

We have joined with our peers, NCSEA, CASE, and SECB, to form the Structural Engineering Licensure Coalition (SELC) to advance implementation of the SE license as a post-PE credential. This fall, five years since the publication of the original vision, SEI will formally evaluate where we stand, what we have accomplished to date, and what we need to do to continue to achieve the Vision’s goals.▪ You can help advance the Vision for the Future of Structural Engineering in two ways. Take the first step of influencing our structural engineering profession by joining an SEI committee at www.asce.org/structural-engineering/sei-committees and, secondly, support the vision with a gift to the SEI Futures Fund which provides critical funding support to many of our initiatives.

Create a New SEI Global Activities Division A business plan and budget have been approved by the SEI Board of Governors for the newly formed Global Activities Division. This new division will address the needs of a worldwide membership, and position our members as global leaders in structural engineering research and practice. Meetings have also been initiated with Structural Engineering organizations outside the United States to seek ways to collaborate. STRUCTURE magazine

If you would like to learn more about the SEI Vision for the Future of Structural Engineering or SEI, please contact SEI Director Laura Champion at lchampion@asce.org or SEI President Andy Herrmann at aherrmann@hardesty-hanover.com.

June 2017

Structural DeSign design issues for structural engineers

his article is the conclusion of a twopart series which discusses the seismic design provisions for nonbuilding structures found in Chapter 15 of ASCE 7-16, Minimum Design Loads and Associated Criteria for Buildings and Other Structures. The previous article (Part 1, STRUCTURE, April 2017) provided an introduction to the seismic design of nonbuilding structures. Several seismic related issues are unique to nonbuilding structures. This article covers the following advanced topics in the seismic design of nonbuilding structures: • The determination of seismic forces on nonbuilding structures supported by other structures. • The determination of seismic forces on common nonstructural components attached to nonbuilding structures. • The interrelation and overlap between Chapter 13, Seismic Design Requirements for Nonstructural Components, and Chapter 15 of ASCE 7-16. • Special considerations for the seismic design of tanks and vessels.

Seismic Design of Nonbuilding Structures and Nonstructural Components Part 2: Advanced Topics related to ASCE 7-16 By J. G. (Greg) Soules, P.E., S.E., P.Eng., SECB, F.SEI, F.ASCE

J. G. (Greg) Soules is a Principal Engineer with CB&I LLC in Houston, Texas. He is the Vice Chair of the ASCE 7-16 Main Committee, Vice Chair of the ASCE 7-16 Seismic Subcommittee, and Chair of the ASCE 7-16 Task Committee on Nonbuilding Structures. He can be reached at Greg.Soules@cbi.com.

Nonbuilding Structures Supported by Other Structures Section 15.3 of ASCE 7-16 provides requirements for the design of nonbuilding structures supported by other structures for seismic forces, and presents three possible scenarios: • The nonbuilding structure weight is less than 25 percent of the combined weight of the nonbuilding structure and the supporting structure (15.3.1). • The nonbuilding structure weight is greater than or equal to 25 percent of the combined weight of the nonbuilding structure and the supporting structure (15.3.2(1)) – rigid nonbuilding structure (T < 0.06 seconds). • The nonbuilding structure weight is greater than or equal to 25 percent of the combined weight of the nonbuilding structure and the supporting structure (15.3.2(2)) – flexible nonbuilding structure (T ≥ 0.06 seconds). Nonbuilding structures supported by other structures see amplified seismic forces in a similar manner as nonstructural components. To discuss the seismic design of nonbuilding structures supported by other structures, a review of the determination of seismic forces on nonstructural components is important.

June 2017

Nonstructural Components Section 13.3.1 of ASCE 7-16 specifies the use of Equation 13.3-1 (shown below) to determine the seismic design force on a nonstructural component.

(

)

0.4apSDSWp 1+2 z Eqn. 13.3-1 h Rp Ip Fp shall not to be taken as less than: Fp = 0.3SDSIpWp Fp is not required to be taken as greater than: Fp = 1.6SDSIpWp where: Fp = seismic design force ap = component amplification factor that varies from 1.0 (rigid component Tp < 0.06 seconds) to 2.5 (flexible component). Tp is the fundamental period of the component. Rp = component response modification factor (same concept as R for structures) Ip = component importance factor (1.0 or 1.5). Ip is not necessarily the same as the value of IE for the supporting structure. SDS = short period spectral acceleration Wp = component operating weight z = height in structure of point of attachment of component with respect to the base. h = average roof height of structure with respect to the base Fp =

( )

The values of ap and Rp are taken from Table 13.5-1 for architectural components or Table 13.6-1 for mechanical and electrical components. Various terms in Equation 13.3-1 have significant physical meanings. The term 0.4apSDS represents the peak ground acceleration when ap equals 1.0 and the constant acceleration region of the response spectrum (plateau) when ap equals 2.5. The term (1 + 2z /h) represents an additional amplification of the ground motion acceleration due to the elevation of the point of attachment of the supporting structure.

25 Percent Limitation Where the weight of the supported nonbuilding structure is less than 25 percent of the combined effective seismic weights of the nonbuilding structure and supporting structure, the design seismic forces of the supported nonbuilding structure are determined according to Chapter 13 where the values of Rp and ap are determined per Section 13.1.5. Equation 13.3-1 is used to calculate the seismic force, Fp, on the supported nonbuilding structure. The supporting structure is designed to the requirements of Chapter 12, Seismic Design Requirements for Building Structures, or Section 15.5, Nonbuilding Structures Similar to Buildings, as appropriate, with the weight of the supported nonbuilding structure considered in the determination of the

effective seismic weight, W. Section 15.3 represents a clear dividing line between Chapter 13 and Chapter 15 where the nonbuilding structure is supported by another structure.

More than 25 Percent with Rigid Nonbuilding Structure Where the fundamental period of the supported nonbuilding structure, T, is less than 0.06 seconds, the supported nonbuilding structure is considered to be a rigid element. In this case, the supporting structure is designed to the requirements of Chapter 12 or Section 15.5 as appropriate, and the R-value of the combined system is permitted to be taken as the R-value of the supporting structural system. The supported nonbuilding structure is simply taken as another mass in the design of the supporting structure. This procedure is similar to that used for the case where the supported nonbuilding structure is less than 25 percent of the combined mass. The supported nonbuilding structure and its attachments are designed for the forces determined using the procedures of Chapter 13, where the value of Rp is taken as equal to the R-value of the nonbuilding structure

as outlined in Table 15.4-2, and ap shall be taken as 1.0. It is important to note that very few supported nonbuilding structures qualify as rigid elements. There is a great temptation to assume that the supported nonbuilding structure is rigid due to the resulting ease of calculation and lower loads. The period of the supported nonbuilding structure must be honestly evaluated, taking into account such items as fluid-structure interaction and the flexibility of the supporting floor beams. Procedures for taking fluid-structure interaction into account can be found in TID-7024 (1963).

More than 25 Percent with Flexible Nonbuilding Structure Where the fundamental period of the supported nonbuilding structure, T, is greater than or equal to 0.06 seconds, the supported nonbuilding structure is considered to be a flexible element. In this case, the nonbuilding structure and supporting structure are modeled together in a combined model with appropriate stiffness and effective seismic weight distributions. The combined structure is designed to Section

15.5, with the R-value of the combined system taken as the lesser R-value of the nonbuilding structure or the supporting structure. The supported nonbuilding structure and its attachments are designed for the forces determined for the supported nonbuilding structure in the combined analysis. A flexible nonbuilding structure supported by another structure is by far the most common situation. Because the combined structure is designed using the lesser R-value of the supported nonbuilding structure or the supporting structure, the use of a high R-value structural system (e.g. special concentrically braced frame) offers no economic advantage. Of course, a high R-value structural system may always be used to provide better performance. The use of a combined model requires that the structural engineer designing the supporting structure work in close collaboration with the manufacturer of the supported nonbuilding structure. The combined model does not have to be complex. An example of this type of combined model can be found in Appendix 4.G of ASCE Guidelines for Seismic Evaluation and Design of Petrochemical Facilities (2011).

continued on next page

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June 2017

Common Nonstructural Components Attached to Nonbuilding Structures Table 13.6-1 (Mechanical and Electrical Components) and Table 13.5-1 (Architectural Components) contain the basic seismic parameters (ap and Rp) for many common nonstructural components. Occasionally, the engineer will run into cases where specific values for the components are not listed. In this case, it is best to use “other mechanical or electrical components” from Table 13.6-1 or, in the case of an architectural component, use values from “other rigid components” or “other flexible components” from Table 13.5-1. For mechanical or electrical components not listed in Table 13.6-1, the category of “other mechanical or electrical components” provides a simple, although conservative, solution by using ap of 1.0 and Rp of 1.5. Engineers often try to use values for components in Table 13.6-1 that they feel are similar to their component. The engineer takes on some risk in using this approach because the descriptions of the components in Table 13.6-1 are not very detailed. An example can be seen in trying to choose values for a fin fan. A fin fan is a type of air cooler with integral support legs that is often supported on pipe racks. The values listed for fans in Table 13.6-1 (ap = 2.5 and Rp = 6) are not intended for fin fans with integral support legs (these values do apply where fin fans are not supported on integral support legs). Fin fans with integral support legs have been added to Table 13.6-1 (ap = 2.5 and Rp = 3) in ASCE 7-16. It was necessary to specifically add an entry, with significantly reduced values, for fin fans with integral support legs to ASCE 7-16 due to the fans’ poor performance in seismic events, such as the February 27, 2010, Chile earthquake (Soules, Bachman, and Silva, 2016). When in doubt, and when you cannot match your component to an exact description in Table 13.6-1, you should select the “other mechanical or electrical components” category. For architectural components not listed in Table 13.5-1, the multiple choices provided under “other rigid components” or “other flexible components” require engineering judgment. The engineer must first decide if the component is rigid or flexible. This decision should be based on an approximate natural period, Tp, for the component. The engineer must then decide if the elements and attachments of the component are high-deformability, limited-deformability, or low-deformability. Section 11.2 provides definitions of high-, limited-, and low-deformability regarding

the ratio of the ultimate deformation to the limit deformation. These definitions, while precise, are not straightforward to apply. Fortunately, the commentary to Chapter 13 provides some guidance. For example, the commentary notes that high-deformability materials are materials such as steel or copper that can accommodate relative displacements inelastically if the connections also provide high-deformability. Therefore, the types of connections used are critical in the classification process. As an example, steel walkways and steel platforms are commonly attached to nonbuilding structures in industrial facilities. While the steel walkways and platforms are constructed of a high-deformability material, the connections often are not seismically detailed and frequently include short attachment columns with limited ability to absorb inelastic deformations. Most configurations would also qualify as flexible. Therefore, a reasonable recommendation for values of ap and Rp for steel walkways and platforms are ap = 2.5 and Rp = 2.5, which corresponds to “other flexible components” and “limited-deformability elements and attachments.”

Tanks and vessels are nonbuilding structures not similar to buildings. As such, they exhibit a very different dynamic response than building structures. There are four special considerations for tanks and vessels: 1) The importance of anchor rod stretch. 2) The importance of providing seismic freeboard. 3) The importance of providing piping flexibility. 4) Special design requirements for vessel support skirts.

Chapter 13 or Chapter 15?

Anchor Rod Stretch

As described earlier, ASCE 7-16 Section 15.3 provides a clear delineation between Chapter 13 and Chapter 15 for nonstructural components and nonbuilding structures supported by other structures, based on the weight of the supported nonstructural component or nonbuilding structure. Unfortunately, the same cannot be said of certain nonstructural components and nonbuilding structures supported at grade and common to both chapters. The following recommendations attempt to address this lack of clear delineation between Chapter 13 and Chapter 15. The most informative reference for deciding whether to use Chapter 13 or Chapter 15 is Nonstructural Component or Nonbuilding Structure? (Bachman and Dowty, 2008). This resource identifies the common components covered by both Chapter 13 and Chapter 15 as: • Billboards and Signs • Bins • Chimneys • Conveyors • Cooling Towers • Stacks • Tanks • Towers • Vessels Bachman and Dowty also suggest three ways to differentiate between nonstructural components and nonbuilding structures:

Many nonbuilding structures rely on the ductile behavior of anchor bolts to justify the R-value assigned to the structure. Anchor bolts used for tanks and vessels must stretch under seismic loads to provide the required ductility. Section 15.4.9 provides a consistent treatment of anchorage on nonbuilding structures. Anchors must be designed to be governed by the tensile strength of a ductile steel element. Post-installed anchors in concrete or masonry must be pre-qualified for seismic applications. Section 15.7.3 is intended to ensure that anchor attachments are designed such that the anchor will yield (stretch) before the anchor attachment to the structure fails. Under Section 15.7.3, connections, excluding anchors (bolts or rods) embedded in concrete, must be designed to develop Ω0 times the calculated connection design force. Section 15.7.5 requires anchorage to meet the requirements of Section 15.4.9, whereby the anchor embedment into the concrete must be designed to develop the tensile strength of the anchor. The anchor must have a minimum gauge length (stretch) of eight diameters. The load combinations with overstrength of Section 12.4.3 are not to be used to size the anchor bolts for tanks, or horizontal and vertical vessels. Oversized anchors are not able to stretch and, therefore, do not provide the required ductility.

STRUCTURE magazine

June 2017

• Size – nonstructural components are small, usually less than 10 feet in height • Construction – nonstructural components are typically shop fabricated • Function – nonstructural components are primarily designed for functionality while nonbuilding structures are primarily designed to maintain structural stability

Tanks and Vessels

Seismic Freeboard The impact of a sloshing wave on the tank roof or forcing the floating roof into a fixed roof is a continuing source of seismic damage to ground supported storage tanks. Occasionally, external floating roofs are forced outside of the tank shell by the sloshing wave and end up landing on the shell or having the seal catch the shell. Loss of a floating roof in any of these cases often results in a fire. This damage can be eliminated by providing sufficient seismic freeboard.

Piping Flexibility The lack of flexibility in piping connections to tanks is a continuing source of seismic damage to ground supported storage tanks. Therefore, ASCE 7 requires piping systems connected to tanks and vessels to be flexible enough to take specified displacements as noted in Table 15.7-1. The piping must be able to accommodate these movements at allowable stress levels. The piping must also be able to accommodate the amplified movements (Cd times

the values in the tables) without rupturing. Experience shows that systems with little or no flexibility fail in large seismic events and systems with flexibility built-in perform well.

Vessel Support Skirts Skirt supported vessels fail in buckling, which is not a ductile failure mode. Therefore, a more conservative design approach is required. To prevent collapse, ASCE 7 Section 15.7.10 and Table 15.4-2 require skirt supported vessels to be checked for seismic loads based on R/I = 1.0 if the structure falls in Risk Category IV or if an R-value of 3.0 is used in the design of the vessel. The R/I = 1.0 check will typically govern the design of the skirt over using loads determined with an R-factor of 3 in a moderate to high area of seismic activity. The foundation and anchorage are not required to be designed for the R/I = 1.0 load.

components. Key takeaways from this article include: • Seismic forces on nonbuilding structures supported by other structures are determined by the size and stiffness of the supported nonbuilding structure. • The choice of design coefficients for nonstructural components is a function of the deformability of the element and its connection. • The applicability of Chapter 13 or Chapter 15 can be determined based on the size, construction, and function of the component or nonbuilding structure. • The performance of tanks and vessels in a seismic event depends heavily on the anchorage details used, the use of seismic freeboard, the use of flexible piping connections, and the proper design of skirt supports.▪

Conclusion The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

This article provides an overview of some advanced topics encountered in the design of nonbuilding structures and nonstructural

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June 2017

4/28/17 9:21 AM

Structural PracticeS practical knowledge beyond the textbook

arty walls are frequently used in low-rise developments due to their capacity to provide economical structural support in denser areas. They were more prevalent and taller in earlier times when American cities developed. One can find many old party walls in cities along the Atlantic seaboard – from Portland, Maine, to Charleston, South Carolina. In the older areas of New York City, there are over 15,000 existing brick masonry party walls, many in four- or five-story buildings. The shared use by adjoining owners resulted in real property law developments which have added constraints that FEMA 547 Techniques for the Seismic Rehabilitation of Existing Buildings considers overriding. FEMA 547 justifies its lack of guidance: “For conditions along property lines or involving party walls, the two buildings likely have different ownership, and practical and legal issues may be more significant than technical ones.” This article introduces the reader to historical party walls and

Development Along Old Party Walls By Dan Eschenasy, P.E., SECB, F.SEI

presents construction and engineering challenges encountered on construction sites that border old masonry party walls. New York City Building Code (NYCBC) regulations are discussed, but nothing in this article should be interpreted as real property law advice.

A Short History of Regulations

Dan Eschenasy is the New York City Buildings Department Chief Structural Engineer. He is an Honorary Member of SEAoNY and a member of the ASCE Structural Assessment of Buildings Committee.

We know from the Roman engineer, Vitruvius, in his famous treaty De Architectura, that party walls already existed in Rome around 30 B.C. They were used as bearing walls set along property lines to support floors of adjoining structures. Roman legal statutes from that period established the rights and obligations of owners of party walls. The first regulation involving party walls as fire barriers was issued in the aftermath of the Great Fire of London (1666). To combat the “mischief of fire,” laws were promulgated requiring the use of masonry in perimeter walls, including party walls. Thus, the structural bearing function became intertwined with the fire containment function. Party wall legal theories that developed separately in the different states in colonial America are the source of differences from state to state in legal practices that exist today. In 1791, George Washington promulgated a party wall regulation for the District of Columbia. The wood frame construction of the original New York City (NYC) buildings allowed several major fire conflagrations to occur. As a consequence, use of brick or stone in perimeter

12 June 2017

walls was mandated by fire ordinances dating as far back as 1830. By 1900, row houses using party walls covered block after block. Even when building heights had reached six or seven stories, as long as the construction occurred on narrow lots, masonry party walls remained the choice of builders. Since the 1870s, successive NYC building regulations have included specific instructions for party walls. There were 20 entries in the 1901 NYCBC that were mostly prescriptions for the construction of party walls. Around the turn of the 20th century, fire science had developed as a separate field, and the 1905 National Building Code developed by the National Board of Fire Underwriters (NBFU) and the American Insurance Association had 9 entries similar to those in the 1901 NYCBC. Up to the 1916 NYCBC, the mandated party wall thickness was the result of empirical structural considerations related to applied weight. Following the lead of the NBFU, the 1916 NYCBC introduced the concept that firewalls be specified according to mandated fire tests. Several technical developments led to the near elimination of specific party wall structural provisions in building codes, such as when the empirical design of masonry was replaced by an engineered design of masonry, and when steel and concrete frames replaced masonry as bearing systems for tall structures. The 2014 NYCBC uses the IBC definition: “Any wall located on a lot line between adjacent buildings, which is used or adapted for joint service between the two buildings, shall be constructed as a fire wall (sic).” Although not always explicitly stated in codes, a party wall needs to meet both firewall and material specific structural design requirements; in the case of a fire in one building, the wall is expected to maintain its structural stability and stop the fire from spreading to the adjoining building. The usual problems of demolition and excavation along buildings on lot lines are amplified when a party wall lies on that line. The structural function of the wall needs to be preserved. Also, weatherproofing needs to be added. Almost all of the references in the 2014 NYCBC are prescriptions for the protection of existing party walls during construction or demolition. Several other jurisdictions (Philadelphia, Washington D.C., and more) have similar regulations for the protection of party walls.

Existing Party Walls Attached unreinforced masonry buildings sharing a party wall constitute a more stable unit than separated buildings. The larger footprint of the attached

PARTY WALL

BOWED FAÇADE WALL

Figure 1. Crack in party wall.

Locating Existing Party Walls In addition to legal constraints, old party walls might constitute spatial constraints as their presence limits buildable area or influences the location of columns or shear walls. The “discovery” of a party wall in the construction phase results in complicated changes and significant delays. It is essential to recognize the presence of such walls at the preliminary stage of a project. When one plans to develop a new building on a lot that is still occupied, an exterior topographical survey may not reliably determine the presence or even the exact thickness of a party wall. Even more, such a wall might not lie exactly centered on the property line. Rarely do original construction drawings exist, and current owners may be unaware and may only infer the presence of a party wall. Since the thickness of two abutting but independent walls can be larger than that of a party wall, a simple probe may be sufficient to elucidate the situation. In many cases, adjoining owners

ORIGINAL PARTY WALL

EXISTING UNDERPIN COLLAPSED OLD PIN

Figure 3. Old pin collapsed.

may not accept acquiring additional data, especially when the measurement requires destructive probes. In older New York City neighborhoods, as a result of successive development on the same site, it is not uncommon to find party walls extending or incorporating older party walls (Figure 2). Depending on the shape of the original building, these remnant walls may not be continuous or may not span the entire length of the more recent wall, and may be missed by probes. A serious accident that took place during the underpinning of the foundation of a fifteen-story loft in midtown Manhattan exemplifies the danger of not exploring the layered history of party walls adjoining a construction site. The steel frame loft had been erected around 1926 on a site previously occupied by a masonry building that belonged to a group of attached 1880s tenements. When originally constructed, the loft’s masonry incorporated the party wall of the attached remaining tenement building. Also, this 1880s wall was underpinned to

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structure leads to much longer shear walls. In addition, whatever the direction of the out-ofplane load on the party wall, these loads can be transferred by compression to the floor diaphragms on the opposing side. An incident involving an underpinning operation provided proof of the higher reliability of party walls. An improper construction procedure led to the partial collapse of a rubble foundation that was supporting a party wall separating two historic buildings. A bottom section of approximately 25 feet collapsed, leaving about 15 feet of rubble foundation standing at each end. The soil underlying this remaining foundation was competent and the three-story unreinforced masonry wall was able to turn into an arch spanning 25 feet above the collapsed area. Most independent walls would have failed, but the wood floors on both sides of this wall maintained its geometric stability and the restraint provided by the floors allowed the wall to sustain the significant increase in compressive stresses. Many owners may not be aware of the obligation to maintain, in common, the structural as well as the fire separation functions of party walls. An interesting incident occurred in 2009 when the inspectors determined, while responding to a complaint about a facade bowing,

that the facade was common to two attached buildings. The interior inspection revealed a 5-inch crack (Figure 1) that had developed along the line where the party wall used to be keyed into the facade. Each building was owned separately, and the owners could not reach an agreement to repair the crack. The fire separation was compromised and the collapse of the facade onto the street was imminent. As the engineers could not find a solution to arrest the evolution of the bowing, the facade was ordered demolished.

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ORIGINAL PARTY WALL (in white)

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Figure 2. Building incorporating old party wall.

STRUCTURE magazine

June 2017

older attached buildings were erected by the same developer and each building was subsequently sold to different individuals. Although no Party Wall Agreement was signed, the owners’ legal rights and obligations are implied by the mere existence of the wall and its common use.

LOT LINE ANCHOR

FILL GAP MATERIAL 25 PSI COMPRESS

WOOD FLOOR CONCRETE SHEAR WALL

Adjoining Demolition

Figure 4. Wall collapsed due to lack of anchorage to the diaphragm.

accommodate the deeper basement of the loft. In 2014, the remaining tenement was demolished and excavation for a new building commenced. The basement of this new development ran even deeper than the loft’s basement, and an underpinning installation started. It appeared that the contractor was not fully aware that he was, in fact, undermining an older underpinning job. While he was digging, an old unattached pin overturned and collapsed with tragic consequences (Figure 3). When the existence of a party wall is confirmed, the proper course of action is to obtain acceptance from the adjoining owner for any construction involving this wall. The acceptance of the proposed work is formalized in a legal document called the Party Wall Agreement. Recently, the NYCDOB issued a standardized form. Such agreement is necessary when owners of adjoining lots decide to build a lot line wall to be used in common. In many cases,

Figure 5. Noggin wall.

Demolition along a party wall results in new structural conditions and exposure to potential adverse weather conditions. As a result, the party performing demolition is required to utilize a series of proactive measures. Since 1968, the NYCBC required: “where the floor beams of the adjacent building bear on the party wall, the person causing the demolition shall ascertain that such beams are anchored into the wall and, where such anchorage is lacking, shall provide anchorage or otherwise brace the standing wall.” The validity of this requirement was confirmed by several recent wall collapse investigations that determined that these party walls were not anchored when the adjoining building was demolished (Figure 4). As explained in the Collapse of Masonry Structures under NonExtreme Loads [Eschenasy, Second Applied Technology Council’s (ATC) & the Structural Engineering Institute’s (SEI) Conference on Improving the Seismic Performance of Existing Buildings and Other Structures], forensic investigations determined that the collapses were not due to over-stress but to loss of stability. An engineering investigation is required before placing any special load on a party wall, for instance when the basement of a demolished building is filled with soil. Additional bracing of a wall may become necessary when the wall is not plumb or starts to lean following demolition or excavation. Demolition adjoining “nogging” party walls poses special difficulties. The noggin (or nogging) wall is an assembly where the space between wood studs is filled with random bricks (Figure 5). Building regulations up to 1916 allowed this type of construction to be used as a building separation. Today, there are still a good number of such walls that exist. Vinyl siding has been, and still is today, the owners’ primary choice for weather proofing noggin walls. Investigations of several nogging wall collapses blamed significant rot of the studs resulting from rainwater that had penetrated behind the vinyl cladding. In contrast to regular wood stud walls where the air around the studs allows evaporation of the water, the bricks that are set tight against the studs preclude evaporation. Even worse,

STRUCTURE magazine

June 2017

RUBBLE WALL EXISTING STONE SHELF UNDERPIN NEW SOIL ANCHOR

CUT TOE NEW BASEMENT

Figure 6. Typical lot line condition.

the deteriorating conditions are not usually observed in time due to the mask provided by the vinyl siding. In addition to improving weather proofing, a proper solution to stabilizing these noggin walls is to brace them with properly nailed wood boards.

Developing an Adjoining Structure As long as masonry was the main structural bearing material, the vertical extension of party walls was permitted under the condition that the existing wall was lined with additional brick to meet the total required thickness for the new height. When, at the turn of the 20th century, masonry bearing walls for high-rise buildings were abandoned in favor of steel frames, engineers faced different challenges when working in the vicinity of masonry party walls. As early as 1912, the Kidder Parker Handbook observed that “when buildings of skeleton construction are erected without a party wall agreement, it is usually impossible to obtain a symmetrical foundation directly under the columns supporting the side or party wall.” The Handbook recommended cantilever foundations. The 1915 National Building Code indicated: “Where an existing party wall is to be incorporated in a new building of skeleton or curtain wall construction, the vertical extension of the existing party wall shall be supported entirely by columns and girders.” Many incidents and delays occur during excavations when the configuration of the party wall foundation is unknown. It is essential to use an exploratory pit to understand the shape of the existing footing.

well. At the time of their erection, these walls saved space. However, today, when the same lots are used for redevelopment, the presence of party walls reduces the buildable area, especially at the basement level. The presence of such old party walls places additional obligations on the developer and increases the risk of foundation excavation accidents. When a project occurs in an area occupied by old masonry buildings, the presence of a party wall along a lot line needs to be probed as early as possible.

THE LEGACY ADVANTAGE

Given the ubiquity of attached masonry structures in older U.S. cities, it is surprising that recent standards and model codes that cover existing buildings, i.e. the International Existing Building Code (IEBC) and the American Society of Civil Engineers ASCE 41-13, Seismic Evaluation and Retrofit Rehabilitation of Existing Buildings, do not devote specific prescriptions for party walls. Hopefully, this will be addressed in the future.▪

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Conclusions In their millennial existence, party walls have saved material and have performed STRUCTURE magazine

June 2017

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Typically, unreinforced masonry walls were supported by a continuous rubble stone wall that was approximately six inches wider than the wall above. Some weaker soils required this rubble wall be placed on an enlarged base – on a stepped masonry, stone, or concrete shelf – at least six inches wider than the wall above. Obviously, such enlargement encroaches on the intended location of the new foundation wall (Figure 6). If the toe needs to be removed, this must be indicated in the Party Wall Agreement. Sawing off the toe needs to be preceded by an underpinning solution that prevents the tendency of the foundation to rotate as a result of the eccentric application of the wall’s gravity force resultant. In a recent case, where the underpinning and removal of the toe were not properly engineered, a substantial increase of the existing building’s lean occurred. It required extensive shoring. Concerns about buildings “pounding against each other” during a seismic event have led to the introduction of “structural separation” requirements in engineering standards. Structural separation is not necessarily the opposite of party wall construction since, from a structural engineering point of view, a group of attached buildings forms a single structure. Following the 1995 introduction of the building separation requirement in NYCBC, it was observed that when a new building was built, separated by the code required gap from an existing unreinforced bearing masonry structure, the latter could start to lean and close the gap. To prevent this lean, the code now requires the structural separation along unreinforced masonry structures to be filled with a material that has a minimum compressive strength of 25 psi. Commonly, in new tall buildings, concrete shear walls are built along lot lines and function as firewalls and as an envelope enclosure. When adjoining existing party walls, these shear walls need to be placed further away from the lot line by a distance equal to the protruding wall dimension. Owners perceive this as a loss of rentable space.

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eismic assessment of earthquake-damaged buildings typically involves damage surveys and structural analyses. In this unique project, the evaluations also included the seismic testing of nine full-scale beam-column connections and two portions of shear wall slab connections, extracted from reinforced concrete buildings with minor damage. The concrete specimens had to be tested in less than a year. As the number of specimens and limited time exceeded the capacity of available testing facilities, a local warehouse was turned into a temporary testing facility. With the lack of a strong floor and strong wall, the project team designed a self-equilibrating steel test frame so that the reaction forces transferred to the warehouse slab were minimized. The innovative test frame accommodated all the different specimens in earthquake-type cyclic tests while simulating the boundary conditions of the specimens inside the buildings.

Innovative Test Frame The test frame, shown in Figure 1, consisted of two 20-foot-high steel towers separated by 20 feet and supported by 3-foot-thick steel-concrete (SC) composite footings resting on top of the warehouse slab. Horizontal steel diaphragms connected the two towers at the top and bottom. The top diaphragm was made removable to facilitate installation of the heavy specimens inside the frame. The bottom diaphragm was elevated from the floor so that the specimen reactions were transferred to the end towers without damaging the warehouse slab. Removable V-braces connected the bottom diaphragm to the two towers. The test frame was designed under strict deflection requirements so as not to influence the accuracy of the test results. The maximum lateral and vertical deflection were limited to 1⁄8 inch and 1⁄16 inch (downwards), respectively, while no uplift was allowed. The efficient structural system of the frame provided the necessary lateral and vertical stiffness. The two end towers were comprised of Universal Column (UC) sections UC310 and UC200, which were braced at different levels and stiffened by 3⁄8-inchthick steel plates. The middle tower columns, which

Figure 1. Self-reacting test frame.

Structural teSting Figure 2. Test frame deflected shape under maximum actuator load.

issues and advances related to structural testing

Figure 3. Contact pressure distribution under maximum actuator load.

supported the actuators, were welded-column (WC) sections WC400 and were perforated at various levels to enable adjustment of the actuator to match the different specimen heights. All tower columns were bolted to the SC footings, which consisted of steel boxes filled with 52 yards of high early-strength concrete, each, and reinforced with internal steel diaphragm plates. The SC footings sat on top of the warehouse slab and provided stability and limited vibration during testing. The biggest challenge during the frame design was the low bearing capacity of the warehouse slab, estimated at 14.7 pounds per square inch (psi). The frame was designed to internally equilibrate the applied and reaction forces to overcome this limitation, as shown schematically in Figure 2. The actuator, at its maximum capacity, imposes up to 3,688 kip-feet applied moment at the base of the frame. This moment generates a force couple (shown in red in Figure 2) that applies a high bearing pressure underneath one tower and large uplift forces on the other tower. On the other hand, the specimen reaction forces (in green in Figure 2) generate a reaction moment which equilibrates the applied moment. This equilibrium of forces requires the middle platform to behave in flexure, similar to a strap footing. The middle platform was raised 2 inches from the warehouse slab to function as required. As the reaction forces along the middle platform act in the opposite direction to the applied vertical forces, the bearing pressure applied to the warehouse slab was gradually reduced, as shown in Figure 3 for the case of maximum applied actuator force of 450 kips. The SC footings uniformly spread the applied pressure on the warehouse slab, eliminating any localized stress concentration underneath the towers and ensured contact to the slab at all times (i.e., no uplift). The SC footings were connected by the 20-inch-thick

STRUCTURE magazine

Temporary Testing Facility

Evaluating the Seismic Performance of Concrete Elements By Giulio Leon Flores, P.E., Hussien Abdel-Baky, Ph.D., P.E., and James C. Hays, P.E. Giulio Leon Flores (giulio.leon@ rizzoassoc.com), is a Engineering Manager with RIZZO Associates, a consulting company in Pittsburgh, PA. He was the lead engineer for the seismic tests. Hussien Abdel-Baky (habdelbaky@shockeyprecast.com), is a Senior Structural Engineer with the Shockey Precast Company in Winchester, VA. Previously, he worked for RIZZO Associates and served as structural engineer for the design of the test frame. James C. Hays (james.hays@ rizzoassoc.com), is a Chief Structural Design Engineer with RIZZO Associates. He supervised the design, fabrication, and erection of the test frame.

Figure 4. FE model of test frame with the largest specimen inside.

Figure 5. Assemblage of test frame in the warehouse.

Figure 6. Specimen layout under cyclic testing.

middle platform which consisted of steel beams partially encased by a 16-inch concrete slab that reduced deflection and vibration during the seismic testing. The middle platform beam along the same axis of the actuator column was drilled with anchor points to enable the adjustment of the struts supporting the specimen beam ends. Each anchor point location was designed for an allowable vertical load of 340 kips in tension or compression, and for a maximum vertical deflection of 1⁄16 inch. The finite element analysis of the frame was carried out using STAAD Pro software (Figure 4). Because of the warehouse limitations in height and access, the test frame was designed and fabricated as four separate modules which were assembled at the site (Figure 5): the bottom platform including foundation boxes, the two towers, and the top platform. This design allows future transportation and reassembly at a different location.

American Concrete Institute (ACI) Committee 374.1. The loading consisted of the application of incremental displacements at the top of the specimens, up to a drift of 3% or failure of the specimen. After each cyclic loading, the test was stopped to identify and measure cracks and also take photographic records. The test results consisted of force versus deformation plots at critical sections of one of the specimens. Figure 8a shows the moment versus rotation plot at the beam-tocolumn section of the specimen, shown at different cycles. Figure 8b shows the damaged suffered at the beam-column joint after the last cycle. During the tests, the deformations at critical locations of the test frame were also recorded, being the maximum displacement around 1⁄16 inch at the top of the frame, which was only 1% percent of the maximum applied displacement (+/-5 inches) and thus considered negligible.

Two 20-ton-capacity Franna cranes lifted and installed all the concrete specimens inside the test frame. Accurate synchronization of the crane movements was essential for proper lifting and movement of the heavy specimens, which were carefully braced before installation. Two 450-kip double-acting MTS hydraulic actuators applied the lateral forces at the top of the specimens, simulating the earthquake action (Figure 6). For the beam-column connections, the loading represented seismic forces acting on the specimens as if they were inside the buildings.

Secondary supports consisting of heavy steel caps at the bottom and top of the column, and double-pinned struts at the beam ends, forced the specimen to deflect as during a seismic event. In other words, a zero moment at the beam and column ends (pin connections) and free translation except at the column bottom, which is the reference point for relative displacement (Figure 6). The double-pinned struts consisted of 10- x 10-inch-square hollow sections with 80-millimeter-diameter pins attached at both ends. The bottom pin was bolted to a load cell used to measure the specimen reactions. Pretension rods attached the steel caps to the column, and the double-pinned struts to the beam ends. The axial forces in the pretension rods in the columns represented the gravity load carried by the column inside the building. The slab-wall specimens consisted of portions of a concrete shear wall with its tributary slab and were intended to test the shear transference at the slab-wall interface during a seismic event (i.e., shear friction). In this test configuration, the walls of the specimen were placed flat on the test frame floor while the slab stood upright (i.e., in the vertical position). A secondary frame consisting of UC310 sections with lateral bracing was designed to transfer the actuator force from the top to the lower part of the steel columns and then to the slab-wall interface via grout pads (Figure 7). The beam-column joint regions and slab-wall interfaces were externally instrumented with high accuracy string potentiometers to record beam and column flexural rotations and shear deformations. The specimens were subjected to quasi-static reversed cyclic loading in accordance with the

Figure 7. Slab-wall specimen in test frame.

Figure 8a. Moment vs. rotation results.

Figure 8b. Specimen damage after cyclic testing.

Testing Process

STRUCTURE magazine

June 2017

Conclusions The temporary testing facility, equipped with a self-reacting steel test frame, proved to be a successful alternative to performing seismic testing of large concrete elements on time and within budget. The innovative self-reacting frame, which can be transported and re-assembled at a different location, eliminates the need for strong-floor and strong-wall, thus dramatically reducing costs. The test results provided valuable information to evaluate the buildings’ seismic performances.▪

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Structural SyStemS discussion and advances related to structural and component systems

hat does stiffness have to do with strength? Well, engineers may remember learning about composite beam design, column design, or beams with compression reinforcing. Courses that included these topics introduced the “n” Factor. That is the factor that defines the ratio of the moduli of elasticity between materials. It is used in calculations to convert the stiffness of different materials to a common one. This was the foundation for understanding that stiffness plays a role in the distribution of forces in a section. However, material stiffness, the modulus of elasticity, and cross-sectional area are not the only components that influence the distribution of force flow. Geometric constraints, such as length of members, support conditions, and sectional properties all influence the force flow in a system. In fact, all of structural engineering is dictated one way or another by the ratio of stiffness between components in a structure. Force flow is defined here as the distribution of all internal systems forces, such as axial force, moment, shear, and torsion. This article discusses two systems in order to explore the influence of stiffness and strength: 1) structural static systems and 2) connections. Why is this important? Compatibility of deformations of structural components is the foundation of a safe structure. If there is no compatibility between components, they can easily be overstressed and can fail progressively as a result.

Stiffness Versus Strength By Paul A. Gossen, P.E., F.ASCE, and Keith M. MacBain, Ph.D., P.E.

Structural Systems Paul A. Gossen is Principal Emeritus at Geiger Engineers. He is a member of ASCE Committee 19 – Structural Application of Steel Cables for Buildings, as well as Committee 55 – Tensile Membrane Structures. Paul can be reached at pag@geigerengineers.com.

Often, members are connected to carry the same load, although sometimes it may not appear to be

Keith M. MacBain is a Principal of Geiger Engineers and can be reached at kmm@geigerengineers.com.

Figure 2a. Impact of top beam size increase.

20 June 2017

W BEAM

ANALYSIS MODEL

LINK 500

PLF

PICTURE 1

ANALYSIS MODEL

PICTURE 1

500

PLF

Figure 1. Analytical model – connected beams.

the case. In these instances, the load is distributed to the members in proportion to the ratio of their stiffnesses. A simple model can demonstrate the interactive effect of members with different stiffnesses (Figure 1). The beams were first sized as independent, unconnected beams based on a linear load of 500 pounds per linear foot (plf ). From the resultant moment, a W 10 x 15 was selected for the 20-foot long top beam and a W 10 x 30 was selected for the 30-foot long bottom beam. When connected, the top beam moment increased by 155% and the bottom beam moment decreased by 33.5%. This disparity is the result of the top beam being significantly stiffer, due to its shorter length, than the bottom beam. It essentially “robbed” the load from the bottom beam. The top beam’s size had to be increased to eliminate the overstress while the bottom beam size remained unchanged. It was necessary for the top beam to be increased to a W 10 x 30 to obtain the strength to carry the increased load that was shed from the bottom beam. Increasing its section results in a larger section modulus and, more so, a larger moment of inertia. Looking at the terms of both, the section modulus and the moment of inertia, one can see that the moment of inertia, and thus the stiffness, increased by a factor of 2.5 while the section modulus increased by 2.3, which is the opposite of what was desired. (Section Modulus S = bd2/6; Moment of Inertia I = bd3/12; b = width, d = depth) (See Figure 2a). Another solution would be to increase the stiffness of the bottom beam. In this scenario, the bottom beam is increased to a W 14 x 38 to obtain the stiffness that eliminates the shedding of load to the top beam, while the top beam size remained unchanged. Figure 2b shows the interaction of the two beams. Note that the second solution is more economical. In the first

Figure 2b. Impact of bottom beam size increase.

scenario, 300 pounds was added to the structure while the steel weight was increased by only 240 pounds in the second scenario. However, the beam depth increased from 10 inches to 14 inches. The 14-inch deep beam was chosen because the moment of inertia and stiffness increases significantly without a large increase in weight. If the depth could not be more than 10 inches, then the beam size would have had to become a W10 x 68. A third solution would have been to change the boundary condition of the lower beam, i.e. restrain the rotation at the support. So, where do these conditions occur in practice? They are more widespread than one may realize. A simple example is a wooden roof structure to form dormers, frequently used in residential buildings (Figure 3). The ridge beam is linked to the valley rafters by the sloped rafters. The roof sheathing is 2 x 6 tongue and groove decking which provides

very limited diaphragm action. Often members in these structures are designed based on the contributing load and spans without regard to the fact that they are interconnected. Thus, load sharing is ignored. However, if the structure is analyzed three-dimensionally, which reflects load sharing among the members, it is evident that the assumed load allocation is incorrect and that the ridge member, which was properly sized for the assumed “contributing” load, is overloaded. The behavior is similar to that demonstrated in the example above. Again, the reason is that the ridge member is stiffer compared to the valley rafter and attracts a greater share of the total load. Most engineers would increase the ridge member size, usually its depth, because this is more effective in resisting moments. However, what is needed is to modify its stiffness as well as its strength. As shown in the example above (Figures 2a and 2b), the increase in size, specifically the depth, increases the section modulus. At the same time, the moment of inertia is also increased by a larger factor. Increased stiffness attracts even more load, resulting in larger moments. Investigating the compatibility of stress and strain, an indicator of material stiffness and cross section, is essential when combining sections of different stress capacities and or stress/strain behavior. An example is a cable roof clad with prefabricated roof panels (Figure 4). The roof panels are

composed of joists that support metal roof deck. Edge members around the panels form borders that support the deck on the long sides and the joists on the short sides. The panels were prefabricated and lifted onto the cable net. The members along the cables are clamped to the cable to prevent the panel from sliding and to resist uplift from wind. The modulus of elasticity (E) of the cable is 20 x 106 kips per square inch (ksi), and its working strength is 90 ksi, with a factor of safety of 2.2. The edge member that is clamped to the cable is a structural section with a modulus of elasticity of 29 x 106 ksi and a working stress 21.7 ksi. If the cable and the edge member are rigidly bolted together, the strain compatibility must be checked. The cables are pre-tensioned to 50% of their allowable capacity before placing the panels. The differential strain between preload and maximum load for the cable is (0.5 x 90)/(20 x 10 6) = 2.25 x 10-6 and for the edge member it is 21.7/ (29 x 106) = .75x10-6. This means that the strain in the cable under full load is 3 times greater than the allowable strain in the edge member, which would cause failure of the edge member. The solution was to rigidly clamp the edge member of a panel to the cable at only one location and allow the other connections to slide. These additional connections were required to resist wind uplift. The detail for the sliding connection consisted of a split pipe with a neoprene liner between the surface of the cable on the inside face of the pipe. The split pipe was bolted on the cable using “U” bolts. There are a few structural systems where deformations increase their capacity. In catenary systems, elastic deformations amplify the sag and, thus, its load carrying capacity. Membrane and cable structures often incorporate catenaries.

RIDGE BEAM

SLOPED RAFTER VALLEY RAFTER

WOOD ROOF

PICTURE 3 Figure 3. An example of load sharing members in a wood roof.

STRUCTURE magazine

Figure 4. “Floating” roof panels in a cable roof.

June 2017

continued on next page

3/8"

3/4" THICK COVER PLATE

3/8" 1" TOP FLANGE OF W 30 x 132

END OFofCOVERPLATE FigureDETAIL 5. End detail a cover plate.

Figure 6. Wooden scarf joint connections.

Connections The importance of exploring compatibility in connections cannot be overstated. However, detailed free-body analyses of load transfers in connections are often ignored. An example of connection compatibility is the connection of a cover plate that reinforces a girder. The end connection of a cover plate to the flange of a girder is the critical area because the cover plate strain has to “catch up” with the strain in the girder flange. The load is transferred from the top flange to the cover plate through welds. The welds provide a shear connection between the two components. Thus, the shear capacity of the weld is critical for the integrity of the connection. Using a free body diagram and looking at the strain compatibility between the connected area of the cover plate and the flange, the weld, as well as the shape of the end of the cover plate, can be determined. The flange is under load and, thus, has deformations. The cover plate needs to adapt to this deformation for a safe load transfer. The limit of how much load can be transferred is defined by the capacity of the welds that connect the plate to the flange. The local stiffness of the cover plate section dictates how much load is attracted. In the design, one can tailor the shape of the end of the cover plate, thereby manipulating

the magnitude of load transfer to match the capacity of the welds or match the welds to the stiffness of the cover plate. Here is an example: A W30 x 132 beam has a moment of 665 kip-feet resulting in a bending stress (fb) of 21 ksi. A ¾-inch-thick cover plate is to be welded to the flange of the beam. The strain resulting from the stress in the beam flange must become the same in the cover plate. The plate is welded to the flange with a 3 ⁄8-inch fillet weld. The capacity of the weld is 5.6 kips per inch. For welds on both sides of the plate, the load that can be transferred is F = 2 x 5.8 = 11.6 kips per inch. The locally increased area of the cover plate at this location should not exceed A = F/fb; A = 11. 6/ 21 = .552 square inches. The stress of the flange is used here because the strain in the cover plate must match the one in the flange. The modulus of elasticity (E) is the same for the flange and the cover plate. Thus, the differential increase of the ¾-inch plate width over 1 inch is .552/ 0.75 = .736. Both sides of the plate edges are tapered in plan. Thus, the taper is 2/.736 or 1: 0.368 or 1 inch: 3⁄8-inch (Figure 5). The magnitude of the total load that is transferred depends on the width of the plate in this example. As more and more load is transferred to the plate, the stress level in the beam flange is reduced and, theoretically, the slope of the taper could be increased. A more practical solution would be to reduce the weld size.

Scarf splices in wood connections (Figure 6) use the same principal. By tapering the connected section against each other, a smooth load transfer is achieved without overstressing the glue line. They are usually used to extend the length of a member. Failure in these connections usually originates at the apex of the tapered joint. If the glue line fails at this location, the joint “peels” apart. The cause is an increase of stress in the remaining glue line. Another example of how the stiffness of a component in a connection affects its performance can be shown at a simply-bolted splice of a wide flange beam. The force in one beam is transferred to another beam by splice plates. Plates on each side of the web, on the outside of the flanges, and on the underside of the flanges are bolted together to form the connection (Figure 7). The connecting bolts are loaded in double shear. The plates on the outside of the flanges differ from those on the underside of the flange. The width of the outside plates can be the full width of the flanges. The plates on the underside of each flange are placed on each side of the web to clear the web and fillet; thus, the combined width of these plates is less than that of the top plate. The full capacity of a bolt in double shear can only be realized if both bolt shear planes are stressed to their maximum allowable shear stress. This can only happen when the strain of top flange plate is equal to that of the underside. A Consequently, the area of the top plate must equal the sum of areas of the “under-the-flange plates.” The total width of the underthe-flange plates is less than that 0.5A of the top plate due to the interference of the web and fillet. By increasing the thickness of the under-the-flange plates, the same area of that of the top plate can SECTION be obtained resulting in an equal strain in the connection plates.

SIDE VIEW

TYPICAL SPLICE Figure 7. SteelBEAM beam splice.

PICTURE 7

STRUCTURE magazine

June 2017

STRESS

GLASS & CERAMICS

The Final Word The stiffness distribution in a structural system affects its force flow. By manipulating its stiffness, the force and moment distribution can be rearranged. Components in structures are rarely isolated, even though it is often assumed that they are. Looking at the deformations and strain compatibility between components, one may get a better understanding of the force and moment distributions that may result in a

HIGH STRENTH WIRE

MILD STEEL PLASTIC RANGE

more economical and safer design. There are computer programs that analyze these relationships and more that can incorporate the non-linear behavior of materials. However, finite element analyses of connections often still lead to unrealistic results, because these analyses do not reflect the plasticity of the materials. The evaluation of strain compatibility in the design of pure elastic materials is essential. Without that, the possibility of sequential or progressive failure is high.▪

STRAIN

Figure 8. Stress-strain curves for different materials.

Materials

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All the above examples are based on materials with pure elastic behavior. Elastic behavior means that the stress/strain relationship is linear. Most designs assume that the materials behave linearly, which allows one to readily extrapolate the forces and stresses of a structure or connection. It allows the superposition of load effects. However, most materials exhibit non-linear behavior in higher stress ranges. Figure 8 shows the stressstrain curve for steel, high strength wires, and glass and ceramic materials. Note that mild steel in the plastic range (yield range) deforms but gains very little strength. This results in a loss of stiffness. Consequently, in the yield range, a member can shed load to other stiffer members in a structure that are less stressed and whose material have not yet reached the yield stress. The change in geometry in the system may cause instability in compression members that must be examined. Yielding of the steel, however, is limited by the breaking strength of the material as well as strain hardening. Strain hardening occurs with cycled stresses in the yield range and reduces the ductility of the material. Fillet welds would not work without yielding since strain compatibility is rarely observed between connected sections. Only in bridge work and when fatigue issues are a concern is strain compatibility addressed. The pure elastic behavior of a material (without any plasticity at any stress level) is brittle behavior. These materials are unforgiving to overloading, and their failure is without warning. It is for this reason that factors of safety in the order of 4 to 6 are applied in the design of glass and stone, and a factor of safety of 2.2 is applied in the design of high strength cables with a limited inelastic range at high stress. These are all in contrast to mild steel, with a sizable yield range and a factor of safety of only 1.6.

June 2017

Practical SolutionS solutions for the practicing structural engineer

n this article, a wood frame shear wall is analyzed and compared per American Wood Council’s 2015 Wood Frame Construction Manual (WFCM) and 2015 Special Design Provisions for Wind and Seismic (SDPWS). The difference between the segmented and perforated shear wall design approach and an overview of various hold-down design methods is discussed. Designers will likely find that the 2015 WFCM contains time-saving features for calculation of loads and design of shear walls and other building systems and components within the scope of the document.

2015 WFCM The 2015 WFCM is referenced in both the 2015 International Residential Code (IRC) and the 2015 International Building Code (IBC). The WFCM includes prescriptive and engineered design provisions for wood wall, floor, and roof systems and their connections. A range of structural elements

updated provisions to provide consistency with ASCE 7-10 and is also referenced in ASCE 7-16. Additional information concerning changes to 2015 SDPWS appeared in STRUCTURE magazine’s July 2015 issue. To facilitate the following design example, a free non-printable PDF version of the 2015 SDPWS and 2015 WFCM is available from the AWC website for those who do not already have a copy of the standards.

Wood Shear Wall Design Example The following design assumptions are used for development of a comparison of shear wall designs using the 2015 WFCM and 2015 SDPWS. Gypsum is assumed as interior shear wall sheathing, but the approach will show the difference when not including its capacity. Both the segmented shear wall (SSW) approach and perforated shear wall (PSW) approach are utilized and compared. Note this is not a comprehensive shear wall design. Issues such as deflection, wind uplift, base shear, and summing hold-downs from upper floors are not addressed.

Wood Shear Wall Design Examples for Wind Per 2015 WFCM and 2015 SDPWS By John “Buddy Showalter, P.E.

John “Buddy” Showalter is Vice President of Technology Transfer for the American Wood Council and serves as a member of the STRUCTURE magazine Editorial Board. He may be reached at bshowalter@awc.org.

is covered, including sawn lumber, structural glued laminated timber, wood structural sheathing, I-joists, and trusses. While the WFCM is geared primarily to one- and two-family dwellings, IBC 2309 also allows the WFCM to be used for small commercial applications assigned to Risk Categories I and II. As an example, a singlestory, slab-on-grade, light commercial structure with building length and width less than 80 feet (i.e., restaurants, office buildings, etc.) could be designed for lateral (wind or seismic) and gravity loads per the WFCM. ASCE 7-10 Load Provisions Tabulated engineered and prescriptive design provisions in WFCM Chapters 2 and 3, respectively, are based on the following loads from ASCE 7-10 Minimum Design Loads for Buildings and Other Structures: • 0 to 70 psf ground snow loads • 110 to 195 mph 700-year return period 3-second gust basic wind speeds • Seismic Design Categories A – D Additional information concerning changes to 2015 WFCM appeared in STRUCTURE magazine’s February 2015 issue.

2015 SDPWS The 2015 SDPWS contains provisions for design of wood members, fasteners, and assemblies to resist wind and seismic forces and is referenced in the 2015 IBC Chapter 23 on wood for the design of lateral force resisting systems. It contains many

24 June 2017

Summary of Design Assumptions • 130 mph (700-yr, 3-second gust) Exposure B • L=36 feet • W=30 feet • 5/12 roof pitch • Top plate to ridge height = 6.25 feet • 2-story • 8-foot wall height • 6.75-foot door height • 4-foot window height • Wood Structural Panel (WSP) Exterior Sheathing • Vary interior walls – with and without ½-inch gypsum Figure 1 provides an elevation view showing window and door openings. Only first-floor shear walls and hold-downs will be analyzed in this example. 2015 WFCM Prescriptive – Segmented Shear Wall The 2015 WFCM Table 3.17A gives length requirements for an SSW resisting wind loads. For a 30-foot building wall length, W, the interpolated value is 12.3 feet. Table 3.17A, Footnote 4, provides required sheathing length adjustments based on wall heights and top plate-to-ridge heights other than what is assumed in the table – which is 10 feet for each. For this example, the adjustment is 0.68. Therefore, an SSW would require 8.4 feet of full-height wall segments. Figure 1

Segmented Shear Wall

Perforated Shear Wall Figure 1. Elevation view of the structure used in a wind load design example comparing shear wall design procedures.

shows 4 feet of shear wall at each corner and 2.5 feet between the windows for a total of 13 feet, which is sufficient. Hold-downs are required at the ends of each segment and will be designed later (Figure 2). WFCM Table 3.17D facilitates variation of exterior and interior sheathing materials, nail diameter and spacing, panel thickness, and stud spacing. WFCM shear wall baseline assumptions are 7⁄16 wood structural panels on studs at 16 inches on-center, 8d common nails, 6-inch panel edge nail spacing, and 12-inch panel field nail spacing (6 and 12). Shear wall tables for wind also assume contribution of half-inch unblocked interior gypsum. Note the allowable stress design (ASD) unit shear capacity of 436 plf for wind and a maximum shear-wall-segment aspect ratio of 3.5:1. All shear wall capacities in Table 3.17D are derived from the 2015 SDPWS. Table 3.17D, Footnote 2, requires blocking of gypsum wallboard edges where the aspect ratio exceeds 1.5:1. SDPWS does not contain aspect ratio limits for the case where exterior and interior sheathing materials are combined for wind resistance. SDPWS does state that unit shears can be combined. Prior practice within WFCM for aspect ratios has been to limit use of the combined materials to the higher aspect ratio material. For example, WSP alone has a maximum aspect ratio of 3.5:1. Blocked gypsum wallboard has a maximum aspect ratio of 2:1. WFCM has permitted the use of combined resistance resulting from WSP exterior sheathing and blocked GWB interior sheathing on walls with an aspect ratio up to 3.5:1. What if interior gypsum capacity is excluded from wind design? There may be cases, such as unfinished garages, where there is no interior gypsum. Note also, if contractors don’t install gypsum with assumed nail spacings, it doesn’t provide additional

Figure 2. Comparison of shear wall requirements for a segmented vs. a perforated shear wall using 2015 WFCM prescriptive design provisions and assuming contribution of gypsum capacity.

shear capacity for which it is designed. The capacity shown in Table 3.17D is 336 plf (gypsum is assumed to have 100 plf capacity for wind), and an adjustment factor of 1.3 is tabulated which can be used to adjust sheathing length requirements that were calculated from Table 3.17A earlier. That would then require 10.9 feet of sheathing length rather than 8.4 feet. In this particular example, if interior gypsum is excluded, the 13 feet of full-height segments determined earlier is sufficient. As noted, hold-downs are required at the ends of each segment (Figure 2). 2015 WFCM Prescriptive – Perforated Shear Wall For the PSW approach, the entire wall is sheathed on one or both sides with wood structural panels. For wind design, interior gypsum can also be used additively with exterior wood structural panels. The contribution of sheathing above and below window openings, and above the door opening, can also be included. Nail spacing requirements for WSPs may be decreased (e.g. from 6 and 12 to 4 and 12). By increasing wall capacity, hold-downs can be eliminated around window and door openings. This is a major benefit of the PSW method. WFCM Table 3.17E is used to determine the PSW full height sheathing adjustment. There is a 6.8-foot door opening in the middle of the wall, which will be used as the maximum, unrestrained opening height. The full-height sheathing length required for the SSW was 8.4 feet when including gypsum. The PSW length adjustment is based on the tabulated length of SSW required. So, 8.4 feet divided by the full wall length is 23% of full-height sheathing. Interpolation gives a factor of 1.86. Multiplying by the length required for the segmented method results in 15.6 feet.

STRUCTURE magazine

June 2017

Without interior gypsum, the segmented method required 10.9 feet which is 30% of full-height wall sheathing. So, the 1.72 factor from WFCM Table 3.17E results in 18.7 feet required for the PSW. Either one of these results is sufficient because there are 21 feet of full height wall sheathing. The 21 feet of full-height sheathing is based on wall portions that meet maximum aspect ratio limits. As noted, a major benefit of the PSW method is that, for examples such as this, hold-downs are only required at the corners. By simply accounting for the benefit of a wall fully-sheathed with WSPs, which is fairly common in most parts of the country, holddowns can typically be eliminated around window and door openings (Figure 2). 2015 WFCM Prescriptive – Hold-Down Capacities In the WFCM, overturning loads are differentiated from uplift loads. Overturning moments result from lateral loads which are resisted by shear walls. Uplift forces arise solely from wind uplift on the roof and are transferred directly into walls supporting roof framing. A conservative assumption of WFCM tabulated hold-down capacities is that ASD unit shear capacities for the reference shear wall are multiplied by wall height to determine maximum hold-down capacity. Therefore, the same hold-down capacity can be calculated for both the SSW and PSW. Calculating the wind hold-down capacity in WFCM Table 3.17F is based on both wood structural panels and gypsum, resulting in 3,488 pounds for this example. If gypsum is excluded, then the capacity is lower. Table 3.17F, Footnote 1, states that the tabulated hold-down capacity is divided by the sheathing type adjustment factor from Table 3.17D, which is 1.3 as determined earlier, so a 2,683pound hold-down is required if excluding

Reference SDPWS Capacities and Adjustments V = 2,985 lbs v = 436 plf (w/ blocked gypsum) v = 336 plf (w/o gypsum) %FHS = Li / Ltot Li = 16’ + 2[2(2.5)/8]2.5’ = 19.1’ Ltot = 36’ %FHS = 19.1’ / 36’ = 53% Interpolated Co Factor = 0.59

Perforated Shear Wall

436(0.59) = 257 plf 2,985/257 = 11.6’ < 21’ (w/ blocked gypsum) 336(0.59) = 198 plf 2,985/198 = 15.1’ (w/o gypsum)

Segmented Shear Wall Figure 3. Comparison of shear wall requirements for a segmented vs. a perforated shear wall using 2015 WFCM engineered design provisions and assuming contribution of gypsum capacity.

gypsum. Note also that hold-down capacities are tabulated per story. Required hold-down capacities need to be summed from the story above, but are not shown in this example for simplicity. Another conservative assumption in WFCM Prescriptive Design provisions (Chapter 3) is that design dead load is only used to offset uplift loads and not overturning loads. However, WFCM Engineered Design provisions (Chapter 2) allow for up to 60% of design dead load to offset overturning. Of course, engineering judgment is required to determine what portion of design dead load is tributary to the holddown, which is a major reason for the conservative approach. 2015 WFCM Engineered – Segmented Shear Wall Engineered requirements in WFCM Chapter 2 allow calculation of loads that are assumed in the prescriptive requirements of WFCM Chapter 3. WFCM Table 2.5B shows lateral loads on the roof and floor diaphragm. With a 5/12 roof pitch, a roof span of 30 feet and loads parallel to the ridge, interpolate 94 plf.

Note: Li per SDPWS 4.3.4.3 adjustment = 2bs/h

Figure 4. Summary of perforated shear wall calculations per 2015 SDPWS Table 4.3.3.5.

Floor diaphragm load is 105 plf (after a wall height adjustment per Footnote 2). Add those up for 199 plf, multiply by building width, and divide by two because half the load goes to each shear wall. The result is 2,985 pounds at the top of the first-floor shear wall. The 2015 WFCM references 2015 SDPWS for shear wall capacities. However, as discussed earlier, WFCM Table 3.17D tabulates 436 plf and 336 plf shear wall capacities for this example, with and without gypsum, respectively. Therefore, length requirements are: • 2,985/436 = 6.8 feet (with blocked gypsum) • 2,985/336 = 8.9 feet (without gypsum) Based on these results, the 4-foot segments at each building corner are sufficient if blocked gypsum is included in the shear wall capacity (Figure 3). This also shows that the WFCM engineering provisions provide more efficiency in the design process than the prescriptive design provisions. 2015 WFCM Engineered – Perforated Shear Wall Figure 4 summarizes PSW calculations per SDPWS Table 4.3.3.5. Figure 3 shows the

Shear Capacity Adjustment Factor h = 8’ Capacity Adjustment Factor Shear = 8’ 16’ + 2[2(2.5)/8]2.5’ = 19.1’ L hi =

36’+ 2[2(2.5)/8]2.5’ = 19.1’ 16’ Ltot i == A + (5’)(6.67’) = 73.4 ft2 Ltot =4(4’)(2.5’) 36’ o = rAo==0.68 4(4’)(2.5’) + (5’)(6.67’) = 73.4 ft2 C 0.77 (based on total sheathed area) r o==0.68

comparison of shear wall requirements for SSW versus PSW assuming gypsum contribution. 2015 WFCM Engineered – Hold-Down Capacities WFCM section 2.2.4 allows for offsetting hold-down capacity with up to 60% of design dead load. The same approach for determining hold-down capacities based on ASD unit shear wall capacity is used for this example. Therefore, hold-down capacities are as shown earlier for the prescriptive design approach. 2015 SDPWS – Segmented Shear Wall SSW design per the SDPWS is identical to what was shown earlier under the WFCM Engineered approach. As noted, the WFCM references the SDPWS for shear wall capacities. 2015 SDPWS – Perforated Shear Wall Figure 5 summarizes PSW calculations per SDPWS Equations 4.3-5 and 4.3-6. These equations provide more accuracy by allowing the total sheathed area to be included in capacity calculations. Note that SDPWS

V = 2,985 lbs (blocked gyp)

2,985 lbs (blocked gyp) hV== V8' =≤1 2,985Vlbs (blocked gyp) gyp) =lbs 2,985 lbs (blocked V = 2,985 (blocked gyp) h= 8' ≤1 h= 8' h = 8' 0.77 Co = 8' == 0.77 C h Co = 0.77 = 0.77 Li o=Co16’ + 2[2(2.5)/8]2.5’ = 0.77 Co16’ Li = 16’ + 2[2(2.5)/8]2.5’ Li = + 2[2(2.5)/8]2.5’ = 16’ + 2[2(2.5)/8]2.5’ Li =LLi19.1' 19.1' L = = 16’ + 2[2(2.5)/8]2.5’ Li = = 19.1' L i 19.1'

i T = 1,624 lbs T =L1,624 lbs =1,624 19.1' T= lbs T i1,624 = lbs

T= Comparison: SDPWS/WFCM Engineered (tabulated) Co = 0.59 C on total sheathed area) o = 0.77 (based

1,624 V =lbs 1,840 lbs (w/o gyp) 4.3.4.3…In the design of perforated shear walls, the length of each perforated shear wall segment with an aspect ratio greater than 2:1 = 1,840 lbs (w/o gyp) = 0.59 Comparison: SDPWS/WFCM Engineered (tabulated) CV /h for the purposes determining L and ∑ Lof each shallthe be multiplied 4.3.4.3…In design by of 2b perforated shearofwalls, the length T = 1,001 V = 1,840 lbs (w/olbs gyp) V = 1,840 lbs (w/o gyp) o s i 4.3.4.3…Inthe thedesign designofofperforated perforated 4.3.4.3…In shear walls, walls, the thelength lengthofofeach each

Note: Li per SDPWS 4.3.4.3 adjustment = 2bs/h Note: Li per SDPWS 4.3.4.3 adjustment = 2bs/h

perforated shear wall segment with an aspect ratio greater than 2:1

perforatedshear shearwall wallsegment segment with with an aspect aspect ratio perforated ratiogreater greaterthan than2:1 2:1 forpurposes the purposes of determining L and ∑ L shall be multiplied by/h/h2b TT==T lbs for/h the purposes of LLand shall multiplied by2b 2b = lbs shall bebemultiplied by for the of determining determining and∑ L∑ L 1,001 lbs V1,001 =1,001 1,840 lbs (w/o gyp) 4.3.4.3…In the design of perforated shear walls, the length of each ss

i i

i ii

perforated shear wall segment with an aspect ratio greater than 2:1

purposes of determining be multiplied by 2bs/h for the T= lbs Figure 5. Calculation summary of PSW shear capacity adjustment factor per 1,001 Figure 6. Summary of perforatedshall shear wall calculations of hold-down capacity Li and ∑ Li 2015 SDPWS Equations 4.3-5 and 4.3-6. per 2015 SDPWS.

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June 2017

Table 4.3.3.5 and WFCM Table 3.17E both use maximum opening height and, for example, do not account for sheathing below window openings. Adjusting the shear wall capacity of 436 plf (with blocked gypsum) and 336 plf (without gypsum) by Co = 0.77 results in shear wall lengths of 8.9 feet and 11.6 feet, respectively. 2015 SDPWS – Hold-Down Capacities There is a difference in how hold-down capacities are calculated in SDPWS versus WFCM. Per SDPWS Equation 4.3-7 for SSW, hold-down capacity is based on induced unit shear. The 2,985-pound load and 8-foot shear wall length result in 347 plf. The shear wall height of 8 feet results in a hold-down capacity of 2,985 pounds. Excluding gypsum, a hold-down of 1,840 pounds is calculated. For the PSW, SDPWS Equation 4.3-8 can be used, which accounts for the shear capacity adjustment factor determined earlier and the 2bs/h adjustment per SDPWS 4.3.4.3. Figure 6 summarizes PSW hold-down calculations.

Summary and Conclusion Table 1 provides a summary comparison of various methods used for this wood shear

AWC Standard

Segmented

Perforated

Hold-Downs, lbs

2015 WFCM Prescriptive

8.4’ (10.9’)

15.6’ (18.7’)

3,488 (2,683)

2015 WFCM Engineered

6.8’ (8.9’)

11.6’ (15.1’)

3,488 (2,688)

2015 SDPWS

2,985 (1,840) [SSW] 1,624 (1,001) [PSW] (Parenthetical values assume NO interior gypsum capacity) 6.8’ (8.9’)

8.9’ (11.6’)

Table 1. 2015 WFCM/SDPWS Shear Wall Length Comparison – 1st of 2-story; 30-foot roof span with 5/12 pitch; 130 mph Exposure B wind load.

wall design example. WFCM prescriptive design provisions have an engineering basis in SDPWS and ASCE 7, even though simplifying assumptions are made which lead to more conservative results. WFCM engineered and SDPWS results are nearly identical, which is expected, except for the PSW design. The difference in the PSW results is due to SDPWS equations being used to calculate the shear capacity adjustment factor, rather than using tabulated values. The segmented approach results in less shear wall length in all cases; however, hold-downs are only required in the corners for the PSW. By simply accounting for the benefit of a wall fully-sheathed with WSPs, which is fairly common in most construction today, hold-downs can typically be

eliminated around window and door openings. Not surprisingly, where the most efficiency is gained is with the pure SDPWS engineering approach. Shorter shear walls and smaller holddowns result. However, designers might find the WFCM tables to be a time-saving feature, especially for calculating ASCE 7 wind loads.▪ This article is based on an NCSEA Diamond Reviewed program titled DES413-2 – Wind Shear Wall Design Examples per 2015 WFCM and 2015 SDPWS available for free on the AWC website. Earn continuing education credits or find more details about this shear wall example at www.awc.org.

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Structural FailureS investigating failures, along with their consequences and resolutions Failure caused by compacted snow pile, Secaucus, NJ.

tructural engineering always has been a conservative profession. This is particularly evident in the structural building codes and standards. These regulations are intended to protect the public from deficiencies in design and construction. For the most part, they are effective. Why then do so many parking garages collapse? In 2015, at least six parking garages partially collapsed in the United States. More collapsed in 2016. Some failures might be due to deficient design or construction, or to inadequate maintenance, but most are due to excessive loading. For decades, the codes and standards have stipulated that parking garages be designed for a minimum live load of 50 psf. This is more than adequate for sedans, SUVs, and half-ton pickup trucks. The average vehicle in the United States today weighs 4,009 pounds and can safely be parked in any code-compliant garage. With the margin of safety inherent in the codes and standards, failure becomes imminent only when actual live loads approach twice the minimum. Some parking garages post maximum vehicle weights at their entrances. More commonly, the entrances limit vehicle height (thus, vehicle size/ weight) with clearance bars typically set at 7 feet or less. Here are four examples where parking garages collapsed due to excessive loading.

Saturday, 01/24/15, Secaucus, NJ A 3-level garage with 600 parking spaces serves the Harmon Plaza office tower, the Clarion Empire Meadowlands Hotel, and the Osprey Cove apartments. A Bobcat utility vehicle was plowing the top deck of the garage following a snowfall of 4 inches or less. The weight of the Bobcat was not a problem, nor was the weight of the accumulated snow. However, compacted snow weighs about 20 pcf, and 30 pcf or more if wet. The snow had been pushed into a compacted pile. At 7:00 am, with the pile more than 4 feet high, the top deck of the garage collapsed under the combined weight of the snow pile and the Bobcat. The resulting opening swallowed the Bobcat and one vehicle parked below was crushed. No one was injured except the Bobcat driver, who had a mild concussion. On weekdays, the garage would have been filled with vehicles and people.

Failure of Imagination

By Stan Caldwell, P.E., SECB

Friday, 05/01/15, Washington, D.C. A 3-level garage serves the iconic Watergate mixed-use complex. Now more than 50 years old, the complex was undergoing a comprehensive restoration. A landscaping contractor placed soil and debris on the top deck of the parking

Garage collapse, Washington, D.C.

STRUCTURE magazine

Stan R. Caldwell (www.StanCaldwellPE.com) is a consulting structural engineer in Plano, Texas. He can be reached at StanCaldwellPE@gmail.com.

Pancake failure, Dallas, TX.

garage. Soil weighs about 75 pcf if dry, and up to 125 pcf if wet. Thus, only one foot of soil might cause failure. At 10:00 am, the top deck of the garage collapsed beneath the soil and debris. Everything on the two levels below was crushed in a pancake failure. Two people were injured, one critically, and about 35 vehicles were destroyed. Friday, 10/23/15, Dallas, TX A 7-level garage with 800 parking spaces serves the upscale Renaissance on Turtle Creek condominiums. The pool and amenities deck at the top of the garage were being renovated, and the contractor piled soil and miscellaneous debris at the toe of a ramp near one corner of the top deck of the garage. A video taken just a few hours before the collapse shows that the pile was about 3 feet high and was covered with plastic sheeting. During a rainstorm, water flowed down the ramp and partially saturated the pile. At 5:00 pm, the top deck of the garage collapsed beneath the pile. Everything on the six levels below was crushed in a pancake failure. Amazingly, no one was injured. Numerous vehicles were destroyed, and about 250 vehicles and their contents remained inaccessible to their owners for 58 days. The garage remained closed for several months until reconstruction was completed. Friday, 04/22/16, Houston, TX An underground garage with an at-grade top deck serves adjacent office buildings on Town & Country Boulevard in Houston. The garage had been flooded during a period of record rainfall. A tanker truck was summoned to the site to pump out the water. The entrance to the top deck was blocked by a clearance bar proclaiming “MAXIMUM WEIGHT 4,000 LBS”, so the truck driver backed his tanker over the curb and onto the deck a few feet to the right of the entrance. The tanker has a capacity of 5,800 gallons. That amounts to 48,400 pounds of water, not including the weight of the truck. As the tanker filled with water, the top deck

eventually failed under the weight of the rear wheels. Fortunately, no one was injured.

Failure of Imagination In the wake of the terrorism that jolted the United States on the morning of September 11, 2001, some smart person declared that the attacks succeeded in part because of a “failure of imagination” by multiple American security agencies. No one imagined that a zealous group of young men would simultaneously hijack four airliners full of fuel and fly them into predetermined targets using nothing more than box cutters, which were entirely legal to carry on flights at that time. The parking garage collapses described in this article are not directly related to any known design, construction, or maintenance deficiencies. Nevertheless, with hindsight, all were preventable. In each instance, it could be argued that there was a failure of imagination. That is, a failure by parking garage designers and owners to imagine the utter lack of common sense among some of those working in and around the garages. In Secaucus, the garage probably did not need to be plowed. The light snow likely would have melted within a day or two. If snow removal was necessary, it should have been done in a manner that avoided creating compacted piles. In Washington and Dallas, the owners and the landscaping contractors should have

Deck failure, Houston, TX.

STRUCTURE magazine

June 2017

contacted structural engineers for approval before the placement of any soil or debris, even temporarily, on the top decks of the garages. In Houston, an above-grade security fence or rail at the perimeter of the underground garage footprint would have kept heavy trucks off of the top deck. Imagination is a difficult thing. The building codes and standards will never adequately address that which is unforeseen. This shifts the burden to structural engineers, and simply designing for large imaginary loads is clearly not the answer. Perhaps structural engineers should consider changing their approach to future parking garage projects. In conclusion, the following describes two imaginative suggestions. First, to reduce the likelihood of overload, structural engineers should seek “buy-in” from the owners of parking garages regarding loading limitations. Most manufactured products come with a manual that clearly defines intended use and a statement that the manufacturer is not liable when their product is used in other ways. Parking garage designers should consider a similar approach. There is a precedent for this. In areas with expansive clay soils, prudent structural engineers require building owners to sign off on the acceptable differential movement of slabs on grade. A structural engineer could write a project-specific loading statement for a parking garage and require the owner to sign a brief document indicating that he or she has read, understands, and accepts the loading limitations for that garage. Second, to limit the damage of a collapse, structural engineers should design parking garages to avoid progressive collapse or pancake failure. This is particularly important at the level immediately below the top deck. While not easily or inexpensively accomplished, this is doable and would substantially limit any damage caused by an overload at the top level.▪

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A Unique Opportunity for a Conventional Tall Building Frame By Mark Sarkisian, S.E., Neville Mathias, S.E., and David Shook, P.E.

he Sichuan Airlines Center (SAC) Tower occupies a prominent site in the city center of Chengdu and has become a glittering symbol of the city’s urban transformation. The distinctive 650-foot tall (200 meter) tower is a recognizable city landmark, occupying a pivotal location on the Chengdu skyline. The Tower form is a natural reconciliation of geometry, structure, and space made possible through a systematic approach to its construction. The tower’s design is focused on the creation of flexible, high-quality office spaces and a thoughtful engagement of the public realm, inviting the life of the street into the site. The unique form of the tower folds open vertically, responding to the local climate, capturing light and views for occupants. As the tower rises, its shape is organically transformed in response to the changing relationship of the interior spaces to their elevation; each floor slab changes slightly in shape from the one below at a consistent rate. The perimeter columns follow the building’s changing shape, shifting slightly from one level to the next. To achieve this unique building form, surfaces of the tower skin gently twist as the building rises. This subtle movement is achieved by precisely bending each glazing panel onto the building using adjustable exterior wall anchors. Overall, the result is a single, continuous form that accommodates both the required footprint and setbacks, while extending the southeast corner of the building as a large gentle arc, visually unifying the building’s relationship with the corner of the site and the surrounding civic space. continued on next page

The distinctive 200-meter-tall tower form is a recognizable city landmark, occupying a pivotal location on the Chengdu skyline.

STRUCTURE magazine

June 2017

As the tower rises, its shape is organically transformed in response to the changing relationship of the interior spaces to their elevation.

Sustainability The Sichuan Airlines Centre has achieved LEED® Gold Pre-certification for its commitment to energy efficiency, water conservation, increased ventilation, and improved indoor air quality. The project incorporates a number of sustainability measures which are integral to the building’s overall design intent. Energy use is extensively metered for actual energy consumption monitoring and verification efforts. The project additionally integrates a water efficient landscaping concept where storm water and air conditioning condensation are collected and treated for landscape and irrigation uses. SAC has achieved the highest standard of stormwater design within the project as the landscape irrigation does not utilize any potable water on site. Similar to the design of the structure itself, the development of the site maximizes open space and access to nearby public spaces. The site’s connectivity is positioned near a number of existing public transit opportunities including metro train and bus networks; it is situated just one block from a subway interchange station. This contributes to the building’s high Walk Score rating of 94 – where daily errands do not require a car. A new, extensive public transit system has also been proposed for the Central Chengdu region, which will further increase alternative transportation options for building occupants.

interior partition walls to meet the exterior wall. Buildings with complex and varying perimeter geometries often lose this organization of the column and exterior wall mullions. Thus, when an interior partition meets the exterior wall, it often ends up being located between mullions and columns. This greatly diminishes the usability and quality of the interior spaces. For the SAC tower, this condition was minimized by shifting the columns slightly on each floor and following the mullions. Thus, an innovative and integrated building form was achieved without compromising the quality of the interior spaces. The shift in column location from floor to floor was small enough that conventional reinforcement detailing could be used and significant eccentricities were avoided. Since the frame column spacing was tighter towards the top of the building, it could be designed more efficiently and could have higher participation in seismic events, alleviating some of the demands on the central core. At the corners of the floor plans, much longer spans, up to 42 feet (13 meters), were introduced to provide expansive, uninterrupted views.

Safety in Seismic Zones The city of Chengdu is the capital of Sichuan province in China and was tragically struck by a magnitude 8.0 earthquake in 2008, only three years before the start of the SAC Tower design. With nearly 70,000 deaths, 15,000 missing, and 375,000 injuries, safety was of paramount importance to both the developer and the city of Chengdu. In recent decades, the Chinese building codes have significantly evolved and have been shown to yield safe buildings in seismic zones. The issues encountered in the Sichuan earthquake of 2008 were due to much older buildings which pre-dated many of the current Chinese code provisions pertaining to seismic safety and poor construction. The seismic force resisting system of the SAC tower is a dual system composed of a centrally reinforced concrete floor supplemented by a perimeter reinforced concrete moment frame. The combined behavior of the two systems results in a building period of 5.3 seconds with satisfactory inter-story drift behavior.

Form and Structure Creating Quality Spaces

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Building forms which are highly responsive to energy performance, quality views, and aesthetics often create complex geometric interior spaces. This frequently results in poor office layouts and irrational relationships of building structure and exterior wall. Even in cases where the structural geometry is responsive to the exterior wall, the quality of the interior office space is often left lacking. The exterior wall of SAC is highly respectful of solar exposure and expansive views. A traditional moment frame would have approximately 30-foot (9-meter) bays with columns extruded vertically through the space. Since the exterior wall changes dramatically from the base of the building to the roof, the column relationship to the exterior wall would result in varying and often very poor office space conditions, dramatically diminishing views and adversely affecting energy performance by blocking too much sun in the cold winters prevalent in Chengdu. A simple, but novel change to traditional frame geometry was incorporated which gradually reduced the spacing of columns from 30 feet at the building’s base to 20 feet (6 meters) at the top, and shifted in their plan location slightly at each floor. Buildings with extruded rectangular geometries typically have the columns and exterior wall mullions aligned. This creates a natural place for STRUCTURE magazine

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June 2017

a general review of all aspects of the lateral force resisting system. Under the provisions of the Chinese building code, in addition to other special design checks, a nonlinear time-history analysis was conducted by China Southwest Architectural Design and Research Institute and presented to the EPR panel. Although different than the performance-based design process established in select cities of the West Coast of the United States, the EPR process serves as a highly valuable component of the Chinese building code review process in producing buildings of high seismic safety.

Conclusion The Sichuan Airlines Center tower is an example of transformative urban architecture, quality office space, integrated design, and seismic safety. Its construction in a rapidly evolving Chengdu city skyline will provide a quality precedent for future construction in Sichuan Province.▪ The site unifies the building’s relationship with the corner of the site and the surrounding civic space.

The Chinese regulations require buildings that exceed code provisions to undergo “expert panel reviews” (EPR). The level of review depends on a building’s code-exceeding characteristics such as structural height, the uniqueness of the lateral force resisting system, building complexity, inherent torsion, etc. Due particularly to the height of the building, which exceeded code limits for the structure type, a regional level EPR was required, as opposed to a more stringent national level EPR. The regional EPR process is

Mark Sarkisian, S.E., is the Structural and Seismic Engineering Partner of the San Francisco office of SOM. He can be reached at mark.sarkisian@som.com. Neville Mathias, S.E., is an Associate Director at the San Francisco office of SOM and can be reached at neville.mathias@som.com. David Shook, P.E., is an Associate Director and Structural Engineer at the San Francisco office of SOM. He can be reached at david.shook@som.com.

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New Clearwater Opal Sands Resort on South Clearwater Beach, Florida.

A Challenging Wind and Flood Design

pal Sands Resort is a new 17-story plus, 380,000-squarefoot premier hotel located directly on the Gulf of Mexico in Clearwater Beach, Florida. The recipient of the 2016 ENR Southeast Regional Best Project Award, the Opal Sands Resort was born from the vision of the owner/developer, Ocean Properties, LTD (OPL). The owner desired a new flagship hotel to add to their portfolio of international resort hotels. Soon enough, that vision became a reality because of the innovative work by project architect, Nichols Brosch Wurst Wolfe & Associates, Inc. of Coral Gables, Florida, and their team of consulting engineers, including the structural engineers, McCarthy and Associates, a Division of Pennoni, and the general contractor, Moss & Associates LLC.

The Challenges From the beginning, the apparent structural design challenges were numerous: 1) Design a 17-story curvilinear building, with a coastal exposure, to resist Florida’s high hurricane wind loads. 2) Design a building located in multiple flood zones including two different FEMA V-zones, the Pinellas Gulf Beaches Coastal Construction Zone, and the Florida Department of Environmental Protection (DEP) Coastal Construction Control Line (CCCL). STRUCTURE magazine

3) Design a heavy concrete building on a site overlaid with loose beach sands. 4) Eliminate a critical transfer beam over the lobby during construction and after many of the upper floors had been constructed.

The Building The Opal Sands Resort consists of 12 floors of hotel rooms, each of which has a magnificent view of the Gulf of Mexico. The hotel rooms sit above a spacious lobby, restaurant, numerous meeting room floors, and two levels of parking. At the lobby level, on the west waterfront side of the main tower, is a plaza deck with swimming pools, volleyball courts, and tiki hut bars. A large portion of that deck is cantilevered out over the water. A sand ramp was constructed, in cooperation with the City of Clearwater, to allow for visitors to easily access the beach from the deck above. On the east side of the main tower is another expansive space that contains a large ballroom, meeting rooms, and escalators from the main entrance up to the lobby level. Valet parking for guests is required because the hotel takes up the entire site and access to the parking levels is only available from a ramp on the street side and two large vehicle elevators. Considering the sheer size of the building, along with its ideal location and innovative design, it comes as no surprise that the hotel was an instant success upon opening in February 2016.

June 2017

The General Design Approach The entire building superstructure and foundations were modeled in 3D using RAM Structural Systems. The output from that model was then sent directly to the 3D Revit model and refined to make the structural drawings more readable and consistent with industry standards. Due to the unusual shape of the floor plans and column bay spacing provided by the architects, unbonded mono strand posttensioned concrete slabs supported by concrete columns and shear walls was an obvious choice for the superstructure. Post-tension design also helped to minimize floor heights and has been a popular system for multi-story residential buildings in the state of Florida for a long time. The delegated engineer for the post-tension system was Structural Technologies/VSL. Although the architects were very accommodating in locating columns to best support the structure, there were instances where it was necessary to introduce transfer beams to support discontinued columns from above. For example, a large transfer beam (3 feet x 4 feet), reinforced with a combined mild steel and bundled post-tension cables, was required over the lobby where the engineers were directed to provide a column-free space. A very large cooling tower is located on the roof of the main tower and surrounded by hot dipped galvanized steel framed walls that are elliptical in plan. The walls are 19 feet 6 inches high and self-braced against wind tunnel test-wind pressures, reaching 125 pounds per square foot (psf ), with a grid system of steel beams at the top of the walls. The steel framing is designed and detailed to be removable for

a load test was conducted on a sample shaft to confirm the foundation design recommendations. A pilot hole was then drilled at each shaft location to confirm there was no variance in the subgrade.

Wind Design

The governing code for the design of the Opal Sands Resort was the 2010 Florida Building Code (FBC) which, by reference, incorporated much of the ASCE 7-10 Minimum Design Loads for Buildings and Other Structures for wind design. According to a Pinellas County local technical amendment, the minimum design wind velocity for this site is 145 mph with an ultimate 3-second gust, Risk Category II, and Exposure Category D as defined by the FBC. In addition to wind forces, the exterior glazing must also meet wind-borne debris or missile impact criteria and be certified by the manufacturer with a Florida Product Approval Number or Miami/Dade Notice of Acceptance. Due to the curved shape of the main tower and in anticipation of cost savings, OPL agreed to McCarthy’s recommendations to complete a wind tunnel test. CPP, Inc. of Fort Collins, CO, was retained to build a model of South Clearwater Beach and the proposed Opal Sands building, and then subject that model to hurricane force winds in a wind tunnel. The resulting pressures were less than what would have been required by code and were used by McCarthy in the design of the building for both the main wind force resisting system and components and cladding. Additionally, the wind tunnel testing provided the architects with accurate feedback on locations around the base of the main tower where wind accelerations are likely to affect pedestrians adversely. Screen walls, landscaping, and other types of buffers were then incorporated into the building design to mitigate this concern. The initial building design included multiple shear walls to resist powerBy E. Michael McCarthy, P.E., M.ASCE ful coastal winds but, as the design progressed, many of those walls were future replacement of the equipment. The exterior walls of the build- reduced or eliminated to accommodate interior space planning. It ing are a combination of glazing, in-fill masonry, and concrete shear soon became apparent that the remaining shear walls were no longer walls. The size and configuration of the building dictated the need adequate. The 3D RAM analysis/design model was changed to supplefor expansion joints. The most logical location was between the main ment the shear walls with the natural stiffness of the columns and slab tower and lower building components on both the east and west sides. framing to develop shear wall/frame interaction. While this solution The building foundations were designed in accordance with a geotech- solved the lateral bracing problem, the reinforcing steel and shear head nical report prepared by the project geotechnical engineers, Driggers reinforcing in the slabs and columns had to be increased to account Engineering Services, Inc. (DESI). DESI conducted a series of deep for the additional bending moment transfer at each column-to-slab borings and discovered that the site is underlaid with mostly sandy intersection. The model did not indicate excessive diaphragm shear soils of varying densities down to lime rock deep underground. Based force transfer at the upper level(s) of the building at the interaction on those findings and point of the flexural deflection of the shear walls as they tried to the column and wall exceed the lateral shear distortion of the frame. All wind pressures loads provided by were input into the RAM 3D model as load cases in the overall design McCarthy, DESI rec- of the building. ommended reinforced concrete drilled shafts Flood Design or caissons for building support. The The 2010 FBC incorporates by reference ASCE 24-05 Flood Resistant drilled shafts varied Design and Construction which was used in conjunction with the in diameter from 2 national, state, and local flood regulations in the design of the Opal feet to 5 feet and were Sands Resort. These included two different FEMA V-zones, Pinellas drilled into the rock Gulf Beaches Coastal Construction Code, and the Florida DEP about 70 feet below CCCL. The lowest horizontal beam supporting the first elevated and the ground surface. occupied floor of the building was required to be either at or above Opal Sands with Sand Key Bridge in the background. Before construction, the design flood elevation (DFE), derived from the FEMA maps

Beach Hotel

STRUCTURE magazine

June 2017

Post-tension slab construction.

and the appropriate freeboard, to comply with these complex flood regulations. Also, all construction below the DFE must allow storm water to flow through the building unimpeded. Specifically, any wall that is not a shear wall, stair, or elevator wall was designed to withstand wind forces but must fail or breakaway under storm-driven wave action. In this case, the lower floors of the building are primarily open parking, making it easy to comply with the Flow-through requirements. Additionally, the mechanical and electrical equipment that is normally placed on lower levels of buildings had to

be located on platforms above the DFE. The most stringent flood code at the Opal Sands site is governed by the DEP and affects all construction located seaward of the CCCL. Fortunately, the CCCL wrapped around the west and north property lines of the site so that most of the building was not affected. However, the portion of the pool deck and tiki hut deck that was designed to cantilever out over the water was subject to CCCL requirements. Those decks had to be elevated above the cresting wave elevation established by the DEP to minimize the potential for damage from wave forces. Because this building has direct coastal exposure, the open plaza deck and two levels of parking below on the west side are reinforced with a special two-layer zinc coated rebar, and the concrete included a corrosion inhibitor admixture. Additionally, all hydrostatic and hydrodynamic forces were input into the RAM 3D model as a load case, in addition to waterborne debris impact forces, in order to account for these flood zone forces in the overall design of the building. This array of complex flood design codes required the use of unique and creative solutions in the design of Opal Sands.

Construction At McCarthy’s recommendation, OPL agreed to have a full-time representative from McCarthy on site during construction of the structural building components. That representative worked with the contractor on a daily basis in responding to RFI’s, reviewing shop drawing submittals, and reviewing the ongoing construction for conformance to the drawings and specifications. In addition to facilitating a smooth construction flow, this level of service at the job site allowed McCarthy to meet the requirements of the Florida Threshold Law as the Special Inspector. Also during construction, a particularly interesting development occurred. OPL noticed the large transfer beam above the lobby and its impact on the ceiling height. For the better of the building, they decided to remove the beam and insert the lobby column, which the transfer beam initially replaced, back into the floorplan. That column was originally designed to extend from the foundation to the ballroom roof above the 6th level but was discontinued at the 3rd-floor lobby as STRUCTURE magazine

RAM structural systems model.

an interior design decision. At the time it was decided to eliminate the transfer beam, all the upper slabs had already been poured, complicating the change. It was an “all hands on deck” moment with McCarthy’s engineering staff to develop a solution that could be completed quickly and without affecting ongoing construction. The biggest challenge was to support the upper slab levels from the time the transfer beam was demolished until the new column was in place and ready to carry the load. Other difficulties included de-tensioning post-tension cables in the beam and strengthening the drilled shaft foundation to handle the additional loads. Fortunately, the column stack below the lobby floor was found to have enough excess capacity and did not need to be reinforced. The final solution involved the following steps, all of which were completed successfully: 1) Install temporary shoring from the foundation up to the 5th floor. 2) De-tension the post-tension cables in the transfer beam. 3) Install new steel micropiles under the direction of the geotechnical engineer and expand the pile cap. 4) Demolish the transfer beam using a Tru-line hydraulic chipping hammer operated by remote control on the lobby floor. 5) Form, reinforce, and pour back the new column. After all this, the interior designers could work the column into the lobby space design after all.

Summary Construction of the Opal Sands Resort was completed in February 2016 and the building was an immediate success, opening with over 15,000 reservations. It has become the centerpiece of South Clearwater Beach, Florida, and one of OPL’s finest hotels. Looking back, the decision to have a wind tunnel test done was of the utmost importance as it resulted in overall reduced costs as well as a more accurate design. Additionally, having a full-time representative on-site during construction allowed McCarthy to assist the contractor and ensure the building was built in accordance with the structural drawings and specifications. Most importantly, the use of a 3D RAM analysis/design model synchronized with the Revit model allowed McCarthy’s engineers to design an accurate and economical structural system on a challenging site that is subject to high hurricane force winds and complex flood zone regulations.▪

E. Michael McCarthy, P.E., M.ASCE, is an Associate Vice President of McCarthy and Associates, a Division of Pennoni in Clearwater, FL. He is active in the design of buildings throughout Florida and the Caribbean and is currently a member of the ASCE-24 Flood Resistant Design and Construction Committee. He may be reached at mmccarthy@pennoni.com. June 2017

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By Silvian Marcus, P.E., F.ASCE, Gustavo J. Oliveira, P.E., Fatih Yalniz, P.E, and Nicholas Chack, P.E.

slender, high-rise pyramid is the architectural principle behind the 53W53 project, also known as the MoMA Tower or Torre Verre. The 1,050-foot tall high-rise is now under construction and, upon topping out in 2018, it will house 728,000 square feet of ultra-luxury residential condominium apartments. The complex includes amenities such as a 65-foot long lap pool, wellness center, wine vaults, and a private lounge with Manhattan skyline views, as well as 65,000 square feet of additional gallery space for the Museum of Modern Art. The 53W53 project was conceived by Developer Hines and designed by Jean Nouvel. WSP USA serves as both the structural engineer of record and the provider of engineering of building services. The construction manager of this expansive project is Lend Lease. The 53W53 tower is located in the heart of Midtown Manhattan, occupying a 17,000-square-foot midblock lot between Fifth Avenue and Sixth Avenue, with access from both 53rd Street and 54th Street.

Project Description The 53W53 site neighbors the New York City Museum of Modern Art to the west and the 1330 Avenue of the Americas building to the east. The project comprises two levels below grade and 82 floors above street level. The 53W53 tower will rise to a total height of 1,050 feet when complete, including additional architectural elements. The site is approximately a rectangle, measuring 195 feet between 53rd and 54th streets and 87 feet in the East-West direction, which results in a tower with a slenderness ratio of 1:12. Early in 2007, developers Hines, Goldman Sachs, and Pontiac Land Group carried out a comprehensive floor-area ratio analysis which showed that only by combining the air rights of its three neighbors was the 53W53 project able to reach the intended height envisioned by Pritzker Prize-winner, French architect Jean Nouvel, who is the head of Atelier Jean Nouvel and serves as the Design Architect. Also participating in the project are Adamson Associates Architects of Toronto as the Architect of Record, SLCE Architects, and The Office of Thierry W. Despont serving as the architects for interior design. STRUCTURE magazine

Panoramic 3D rendering of the 53W53 Project. Courtesy of Hines.

Architect Jean Nouvel’s concept for 53W53 – a super-slender tower of pyramidal form with northern and southern façades gradually sloping away at two distinct angles from 54th and 53rd streets, respectively – is founded on the aesthetics of exposed elements arranged in an asymmetric, almost random pattern on all façades of the building. Given the number of inclined members forming an unorthodox lattice, these patterns have been termed diagrids. The structural solution envisioned by WSP USA smoothly merged with the architectural intent by using the elements of the diagrid as a continuous system, providing support not only for the building envelope but also adequate in terms of stability and strength for the entire structure.

Foundation System

Installation of Node 4R-6.

The geotechnical composition of the site presented an interesting engineering challenge. While the Midtown area of Manhattan is typically associated with good quality substrate, with bearing capacity up to 60 tons-per-square-foot, the geotechnical surveys and studies performed for the 53W53 tower showed evidence of an old stream towards the west side of the site. The continuous presence of water linked to the stream was attributed to the gradual deterioration of the mechanical properties of the rock, which were evaluated to be only eight tons-per-square-foot. To prevent excessive settlements to the neighboring buildings, and in consideration of the adjacency of the South perimeter of the site to existing facilities of the New York City Metropolitan Transit Authority (MTA) along 53rd Street, the foundation system required the implementation of measures aimed at preventing adverse effects to neighboring underground structures. More than thirty reinforced concrete drilled caissons, each 3 feet in diameter, were required to extend a minimum of 30 feet below the subcellars. In selected locations near the MTA tunnels, the caissons reached 70 feet deep, matching the elevation of the subway track.

Superstructure The superstructure of the 53W53 tower is based on a unique dual-purpose system. In traditional high-rise projects, focus on efficiency and typical construction sequence results in the provision for two separate systems to carry gravity loads and lateral loads. However, the architectural intent of 53W53 required a uniquely different approach. A single, exterior structural system matching the geometry of the diagrid was developed having the ability to carry both vertical loads and those associated with wind and seismic demands. This approach, while efficient from a structural standpoint, presented numerous engineering challenges which were solved with a combination of rigorous analysis and innovative solutions.

The original structural concept considered forming the diagrid with steel members only. A value engineering study resulted in both financial and constructability advantages to using a reinforced concrete diagrid system consisting of vertical columns, inclined elements or braces, and horizontal spandrel beams. This structural solution allowed for the maximization of unobstructed interior spaces in the majority of floors. Perhaps the most challenging aspect of creating a reliable and resilient structural diagrid system was the conception, testing, and implementation of highly specialized nodal connections at almost every intersection of the diagrid. At least three dozen nodes required the interconnection of four structural members or more. On a given node, not only are there multiple diagrid members intersecting, but these elements might also be located on multiple planes and oriented at different angles. Take, for example, the first node installed on site, located on the North face of the sixth floor and the largest in the tower. Node 4R-6, as it was referred to in reference to grid lines and elevation, is the intersection of six major structural elements at various angles, all of which support a significant portion of the tower above. The most efficient solution to address the congestion of reinforcement and realignment of forces at the nodes was to develop a steel core assembly with special connectors to which the high-strength reinforcement, consisting of No. 20 (2.5 inches in diameter) bars, was anchored. The steel core served as the ideal transition element between various layouts and directions of steel reinforcement present throughout the building. The contractor was required to create a full-scale mock-up prior to the actual construction to validate the feasibility of this innovative solution. The outcome of the mock-up was satisfactory in terms of both quality control and reliability for on-site execution. Considering the client’s request for the lower floors of the tower to be used for 53W53 Project 3D Rendering. Courtesy of Hines. additional gallery space for the Museum

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Installation of Node 3A-10 showing a K-frame geometry.

Detail of top (left) and bottom (right) of Node 3P-31 before concrete casting operation.

of Modern Art, shear walls and interior columns were relocated to the periphery, allowing for a highly flexible floor layout. Steel trusses over these large, unsupported spans were used to pick up the loads delivered by the few interior columns and to transfer them to the east shear wall core. Another structural engineering challenge was the need to span over the existing ConEdison emergency generator servicing the Museum of Modern Art, which required a 24-foot long overhang cantilevered from the tower’s northeast corner. The overhang was integrally connected to the exposed members of the diagrid. The required lateral stiffness and strength of the tower was accomplished by placing outrigger walls, effectively connecting the central shear wall core to the exterior diagrid system at Levels 35, 36, and 37, which created the effect of a “belt floor” approximately at mid-height of the structure. These outrigger walls were added to the structure with negligible impact to the marketing plan, as these floors housed the mechanical rooms required to handle the distribution of the building services at that elevation. As the many façade planes converge at different locations and elevations of the building, small pyramids or apexes were created. To minimize the impact of these volumes on interior spaces and to facilitate their construction, it was determined that the five clearly delineated vertices of the building be constructed of steel framing. This solution allowed for the possibility of prefabricating portions of the frames to be erected after the surrounding concrete structure had been cast.

mass was increased towards the top of the building by using a slab thickness of 20 inches at the 73rd, 74th, and 76th floors. Second, a 650-ton Tuned Mass Damper was placed at the double-height space between the 74th and 76th floors.

Integrated Buildings SMEP Design WSP USA provided not only the services for structural design, but was also responsible for the engineering of building services including mechanical, electrical, plumbing, fire protection, telecommunications, and others. This allowed for a very close and effective collaboration among the design and construction teams and good interaction between engineering disciplines.

Construction Progress Foundation construction was completed at the end of 2015. At the time of this publication, the superstructure work has reached the 40th floor, which is approximately 500 feet above street level. As the tower tapers and reduces its footprint towards the top, construction progress is expected to be a 4- or 5-day cycle per floor, standard in New York City’s construction environment. The project is scheduled to top out in the summer of 2018. The installation of curtain wall began in 2016 and will be completed by the beginning of 2019.▪

Wind Tunnel Testing and Structural Analytical Modeling This intricate structure required complex analyses including full three-dimensional sequential construction analysis, with consideration to time-dependent changes in material properties and loading demands, which were used to estimate the required elevation and position compensations to be implemented during construction. In addition to analytical models, a series of comprehensive aerodynamic tests to accurately determine the wind pressures and wind-induced vibrations were performed by RWDI Consulting Engineers. A two-prong approach was necessary to achieve the industry-recommended comfort level for human occupancy in terms of vibration and acceleration due to wind. First, the structural

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Silvian Marcus, P.E, F.ASCE, is Director of Building Structures at WSP USA, Principal in Charge of the project. Gustavo J. Oliveira, P.E., is Vice President of Building Structures at WSP USA, Project Director. Fatih Yalniz, P.E., is Structural Analysis Manager and Vice President of Building Structures at WSP USA. Nicholas Chack, P.E., is Project Manager of Building Structures at WSP USA. Additional Credits Ahmad Rahimian, Ph.D., P.E, S.E, F.ASCE, is Director of Building Structures at WSP USA. Gerardo Aguilar, Ph.D., is Technical Manager of Building Structures at WSP USA.

June 2017

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Salesforce Tower NEW BENCHMARKS IN HIGH-RISE SEISMIC SAFETY

hen completed in 2017, Salesforce Tower will be the tallest building in San Francisco at a height of 1,070 feet (901 feet to the top occupied floor). This super-tall building advances the state-of-the-art of high-rise seismic design through implementation of a number of first-ever design and analysis methods that push limits and set new industry benchmarks. The structural innovations required to create this record-setting, city-defining tower address enhanced performance objectives, foundation challenges, and interactions with adjacent buildings… issues applicable both to this building and future tall buildings in areas of high seismicity.

By Ron Klemencic, P.E., S.E., Hon. AIA, Michael T. Valley, P.E., S.E., and John D. Hooper, P.E., S.E.

Reshaping the Skyline… Redefining the Future In the 1980s, the City of San Francisco enacted policies restricting high-rise development, ultimately creating a skyline that “plateaus” at 500 to 600 feet – the historical maximum zoned heights (Figure 1). The restrictions also limited the city’s annual allotment of space for new office development. These limitations were implemented during an anti-growth era sweeping the nation in the 1980s, fueled by fears of overcrowding and inadequate city infrastructure.

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Figure 1. San Francisco Skyline – Pre-Transbay District.

When studies confirmed that job centralization in a sustainable, transit-oriented downtown core was critical to San Francisco’s continued economic success, a new Transit Center District Plan was developed for a 16-block area referred to as the Transbay District. The plan called for a new Transbay Transit Center to co-locate eleven Bay Area transit systems and related surrounding development. The plan also designated a new zoning district around the Transbay Transit Center, reclassifying parcels to allow buildings from 600 to 850 feet tall versus the existing 500-foot cap. Through this plan, the City sought the development of office and residential buildings that were taller and more densely occupied to spur job and population centralization. The plan also targeted the creation of a “new, sculpted skyline formed by height increased immediately around the Transit Center” to mark the importance of the location and provide a new, vibrant image for San Francisco. To that end, the 50,515-square-foot site bounded by First and Fremont Streets, fronting Mission Street to the north and directly adjacent to the new Transit Center, was designated as the location for the area’s tallest building. At the new 1,070-foot height limit, the record-making building would serve as a beacon for the revitalized area (Figure 2). To help fund the new Transit Center, the rights to entitle and purchase the highly desirable “tallest building” site were awarded through a 2007 global invitation-only design competition sponsored by the Transbay Joint Powers Authority, a public entity created to develop the Center. The winning proposal was submitted by Hines, an international development firm, teamed with Pelli Clarke Pelli Architects. (Developer Boston Properties subsequently acquired 95% of Hines’ stake in 2013.) The team envisioned an iconic 61-story, high-density, sustainable office tower totaling 1.5 million square feet (and ultimately named “Salesforce Tower” for its primary tenant). Above the 26th floor, each elevation of Salesforce Tower curves and tapers away from the street, creating a narrow, slender top finished with a “sculptural crown”. The crown, designed as an unenclosed latticework of structure, continues the expression of metal wrapping the occupied floors of the tower below and contributes to the slender proportions and total height of the building form. The building’s curvature also reduces the apparent height and massing of the building when viewed by pedestrians immediately below. Within the larger context of San Francisco’s future skyline created by the increased building heights in the TCDP area, Salesforce Tower will be the tallest point, marking the significance of the adjacent Transit Center. Construction started in 2013, with completion expected in 2017. STRUCTURE magazine

New Benchmarks in Seismic Safety Given the scale of Salesforce Tower, the calculated number of building occupants will far exceed the building code threshold of 5,000 people, triggering the building’s consideration under Occupancy [or Risk] Category III. Category III buildings require additional safety for wind and seismic demands, thus prompting new challenges for the engineering team. Traditional structural design methods adopt an “enhanced strength” approach when attempting to improve seismic safety and performance. This approach, while easy to implement by applying code-defined seismic forces that have been amplified by an Importance Factor (1.25 for Category III buildings), fails to ensure enhanced building performance when subjected to extreme seismic ground shaking. In fact, in the context of the historical seismic design philosophy of ductility and energy absorption, enhanced strength may, in fact, be detrimental to building performance. Stronger buildings resist more force rather than absorbing the energy of the ground’s shaking. In resisting these higher forces, shear stresses and foundation demands increase to undesirable levels and building performance can be compromised. Instead, a rigorous Performance-Based Seismic Design (PBSD) approach was implemented to allow for, quantify, and control desired building performance at an enhanced level compared to other commercial office buildings. continued on next page

Figure 2. San Francisco Transbay District. Courtesy of Foster + Partners.

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Figure 3. Typical low-rise structural floor plan.

Figure 4. Confirmation of Salesforce Tower’s enhanced performance.

The project’s structural engineer, Magnusson Klemencic Associates (MKA), brought to the project decades of leadership in seismic design in San Francisco, including multiple high-rise towers. The firm also led the development and U.S. adoption of PBSD for high-rise buildings, a methodology that produces buildings that are safer and perform more predictably and reliably than buildings designed following a prescriptive code-based approach. PBSD uses sophisticated nonlinear seismic time-history computer modeling – practical only in recent years thanks to advancements in computing capacity and user-friendly analysis programs – to examine building performance during multiple predicted seismic events. With PBSD, engineers can analyze complex and unique building geometries then precisely allocate appropriate strength and stiffness to achieve an efficient design that meets the desired performance objectives. PBSD methodology not only meets the intent of the building code but explicitly considers and quantifies predicted building performance under multiple ground-shaking scenarios. This explicit examination of building behavior provided through nonlinear time-history analysis, coupled with the application of enhanced performance standards, produces a safer, more reliable building. MKA had already designed five PBSD high-rise buildings in San Francisco, but none that targeted the more stringent Occupancy Category III performance objective. Because PBSD is a relatively new and complex approach, building departments require that such designs be peer reviewed by a second engineering team to verify that building performance meets code intent. This peer review process has been instrumental in the understanding, advancement, and acceptance of PBSD methodologies throughout the industry, as code-prescriptive intent becomes successfully “translated” into approved performance-based designs. After assessing the more stringent Category III code-defined performance objective, and evaluating that intent and application relative to PBSD methodology, the design team targeted a reduction to 6% (from 10%) of the probability of collapse under a Maximum Considered Earthquake (MCE) ground shaking. This is consistent with ASCE 7 Commentary related to Occupancy Category III buildings. As stated in the ASCE 7 Commentary, the performance objective for Occupancy Category III structures is to “reduce the hazard to human life in the event of failure,” which relates most closely to a code-defined performance of Collapse Prevention with MCE shaking. The structural design of Salesforce Tower includes more stringent Acceptance Criteria for MCE shaking as explicit performance objectives, including the following: • Reduced story drift • Reduced coupling beam rotations • Reduced tensile/compressive strains in shear walls • Reduced shear demands on shear walls • Risk Category II acceptance criterion was typically modified to be more stringent by applying a factor of 0.8 The detailed Acceptance Criteria used for the design of Salesforce Tower is shown in Table 1. The structural system features a gravity load-resisting system with structural steel columns and floor framing supporting steel composite deck. The building’s seismic force-resisting system comprises special reinforced concrete shear walls, 24 to 48 inches thick, at the central elevator and stair core. Since the vertical elements of the tower’s seismic force-resisting system include

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only shear walls (Figure 3), the City of San Francisco’s Administrative Bulletin 083 (AB083), Requirements and Guidelines for the Seismic Design of New Tall Buildings using Figure 5. Typical LBE rebar detailing. Non-Prescriptive Seismic-Design Procedures, applied. In addition to technical provisions, AB-083 called for a structural design review by a panel of independent experts. The four-member panel assembled for the detailed review of the analysis and design of Salesforce Tower included two practicing engineers, a research professor, and a geotechnical/seismic ground motion expert. Although wind-loading conditions for the building are not trivial, wind tunnel testing confirmed that demand levels fell below seismic demands, and that occupant comfort standards would be met as judged against international standards. The lateral design of Salesforce Tower was driven by seismic loading conditions for three levels of ground shaking:

• Elastic performance targeted for service-level shaking (with a mean recurrence interval of 43 years) • Moderate structural damage expected for design-level shaking (taken as two-thirds of code-defined MCE shaking) • Collapse prevention, with a reduced probability of collapse consistent with Occupancy Category III, targeted for MCE shaking The nonlinear time-history analyses used to confirm the structural response to MCE shaking employed two suites of 11 pairs of acceleration history. Two suites of ground motions were developed considering a Conditional Mean Spectra approach, targeting the first and second modes of vibration of the tower. This approach was deemed to more rigorously and appropriately test the building’s design, given the importance of the tower’s second-mode-of-vibration response. The suites were developed by performing three-dimensional, nonlinear site response analyses using input rock motions that were selected, scaled, and matched for Conditional Mean Spectra spanning the period range of interest from approximately 0.5 seconds to 9.0 seconds. One suite was conditioned to represent long-period motions, for which 8 of 11 motions include pulse effects. The other suite was conditioned to represent shorter period motions, for which 2 of 11 motions include pulse effects. Pulse effects were considered particularly important for this building given the proximity to the San Andreas Fault. The resulting acceleration histories, taken at the foundation level instead of the more traditional ground surface level, Item Value include kinematic soil-structure interaction effects Story Drift 3.0x0.8 = 2.4 percent taken as the average of 11 analyses; due to base-slab averaging and embedment. A scale 4.5x0.8 = 3.6 percent maximum from any single analysis. factor for each suite of motions was applied so that the corresponding linear response spectra satisfy the Residual Story Drift 1x0.8 = 0.8 percent taken as the average of 11 analyses; requirements of the building code. 1.5x0.8 = 1.2 percent maximum from any single analysis. The results of these 22 earthquake simulations were evaluated and compared against the targeted Coupling Beam Rotation 0.06x0.8 ≈ 0.05 radian rotation limit, taken as the average Acceptance Criteria. Where predicted demand levels (diagonally reinforced of 11 analyses. exceeded Acceptance Criteria, design modifications concrete or steel composite) were implemented. In particular, core wall thicknesses were tuned to reduce and control shear demands Shear Wall Reinforcement Rebar tensile strain is 0.05x0.8 = 0.04 in tension and within acceptable limits at the tower’s base and the Axial Strain 0.02x0.8 = 0.016 in compression, taken as the average of location of a core setback at Level 50. Ultimately, it 11 analyses. was demonstrated that all Acceptance Criteria had been achieved, and the building’s enhanced perforFully confined concrete compression strain is 0.015x0.8 = Shear Wall Concrete mance was confirmed. As shown in Figure 4, story 0.012, taken as the average of 11 analyses. Axial Strain drifts and coupling beam rotations typically fall well within acceptable limits, wall shear demands remain elastic, and vertical wall strains are quite modest with Shear Wall Shear DCR limited to 1.0. Capacity calculated using expected only limited yielding predicted. material properties and a phi factor of 1.0, where ductility demands are modest (ec ≤ 0.005; es ≤ 0.010) or a phi factor of 0.75, where ductility demands are greater. Demand taken as 1.5 times the average demand from each set of 11 analyses. Level P1, P2, and Level 1 Diaphragms

DCR limited to 1.0. Capacity calculated using expected material properties and code-specified phi-factors (ɸ = 0.75). Demand taken as 1.5 times the average demand from each set of 11 analyses.

Basement Walls

DCR limited to 1.0. Capacity calculated using expected material properties and code-specified phi-factors (ɸ = 0.75). Demand taken as 1.5 times the average demand from each set of 11 analyses.

Table 1. MCE-Level Acceptance Criteria for Salesforce Tower.

STRUCTURE magazine

San Francisco’s Deepest Foundations The Salesforce Tower site is underlain with a complex soil strata including fill, sand, San Francisco’s “old bay clay,” and weak bedrock. These geotechnical conditions are subject to potential liquefaction, lateral spreading, excessive settlement, and inadequate foundation support. Given the poor soils and the sheer weight of Salesforce Tower, supporting the building on anything but bedrock was not feasible. Gravity loading and overturning demands at the foundation level from MCE shaking dictated a piled-mat solution. Two foundation systems were considered during the design process: 8-foot-diameter drilled shafts and 5.0x 10.5-foot Load-Bearing Elements (LBEs). As the

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depth to rock from existing grade was approximately 250 feet, and socketing into the rock would require drilling even deeper, the limits of available drilling equipment would be tested for a drilled shaft foundation. The alternate LBE, or “barrette” foundations, were not Figure 6. LBE foundation system. subject to the same depth limitations as the equipment used to excavate the shafts was a combination of a line-supported clam shell and hydrofraise. Ultimately, the LBE foundation system was selected as the most appropriate for the project. An extensive analysis of the LBEs, considering extreme seismic demands, was performed. Reinforcing detailing was incorporated to resist the high tensile, flexural, and shear stresses imposed on the LBEs by MCE ground shaking. Confinement reinforcement was also specified in the upper zones of the LBEs where compressive demands were the highest (Figure 5, page 47). As this was the first time LBEs would be used to support a tall building in San Francisco, an extensive review was conducted by the independent peer review panel. Also, two full-scale Osterberg Load Cells tests were conducted to confirm that the design parameters for the skin friction on the LBEs were appropriate. Through this test program, it was confirmed that the skin friction values were time sensitive (as expected) due to the build-up of filter caking of the bentonite on the side walls of the shafts, requiring time limits to be placed on the overall installation of each shaft. Installation of the foundation system included first the construction of guide walls to control the location and excavation of the LBEs. Excavation then proceeded using a combination of a line-mounted clam shell in the sands and clay, switching to a hydrofraise when denser rock material was encountered. Excavation stability was maintained throughout the process using a recycling bentonite slurry system. After the excavation was completed, full-length, pre-tied reinforcing steel cages were lowered into the bentonite-filled holes, and concrete was placed using dual tremie pipes. All 42 LBEs were constructed from existing grade, rather than the bottom of the 60-foot excavation. A temporary internal bracing system would be required to supFigure 7. Salesforce Tower recent topping out. port the open excavation Courtesy of Magnusson Klemencic Associates/ given the limitation of the Michael Dickter. adjacent Transbay Transit STRUCTURE magazine

Center’s simultaneously open excavation, limiting the ability for heavy equipment to work at the bottom of the hole. As such, concrete placement in the LBEs continued through elevation of the future mat foundation, and the remainder of the shaft was filled with a lean mix for ease of later excavation. The final foundation configuration for Salesforce Tower includes 42 LBEs interconnected by a thick mat foundation to enforce compatibility (Figure 6). The mat varies in thickness from 14 feet at the core to 5 feet at the perimeter. LBEs extend into the underlying Franciscan bedrock, some reaching more than 310 feet below existing grade with rock-sockets of up to 70 feet. The design and construction of this foundation system set new standards for the support of tall buildings in San Francisco’s unique geotechnical and seismic conditions.

Unprecedented Structure-Soil-Structure Interaction (SSSI) Analysis As a condition of the Salesforce Tower site purchase agreement, the Transbay Joint Powers Authority required proof that the tower developed for the site, as well as its interactions with surrounding buildings, would not negatively impact the new Transit Center, especially during the strong ground shaking of an MCE event. An unprecedented Structure-Soil-Structure Interaction (SSSI) analysis was conducted to investigate and confirm the performance of Salesforce Tower and its impacts on the adjacent Transbay Transit Center. MKA’s structural engineering team and the geotechnical engineering team at Arup performed three-dimensional, nonlinear SSSI analyses to assess the interactive performance of Salesforce Tower, the Transbay Transit Center, and another immediately adjacent high-rise tower. This extremely complex assessment involved extensive nonlinear computer models developed in CSI-Perform and LS-DYNA and considered multiple seismic ground motions. The results of these analyses confirmed the satisfactory performance of both Salesforce Tower and the Transbay Transit Center. The details and conclusions of this analysis were ultimately reviewed and approved by the Engineer of Record for the Transbay Transit Center. This was the first time that potential impacts of one building on a neighboring building during strong seismic ground shaking had been considered.

The Road Ahead In addition to its height, the design and construction of Salesforce Tower set new benchmarks for seismic safety, foundation construction, and structure-soil-structure interaction. With many new towers planned in San Francisco and other seismically active cities, the details of Salesforce Tower’s design will help guide and inform future improvements in the safety and performance of high-rise buildings… and change the San Francisco skyline forever (Figure 7).▪

Ron Klemencic, P.E., S.E., Hon. AIA, is Chairman and Chief Executive Officer at Magnusson Klemencic Associates. Ron can be reached at rklemencic@mka.com. Michael T. Valley, P.E, S.E., is a Principal at Magnusson Klemencic Associates. Michael can be reached at mvalley@mka.com. John D. Hooper, P.E., S.E., is Director of Earthquake Engineering at Magnusson Klemencic Associates. John can be reached at jhooper@mka.com. June 2017

Brock Commons By Paul Fast, P.E., P.Eng., Struct.Eng., FIStructE and Robert Jackson, P.Eng.

he tallest contemporary wood building in the world was recently constructed at the University of British Columbia (UBC) in Vancouver, Canada. Brock Commons is an 18-story student residence that is a mass timber hybrid and measures a record 174 feet (53 meters). Fast + Epp are the structural engineers, working in conjunction with Acton Ostry Architects and Hermann Kaufmann Architekten. (The Sakyamuni Pagoda of Fogong Temple built in China in 1056 stands 220.83 feet (67.31 meters) and is the tallest wood building in the world.)

Project Background UBC is experiencing an increase in demand for student housing and has a sustainability commitment to a campus that acts as a “Living Laboratory” where innovation is encouraged, not only in academia but also in building and infrastructure. By pairing this drive with the potential for external funding related to mass timber research, the project was born. STRUCTURE magazine

A Case Study in Tall Timber

The key goals of the project were to create a safe, functional, sustainable, and cost-effective residence for UBC students. Delivering a mass timber building with a construction cost that aligned with the unit cost of a comparable traditional concrete tower in Vancouver was an important goal, demonstrating the viability of wood as a practical material for tall building applications. An integrated design team was assembled by the University to facilitate this effort. The construction manager was appointed, and the timber installer and concrete trades joined the team in a design-assist role, providing real-time feedback on the evolving structural design and offering valuable constructability advice. The structure is comprised of 17 stories of five-ply cross-laminated timber (CLT) floor panels, a concrete transfer slab on the second floor, and a steel framed roof. The CLT panels are point supported on glulam columns on a 9.35- x 13.1-foot (2.85m x 4.0m) grid. Beams were eliminated from the design by utilizing CLT’s two-way spanning capabilities. Two full-height concrete cores provide the lateral stability for the structure.

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Figure 1. Timber erection – Summer 2016. Courtesy of Seagate Structures and Pollux Chung.

Figure 2. Point supported CLT system. Courtesy of Seagate Structures and Pollux Chung.

Structural System The design intent was to keep the structure simple and sensible: develop a prefabricated “kit-of-parts” that could be installed quickly and easily, with minimal labor on site. CLT is often used as a one-way decking system, ignoring the twoway spanning capability afforded by its cross laminations. By utilizing CLT to span in both directions, the design team was able to eliminate beams, significantly reducing the overall structural depth (Figure 2). This created a clean, flat, point-supported surface, allowing for unobstructed service distribution as is commonly found in flat-plate concrete construction. Further, by adjusting the column grid and architectural program to suit the maximum available panel size, the team was able to both minimize the overall number of panels and maximize the efficiency of the system. The primary lateral support for earthquake and wind loading is provided by two concrete cores. Although timber-based lateral forceresisting systems such as CLT walls/cores, timber braced frames, or post-tensioned/self-centering systems were feasible design options for this project, the testing, time, and costs required to obtain regulatory approvals would have negatively impacted the client’s budget and completion date.

Design Challenges Codes and Standards The current British Columbia Building Code (BCBC 2012) limits the height of wood buildings to six stories. As such, a special approval process was required for this project. The design is based on a Site Specific Regulation (SSR), administered by the Building Safety and Standards Branch of the BC Provincial Government, and is applicable solely to this project and site. Due to the complexity of the project, two independent structural peer reviews were completed. The first independent review was timberfocused and was completed by Merz Kley Partner ZT GmbH in Dornbirn, Austria. The second was seismic-focused and was completed by Read Jones Christoffersen Consulting Engineers in Vancouver.

were CNC machined with quality control protocols to better ensure a seamless erection of the timber superstructure. To help achieve a high level of prefabrication for all design disciplines, CadMakers, a third-party consultant, modeled the building and helped coordinate design documents before and during construction. This 3D model, created with CATIA software, includes fully-detailed structural elements and connections, as well as mechanical/electrical systems, architectural fit-outs, formwork panels, and safety guards. The model allowed all CLT penetrations for mechanical and electrical sleeves to be fully coordinated during the design process and successfully converted into fabrication files (CAD/CAM) needed for CNC machining. Point Supported CLT In addition to stiffness and bending requirements, rolling shear stresses at the column supports are typically a controlling factor in two-way, point-supported CLT floor plates. A rolling shear failure is one in which the fibers “roll over” each other, due to shear forces perpendicular to the grain. After designing the custom lamination layup to suit the rolling shear and flexural demands, the design team completed 18 full-scale load tests, at the FPInnovations laboratory in Vancouver, on panels from three prospective CLT suppliers to validate the analysis. There appeared to be some capability for the CLT to redistribute forces, as internal shear cracks propagated through the panel before the critical failure mode occurred. Multiple types of shear/bending failures were observed near the supports (Figure 3). Column Shortening and Shrinkage In tall wood buildings, axial column shortening needs to be considered during design. When properly accounted for, the shortening should not negatively affect the construction, use, or long-term performance of the building. continued on next page

Prefabrication Prefabrication is an essential consideration when designing large-scale wood structures. Well-planned erection and shop drawings are vital to ensuring smooth production and installation of timber elements. This results in fewer errors on the site, less remedial work, and a shorter overall construction schedule. All CLT and glulam elements STRUCTURE magazine

Figure 3. Point supported CLT panel testing apparatus (left); Failure (right).

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Construction Proof of Concept Mock-Up The construction team built a full-scale mock-up of a portion of the building to validate the constructability of the proposed design, 26 feet x 39 feet (8m x 12m) in plan and two stories tall. The mock-up included several connection types to help determine and optimize the details used in the final design (Figure 4). Also, the mock-up was used for the development and evaluation of various building envelope systems considered for the project. Construction Sequencing Figure 4. Proof of concept mock-up.

Several factors affect glulam column shortening: • Dead load elastic axial shortening (Δ = PL/AE) • Live load elastic axial shortening (Δ = PL/AE) • Shrinkage parallel to grain • Joint settlement • Column length tolerances • Wood creep The main concerns surrounding these shortening effects are the impact of the deformations on the vertical mechanical services, as well as the differential movement between the wood superstructure and the stiff concrete cores. The effects of these factors culminate at the roof level, where all columns below contribute to the shortening. A series of 1⁄16-inch thick steel shim plates were added during construction at the column-to-column connections on three strategic levels to mitigate a portion of these effects.

The construction team erected the concrete cores to full height and installed the L2 transfer slab throughout the winter of 2015/16 to facilitate the use of one crane and provide sufficient time for manufacturing and shipping of the heavy timber elements. In June 2016, the timber and envelope installation began. This was completed in four phases. The first involved erecting all columns on one level, diagonally bracing them, and using horizontal spreader bars at the column caps to set the grid (Figure 5). The columns were installed by hand from bundles on the active deck, freeing up the crane for envelope panel installation. The second phase was the installation of the CLT panels, stitching adjacent panels as the active deck moved away from the cores (Figure 6). The third phase was the installation of the steel drag plates at the concrete cores and perimeter angles to support the curtain wall system. The fourth was the installation of the envelope elements on the floor below the active deck (Figure 7). Erection of the timber and envelope panels was completed in just nine weeks, with the four-step installation sequence repeating itself.

Monitoring To better understand the unique behaviors of the building, the structure will be fitted with accelerometers, moisture meters, and vertical shortening string pots. Research teams at the University of British Columbia are undertaking this work as a part of the “Living Laboratory” initiative. The data collected from the accelerometers and inclination gauges will help to verify the building’s performance in a significant seismic event. The string pots will measure the floor-to-floor axial column shortening at strategic levels, which will provide more insight into axial column shortening in highly loaded glulam columns. Lastly, moisture meters and data loggers will be installed in the CLT panels, collecting data from the manufacturing plant to the final installed condition. The meters will continue to measure moisture content throughout the service life of the building which, in a few years’ time, will give a moisture content timeline from fabrication to moisture equilibrium.

Figure 5. Braced columns. Courtesy of Seagate Structures and Pollux Chung.

Conclusion A mass timber building of this scale carries a unique set of engineering challenges, many of which can be mitigated through the use of innovative design strategies and strong quality control protocols. To date, the project has proven cost-competitive with concrete towers in the local marketplace, largely achieved by an integrated design team, real-time input from trades, and structural discipline. This large scale prefabricated project is a testament to fresh thinking and holistic design.▪ Paul Fast, P.E., P.Eng., Struct.Eng., FIStructE, is the Founding Partner of Fast + Epp. He is the structural engineer of record for the innovative 18-story Brock Commons student residence. Robert Jackson, P.Eng., was heavily involved the design and construction of the Brock Commons student residence project.

Figure 6. CLT panel installation. Courtesy of Seagate Structures and Pollux Chung.

STRUCTURE magazine

June 2017

Figure 7. Perimeter envelope installation. Courtesy of Naturally: Wood and Steven Errico.

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1. Planibond ® 3C (corrosion inhibitor) 2. Planitop 11 SCC (self-consolidating concrete mix with silica fume and corrosion inhibitor) or Planitop XS (hand-applied repair mortar) or Planitop 12 SR (low-pressure spray mortar) 3. MapeWrap™ Primer 1 (saturant) 4. MapeWrap 11 or MapeWrap 12 (epoxy smoothing putty) 5. MapeWrap 31 or MapeWrap 21 (epoxy saturant) 6. MapeWrap C Uni Ax 300 or 600 (carbon fiber fabric) 7. MapeWrap 31 (epoxy saturant) 8. Sand broadcast 9. Elastocolor Coat (protective and decorative coating)

1. Mapefer ™ 1K (anticorrosion coating) 2. Planitop 18 ES (rapid-hardening repair mortar) or Planitop 18 (very rapid-hardening repair mortar) 3. Planiseal ® LVB (crack healer/sealer) 4. Planiseal Traffic Coat FS (fast-setting epoxy overlay for bridge decks)

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By Mark Sarkisian, S.E., Eric Long, S.E., David Shook, P.E., and Eric Peterson

y re-examining standards long taken for granted by developers, 350 Mission received LEED® Platinum certification and is reinventing a ubiquitous building type. The form, structure, and systems of this Class-A office tower are generated by rigorous goals of environmental performance, social engagement, and material efficiency. The 30-story volume embodies the higher workforce densities and flexible space planning of 21st-century offices, thanks to floor plates that span nearly 45 feet between core and perimeter. 350 Mission achieves a new paradigm for office tower structures by utilizing post-tensioned long-span flat concrete slabs, a method traditionally reserved for residential construction. The innovative structure lifts the first office floor to create a transparent 50-foot-tall by 43-foot -deep lobby, dubbed the “urban living room.” Energized by a 70- by 40-foot LED screen, the generous public space has 90 linear feet of glass panels that slide open to the sidewalk and blur the threshold between public and private realms.

Performance-Based Seismic Design Dual seismic force resisting systems are required by ASCE 7 for buildings over 240 feet unless appropriate justification is provided. Administration Bulletin 82 and 83 of the San Francisco Building Code prescribe the design and peer review process needed to demonstrate code-equivalence of the core-only seismic force resisting system. This performance-based seismic design approach permits the core-only seismic force resisting system and avoids traditional dual systems which often include costly moment frames. The latest advances in nonlinear time history analysis, seismic design methods, and reinforcement detailing were incorporated into this project, drawing upon the knowledge and experience of the cityappointed seismic design review panel. All gravity framing members and their associated effects on building performance, including P-delta effects and nonlinear behavior, were modeled. Due to the potential interaction of the slabs on the gravity columns and their cumulative effect on the tall lobby condition, this was vital. Additionally, ground STRUCTURE magazine

floor and all basement diaphragms were modeled with shell elements, as the ramp and basement levels are sloped in a ‘corkscrew’ basement.

Long-Span Flat Plate Slab Design A flat plate solution using conventional practices would need to be at least 14 inches thick, with very high quantities of post-tensioning and reinforcement. To reduce the slab thickness to 11 inches and post-tensioning quantities to levels commonly associated with 30-foot residential spans, a cambered solution was proposed in tandem with the post-tensioning system. The primary purpose of the post-tensioning system is to counteract a large portion of the slab self-weight while mitigating flexural cracking. Due to the magnitude of the span and modest post-tensioning, elastic and inelastic creep deflection of the slab, up to 2.5 inches, was anticipated. This would not be acceptable for non-structural components such as partitions and could be visually perceivable in an exposed ceiling condition. Thus, a digitally mapped camber plan and camber values for individual shoring posts were developed with the collaboration of the concrete contractor. Changes in camber slope were also coordinated with the concrete contractor, Webcor Concrete, to ensure the specified geometry could be achieved without significant changes in labor. The direct collaboration of the structural design team with all concrete field superintendents, including those responsible for shoring, forming, placement, and finishing, were vital. Conventional 5,000 psi concrete was utilized along with A615 Gr. 60 reinforcement for both the post-tensioned floor slabs and in non-cambered, mild-reinforced below-grade slabs. Average compressive strength determined in cylinder testing from the below-grade slabs was to be 7,200 psi. The high strength was due to contractor requirements for early strength to tension tendons to keep the project on schedule. Thus, the camber of above-grade decks was adjusted using this more accurate concrete strength. Two comprehensive analytical investigations were conducted in parallel using separate software packages to ensure both strength and serviceability were satisfied. The primary software used for

June 2017

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Chamber overlay diagram.

Exterior at night.

analysis was SAFE® by Computers and Structures, Inc., while a checking analysis model was built in ADAPT Floor® by ADAPT Inc. Both analysis tools calculated cracked section properties based on actual rebar placement for increased accuracy. With the incorporation of camber, post-tensioning material quantities were competitive with shorted 30-foot spans often encountered in high-rise residential towers. For calculation of long-term deflection, methods recommended by ACI Committee 435 in ACI 435R-95, Control of Deflection in Concrete Structures, were utilized. As noted in the ACI 435 document, methods recommended by Graham and Scanion (1986) are appropriate where stiff lateral systems such as shear walls are utilized. The concrete modulus of rupture is lowered from 7.5√fc to 4√fc for calculation of cracked section properties. Also, the computed deflections using cracked section properties are amplified by a factor of 3.5 to determine the total long-term deflection. While both analysis models could amplify ACI 435 recommend values of cracked section deflections, SAFE was found to be more accurate with its age-adjusted modulus of elasticity method which accounts for long-term creep and shrinkage based on methods suggested by ACI 209. Although recommendations of ACI 318 for the modulus of rupture (7.5√fc) and long-term deflection multipliers (3.0) are appropriate when compared to laboratory testing, they do not account for critical construction effects such as early shrinkage cracking due to restraint and early loading of the concrete when shoring is removed. It is highly recommended that appropriate provisions be made in ACI 318 to give guidance to structural engineers designing concrete gravity framing which more appropriately addresses these important issues, and that practicing engineers consider this during design. Slabs are engineered to be flat and to deflect no more than ¾ inch between the core and perimeter 90 days after casting when raised flooring is to be installed. The ¾-inch deflection limit is important as that is the maximum deflection a standard partition slip track can accommodate. Slab elevations were surveyed at casting, as well as 30,

60, and 90 days afterward. The tracking of deflections was determined to be very close to analysis model predictions, which incorporated a novel iterative cracked section analysis procedure recommended by the ACI 435 committee and creep and shrinkage methods recommended by the ACI 209 committee. This design, with contractor collaboration, has set a new standard in office buildings and created a new efficient architecture which gives further enhancement to core-only tall buildings. An important feature of the concrete framing design is a dramatic 30-foot corner cantilever achieved with a 25-inch deep post-tensioned upturned beam. The upturn is concealed in a raised floor system which permitted underfloor mechanical air circulation, electrical conduits, and plumbing lines. Thus, the underside of the slab is kept free of these visual hindrances. The corner post-tensioned beam has a unique tendon profile that differs from conventional post-tensioned layouts. Instead of starting high at the support and draping linearly to middepth at the cantilever tip, it has a slight parabolic drape between the two ends. This was incorporated to avoid sagging at the cantilever mid-span which can occur in very long cantilever conditions. In elevation, the tower superstructure appears as 11-inch plates. The depth of similarly performing steel or waffle-slab construction would measure nearly 3 feet. If the concrete slabs remain exposed overhead, then the typical office floor will reach over 11 feet high, a dramatic increase from the 9-foot heights traditionally associated with Class-A office buildings. Using ultra-thin concrete instead of steel maximizes perimeter glass, achieves more daylight, and supports energy efficient HVAC systems.

STRUCTURE magazine

Innovative Multi-Story Construction Methods Construction methodologies were coordinated among the concrete, rebar, post-tensioning crews, and the place and finish crews. Handset Pro-Shore formwork was used to achieve non-traditional camber profiles and reduce the floor-to-floor cycle from 7 to 5 days. The design team met with the crew superintendents to ensure a full

June 2017

Lobby.

Lobby at night.

understanding of the design intent was conveyed and to receive feedback on improving the specified camber profiles and post-tension layout. The steps required to achieve the soffit profile, before concrete placement, included: 1) Measure deflection of lowest reshored level during placement; 2) Measure shortening of formwork system during placement; 3) Interpolate values based upon shore location relative to supports; 4) Agree upon method of strike-off (screeding) during concrete placement; 5) Add deflection plus formwork shortening to specified camber values; and 6) Survey and record deck soffit formwork elevations before placement. There were preliminary and follow-up discussions regarding what placement methods would be best for providing the indicated camber at the tops of the slabs. Since concrete placement strike-off is done with straight edges, the resulting camber is initially corded. Ordinarily, on flat building slabs, strike-off is done with screeds measuring up to 20 feet long. After consideration of both production requirements and the required surface geometry, it was decided by the contractor and engineer that the maximum screed length would be 10 feet. The direction of screeding was also discussed and agreement made on a standard direction. During the first placement, an additional observation was that ride-on trowel machines, using float pans, had the effect of planing or leveling the surface, which reduces camber. From this point forward, neither ride-on machines nor float pans were used. The camber profile was finished by using walk-behind trowel machines with combination blades. In addition to setting top profiles with a laser level, both 10-foot screeds and the use of walk-behind machines contributed in producing the required camber tolerances at the tops of the slabs. The ongoing process of taking measurements before and after each placement provided necessary feedback to the formwork and placing crews of where the processes needed fine-tuning and helped to keep the quality consistent. Top-of-slab elevation surveys were made at specific locations, relative to Project control, so that they could be exactly repeated for future monitoring. The result was a design and construction methodology that could be applied to other multi-story buildings. Understanding long-term slab deflection characteristics, monitoring as-built tolerances, and incorporating predicted slab performance characteristics into the design of architectural finishes and work scopes after concrete placement and finishing, aligns expectations through both the design and construction processes. STRUCTURE magazine

Field Verified Embodied Carbon The Environmental Analysis Tool™ is an embodied carbon accounting methodology and software published as a free download by SOM at www.som.com. The EA Tool considers embodied carbon from materials, construction activities, and probable seismic damage. Embodied carbon accounting metrics were used in all design decisions and led to an efficient design employing the use of recycled materials. During construction, engineers made weekly site visits, creating detailed reports of all construction activity and equipment used. This information, as well as actual material information, was used to adjust and validate embodied carbon accounting associated with structural materials and construction activity from the beginning of excavation through topping out. Results reveal the EA Tool methodologies are quite accurate, especially for above-grade work. Results also indicate that below-grade work is very carbon intensive, more than twice that of above-grade work.

Performance Performance is often quantified, but not often proven. For 350 Mission, seismic performance has been validated through nonlinear time history analysis, cambered slab deflection analyses have been proven through numerous surveys, and embodied carbon associated with materials and construction activity has been verified through field observations. This assurance in performance is important to advance design techniques towards a new standard in building construction and design methodologies.▪

Mark P. Sarkisian, S.E., is the Structural and Seismic Engineering Partner of the San Francisco office of SOM. He can be reached at mark.sarkisian@som.com. Eric Long, S.E., is a Director and Senior Structural Engineer of the San Francisco Office of SOM. He can be reached at eric.long@som.com. David Shook, P.E., is an Associate Director and Structural Engineer at the San Francisco office of SOM. He can be reached at david.shook@som.com. Eric Peterson is the Construction Manager at Webcor Concrete and can be reached at ericp@webcor.com. June 2017

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TALL BUILDING CONSTRUCTION

New Products and Services Keep Pace with Construction Changes for Tall Buildings By Larry Kahaner

ompanies involved in tall building construction are seeing increased growth thanks to advances in ground improvement, software, and materials. At Hayward Baker (www.haywardbaker.com), Director Jeff Hill has seen two significant changes in the specialty geotechnical construction industry. “The first would be the wide-spread acceptance of ground improvement as a foundation solution. The second is an increased use of geotechnical instrumentation to streamline design and allow for real-time subsurface risk mitigation for the owner and project team alike.” Hill adds a warning, though: “From coast to coast in North America, ground improvement has allowed the use of economic spread footing foundation solutions. Industry experience and satisfactory building performance have increased the willingness of engineers and owners to consider its use further. Perhaps to a fault, however, the acceptance of these techniques has become so widespread that one might argue many ground improvement projects would be better suited as traditional pile supported projects. More bluntly stated, inexperienced contractors are installing ground improvement on projects that should be supported on deep foundations. Of course, deep foundations have been around for centuries and are required where certain unsuitable soils, larger structure design loads, and allowable settlement criteria dictate a structural solution instead of ground improvement. Ground improvement is a terrific tool for SEs to use; however, we must all understand the limitations. Each project’s soil and loading conditions dictate the use of the correct technique.” Hayward Baker has recently acquired Geo-Instruments of Narragansett, Rhode Island, which allows HB to add robust geotechnical instrumentation and real-time monitoring services to their offerings. Says Hill: “Over the past decade, geotechnical instrumentation and monitoring systems have advanced to the point where STRUCTURE magazine

everyone on the project team can now be fully aware of project performance. In some cases, data can be obtained and plotted within minutes or even seconds of collection. Although there is an added cost to provide these cutting edge services, the added value outweighs its cost through the use of more efficient designs and quicker evaluation of the project’s geotechnical risk. Real-time monitoring can be so effective at mitigating or managing risks that projects once deemed un-constructible, or too risky for the owner, may now become viable.” (See ad on page 58.) StructurePoint (www.structurepoint.org), formerly the Engineering Software Group of the Portland Cement Association (PCA), continues to develop and enhance the full range of advanced PCA software and engineering services to model, analyze, and design reinforced concrete structures, according to Tim Schulz, the company’s Marketing Coordinator. “Our newly released spColumn v5.50 program features a dynamic spSection module to create and modify irregular sections, and a 3D failure surface viewer to investigate biaxial runs more thoroughly. Additionally, our spMats program now utilizes a sophisticated FEM solver to expedite solution of large-scale, complex models. Equipped with the latest American (ACI 318-14) and Canadian (CSA A23.314) concrete codes, StructurePoint software assists engineers to work quickly, simply, and accurately.” He says that now, more than ever, engineers are asked to meet the rising demand for new construction as infrastructure spending trends positively. “Clients not only use our continuously developing software, but they benefit from our engineers’ experience in the concrete industry to relieve heavy workloads, augment their staff, and advise on all of their concrete project needs,” says Schulz, who encourages engineers to visit their website. “Our website’s improved layout showcases design examples which aim to expedite the learning process for engineers who are new to reinforced concrete design. For experienced

June 2017

engineers, the design examples are packed with time-saving design aides, detailed code references, estimating guidelines and in-depth solutions for deflection calculations. “ Schulz says that StructurePoint continues to grow as they gain the acceptance and confidence of new users worldwide and retain the trust of established clients. “In a competitive engineering marketplace with tightening budgets and global competition, our focus remains on the improvement of our suite of concrete design software as well as the support and resources we provide. North American engineers are challenged to compete for global projects and to complete them more rapidly than ever before. In these situations, StructurePoint software is employed effectively to design quickly without compromising quality or exceeding costs.” Amy Heilig, CEO, U.S. Office of Dlubal (www.dlubal.com/en-US), would like SEs to know about a recent addition to their software line. “Design per the Aluminum Design Manual 2015 (ADM) was recently added to our nonlinear FEA software. RFEM is now complete with a wide array of multi-material design capabilities per U.S. and Canadian standards, including steel, reinforced concrete, timber, and aluminum. Also, we have designs for specialized niche markets such as glass, cross-laminated timber (CLT), and tensile fabric structures.” She notes: “RFEM continues to be the favorable choice for engineering projects which are not exactly in the ‘simple and easy’ classification. The CAD-like features allow complex element geometries to be modeled, loaded, and analyzed in a matter of minutes. For example, an ASCE 7 Response Spectra Analysis with

seismic equivalent loads is generated on each FE element independent of a structure’s size or shape. Material and geometric nonlinearities also give users the advantage over other software to narrow in on a structure’s complicated behavior.” Heilig says that, since establishing their Philadelphia office in 2015, they see an upward trend for business. “Our customer database has increased with new U.S. and Canadian engineering companies eager to utilize our design software. Our current customers also have continued to add to their existing licenses with additional seats and more design modules, indicating their satisfaction with RFEM and the technical support we have provided them. “ She understands that engineers are growing tired of using structural software that does not meet their demand or is too expensive to consider. “We pride ourselves on a product that was created by engineers for engineers,” says Heilig. “We can guarantee it’s one of the most user-friendly analysis and design program you will ever use. And considering the powerful capabilities RFEM has for contact surfaces, fabric form-finding ability, and material nonlinearities, to name a few, the software is still affordable. RFEM is the global solution for structural analysis, and we intend to continue development to keep up with new structural engineering demands.” To some in the industry for a long time, the idea of what constitutes a tall building has evolved. Geopier Foundation Company’s (www.geopier.com) Director of Business Development, Matt Caskey, says that the company was challenged and excited to work on a 16-story hospital tower in Atlanta in 1992. “Those of us in the ground

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STRUCTURE magazine

June 2017

improvement design-build industry may have a different definition of ‘tall building’ than most structural engineers, but, back in 1992, 16 stories seemed really tall. Before the mid-1990s, ground improvement systems were primarily used for small buildings on the worst-of-theworst sites or where the building design was severely impacted by seismic concerns. Structural engineers began to notice the successful performance of buildings supported on Geopier ground improvement systems and slowly developed confidence that Geopier systems could support the heavy loads associated with taller buildings. Our idea of a ‘tall’ building began to change.”

Now the company finds that, after years of demonstrated success with their ground improvement systems as well as consistent advancements to their technologies, it is supporting more tall buildings – even those on very poor sites. Says Caskey: “Geopier Rammed Aggregate Piers were installed to support the twin 18-story towers of the Marquee at Park Place condominiums in Irvine, California in 2004, where typical PGA values can range from 0.3 to 0.7g. The Geopier Impact technology was used for foundation support and liquefaction mitigation for the 20-story Edificio Parque Manuel Rodriguez building in Concepción, Chile, one of the most seismically active regions in the world. The soil conditions at this site consisted of layered deposits of loose sandy soil.” He notes: “A more recent Geopier rigid inclusion system, the Geopier GeoConcrete Column (GCC), has been used to support tall buildings on sites with deep deposits of very soft clay and peat. GCCs were used to support the 12-story Grand Condominiums in Cambridge, Ontario. The condominiums have column loads as high as 1,500 kips and mat pressures as high as 8,000 psf and were built on top of soft and organic soils to depths of 27 feet. Deep deposits of soft and organic soil also led structural designers to choose GeoConcrete Columns to support the 14-story College Avenue Office Building in New Haven, Connecticut. For this tall building, the GCC system eliminated the need for dewatering and spoils removal from the job site, and supported the foundation bearing pressure of 9,500 psf.” Concludes Caskey: “Developing efficient Geopier design solutions for taller buildings is always an exciting and challenging task for our engineers. The design process works best when the structural designers share load reactions and, sometimes, even help with finite element models to better understand pressure distributions on larger mat foundations. We are honored that the structural design community has confidence designing with GIVE YOUR STRUCTURE STABILIT Y Geopier ground improvement systems for their high-profile, ‘tall’ buildings, and we intend always to validate that confidence.” Work with Geopier’s geotechnical engineers to solve your ground When it comes to assessments, ICC improvement challenges. Submit your project specifications to Evaluation Service, LLC (ICC-ES) receive a customized feasibility assessment and preliminary cost (www.icc-es.org), is the nation’s most recestimate at geopier.com/feasibilityrequest. ognized and accepted product evaluation agency, says William Gould, Vice800-371-7470 President, External Relations and Client geopier.com Services. “We’re an ANSI accredited, a info@geopier.com non-profit subsidiary of the International Code Council (ICC) (www.iccsafe.org), the parent organization that develops the I-codes.”

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GEOPIER GROUND IMPROVEMENT CONTROLS STRUCTURE SETTLEMENT

continued on page 64

STRUCTURE magazine

June 2017

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“Over the past few months, we have issued many evaluation reports to manufacturers of innovative products for compliance with the requirements of the International Building Code (IBC),” says Gould. “We work with both domestic and international building product manufacturers and our service culminates in the publication of an evaluation report. This offers a huge benefit to product manufacturers seeking to grow their business since the ICC-ES Mark of Conformity is so well recognized by Building Officials and inspectors. ICC-ES evaluation reports are also a means of differentiation for manufacturers. Those who obtain them are typically the best manufacturers with the highest quality products. Our evaluation services remain the same, but the scope of products that we evaluate has continued to grow each year. Many new products with ICC-ES evaluation reports are used by SEs in the design of today’s complex structures and tall buildings. Knowing that they have ICC-ES evaluation reports provides peace of mind and assurance that the products meet stringent code requirements. Our product evaluations include examining product quality, strength, effectiveness, fire resistance, durability, and safety according to IBC Section 104.11.” ICC-ES recently issued an evaluation report to a Cross-Laminated Timber (CLT) manufacturer (Structurlam ESR-3631) for compliance of their mass timber product with ICC-ES AC455 Acceptance

Criteria for Cross-Laminated Timber Panels for Use as Components in Roof and Floor Decks. Says Gould: “CLT is being used for floor systems in a number of signature building projects throughout the U.S. and Europe. Although the IBC currently allows CLT for use in buildings up to six stories, it holds great promise for use in taller wood buildings. The ICC Ad Hoc Committee on Tall Wood Buildings is studying this topic and may propose future code changes pertaining to CLT. We have also issued many evaluation reports for wood fasteners and hangers under ICC-ES AC233 Acceptance Criteria for Alternate Dowel-Type Threaded Fasteners and ICC-ES AC13 Acceptance Criteria for Joist Hangers and Similar Devices that are used for attachments to and connections of wood members. Fastener and connector product development continue to be very dynamic as a result of cutting-edge technology, and manufacturers of these types of products are key players in the industry.” How’s business? “Business is strong and indicative of the construction market in general. More product manufacturers need evaluation reports and are applying to ICC-ES for them because they trust us. That’s a good thing for the industry,” says Gould. Business is picking up especially in high-rise buildings and infrastructure, according to MMFX Technologies Corporation’s (www.mmfx.com) President and CEO, Michael Pompay. “The

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STRUCTURE magazine

June 2017

construction trend of building higher with a narrow footprint to make the most of the space is driving demand for high-strength steel products. MMFX is on the forefront of this trend and has been pushing the advantages of using high-strength for years. We are now feeling a market pull due to the demand for high-strength applications and realization of the benefits derived from designing at 100 ksi. We also see a trend toward corrosion resistance. Designers are focused on the longevity of the structures more than ever. Owners, developers and departments of transportation are demanding better quality, longer lasting structures with reduced maintenance.” Pompay explains: “MMFX Steel Corporation offers a Grade 100 concrete reinforcing steel for use in high-rise buildings. ChrōmX 2100, sold under the specification ASTM A1035 CL, can be used at its 100 ksi yield strength to reduce congestion and improve design efficiencies of piles, columns, beams, slabs, mat foundations, and shear walls. “High-strength designs can improve constructability. As real estate becomes more precious and pricey, especially in densely populated cities, multi-use, high-rise buildings are being built taller and narrower than ever before. The resulting rebar congestion limits the constructability and height of these structures when applying lowergrade designs. Designs at 100 ksi, now available through the ACI and ICC design standards, reduce rebar congestion, allowing buildings to reach new heights,” he says. “High-strength, 100 ksi designs also generate construction efficiencies that save money and time. When properly utilized, high-strength designs result in substantial savings in materials, labor, and logistics. For instance, mat foundations are thinned, saving on excavation. Efficiencies are applied to add value to a building. For example, columns are narrowed, increasing usable office or living space.” ChrōmX 2100 is the latest addition to the MMFX high-strength concrete reinforcing product lines that provide over 100 ksi yield strength with varying corrosion resistant properties. “This allows designers to utilize the high-strength efficiencies and select the corrosion resistance that matches the environment and targeted service life of the structure,” Pompay says. “ChrōmX 2100 (ASTM A1035 CL) joins ChrōmX 9100 (ASTM A1035 CS) and ChrōmX 4100 (ASTM A1035 CM) to round out our product offering. ChrōmX 2100 is the lower corrosion-resistant of our products and provides the high-strength properties at the lowest cost, perfect for high-rise construction where corrosion is less of a concern.” John Cooper, District Manager, Complex Composite Group at Vulcraft/Verco Group (www.vulcraft.com), says that business continues to be strong. “Like everyone in the structural steel business, we will always face challenges. However, Vulcraft’s superior product offerings, coupled with a tireless desire to provide unmatched customer experience through every phase of a project, results in greater success for our clients, which means more jobs, more money, and a greater level of assurance for our customers.” The company has introduced the Vulcraft PunchLok II system, which utilizes the PunchLok II tool to achieve higher shear values at less cost through high-quality, side-seam attachments, says Cooper. “Used in conjunction with Vulcraft’s interlocking deck products, the pneumatic PunchLok II tool is a fast, effective, high-volume tool.” It is STRUCTURE magazine

manufactured by Verco Decking, Inc., a Nucor Company, exclusively for use with Vulcraft’s 1.5PLB and 3.0PLN, 1.5 PLVLI, 2.0PLVLI, and 3.0PLVLI steel deck. He adds: “Although the PunchLok II system is new to the Eastern and Midwest states, it has been widely used and proven on the West Coast by Verco Decking, Inc., since 2002. This system was created from Verco’s culture of innovation and a relentless drive to develop a solution with enough diaphragm shear capacity for the high seismic forces in the region.” Los Angeles requires about 1500plf diaphragm shear for a typical big box retail store. “The result,” says Cooper, “was the highest possible shear with the lowest possible cost using the PunchLok II system. The revolutionary high-performance steel deck diaphragm system is recognized by FM, UL, ICC-ES, and IAPMO.” According to Cooper, some advantages of the Vulcraft PunchLok II system are: • Requires no touch-up from either the top side or the bottom side of the deck; • Provides a completely weld-free system when used in conjunction with mechanical fasteners at supports; • 100% effective from the first attachment of the day to the last attachment of the day; • 100% accuracy of inspection of the attachment from the topside of the deck; • Stronger and more effective than conventional button punches; and • Stronger, more effective, and more economical than screwed side laps. As for trends, Cooper says that mechanical fasteners continue to be the deck installers’ method of choice to provide a required attachment pattern of the metal deck to meet project design loads. “The PunchLok II system works hand-in-hand with these fasteners; their allowable diaphragm shear strength values are published in our Design Guide.” (See ad on page 66.) “We are very excited about the latest S-FOUNDATION 2017 release that improves on the ability of structural engineers to easily generate foundations of any kind and shape (including piles), with any number of holes, pedestals, and walls,” says Marinos Stylianou, CEO of S-Frame Software (www.s-frame.com). “When engineers realize the simplicity and power of the modern user interface of S-FOUNDATION, they immediately understand the direct benefits to their organization. Also, recent release updates for S-FRAME Analysis, S-CONCRETE, S-STEEL, and S-CALC added improved functionality and integration.” As far as industry trends, Stylianou sees movement toward design solutions with strong visibility into the analytical solution. “Engineers want the solution quickly and easily, but they also want the ability to verify how the software arrived at the results. That is why S-Frame Software has long incorporated open-solution strategies in its products. Now engineers are looking for something that is one step beyond, a software solution that allows users to customize how the software behaves if the out-of-the-box solution doesn’t meet their needs. That is another major industry trend: customization. Our new generation of products incorporates the ability for users to customize and automate design processes. S-FRAME Software’s implementation of automation

June 2017

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is a first of its kind in this industry. It has taken our users a while to understand the power they now have at their fingertips. When they realize it, however, they immediately understand the direct benefits to their organization. They can customize the software so it follows their own workflow and then deploy it throughout their organization with the built-in library feature.” Stylianou adds: “We have noticed a definite increase in demand for structural engineering analysis and design software since the fourth quarter of 2016, and the trend has continued through the first quarter of 2017. This has been helped, in part, by increased demand from the oil and gas industry which is benefiting from higher oil prices. We’ve seen strong growth in the U.S. and Asia especially from companies that chose to use our advanced structural analysis and design platform for newer trend-setting tall buildings. Having software tools that are versatile enough to work for any industry type or region has been a key differentiator for us.”▪

Tall Buildings guide

JULY - Seismic/Wind AUGUST - Software NOVEMBER - Concrete For interview opportunities on these topics, contact advert@STRUCTUREmag.org.

expertise in tall building design and construction

American Wood Council

CAST CONNEX

Phone: 202-463-2766 Email: info@awc.org Web: www.awc.org/tallwood Product: Tall Wood Description: “Tall wood” defines mass timber buildings that exceed the height limit for wood buildings set by the current International Building Code. Mass timber includes any product currently permitted for use in Type IV construction such as cross laminated timber, structural composite lumber, glued-laminated timber, and large section sawn lumber.

Phone: 416-806-3521 Email: info@castconnex.com Web: www.castconnex.com Product: High Integrity Blocks™ Description: Ultra-heavy, weldable, solid steel sections which exhibit up to 65 ksi yield strength in all three directions of loading and through the full cross-section. Ideal for use within the center of multiaxis loaded connections or where the lamination of multiple steel plates is not advisable.

Phone: 800-929-3030 Email: jong@ctscement.com Web: www.ctscement.com Product: Rapid Set® Cement Products and Komponent® Type-K Shrinkage-Compensating Cement Description: Use Rapid Set cement products for concrete repairs, restoration, and new construction. Achieve high durability, fast strength gain, and structural or drive-on strength in one-hour. Install concrete structures and industrial-size floors using Type-K shrinkage-compensating cement products with no curling, no drying shrinkage cracking, and no intermediate saw cut joints.

Applied Science International, LLC

Concrete Masonry Association of California & Nevada

Dlubal Software, Inc.

Phone: 919-645-4090 Email: support@appliedscienceint.com Web: www.appliedscienceint.com Product: Extreme Loading for Structures Description: A new advanced level of nonlinear dynamic structural analysis; allows users to efficiently study structural failure from any number of actual or possible extreme events. Unlike traditional FEM software, users can easily model high-rise structures composed of reinforced concrete, steel composite, and other structures with as-built and as-damaged details.

CTS Cement Manufacturing Corporation

Phone: 916-722-1700 Email: info@cmacn.org Web: CMACN.org Product: CMD15 Design Tool for Masonry Description: Structural design of reinforced concrete and clay hollow unit masonry elements for design of masonry elements in accordance with provisions of Ch. 21 2010 through 2016 CBC or 2009 through 2015 IBC and 2008 through 2013 Building Code Requirements for Masonry Structures (TMS 402/ACI 530/ASCE 5).

Phone: 267-702-2815 Email: info-us@dlubal.com Web: www.dlubal.com Product: RFEM Description: Complete with USA/International Standards for steel, concrete, timber, CLT, glass, and aluminum, the user-friendly software allows for efficient modeling, a powerful non-linear analysis and highly detailed design results for all multimaterial tall buildings. Direct interfaces with BIM and CAD software incorporate seamless and bidirectional data exchange.

continued on next page

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STRUCTURE magazine

June 2017

Tall Buildings guide

expertise in tall building design and construction

Intergraph

Simpson Strong-Tie

StructurePoint

Phone: 281-890-4566 Email: query.icas@intergraph.com Web: www.coade.com/products/gtstrudl Product: GT STRUDL® Description: For more than 45 years, structural engineers have trusted GT STRUDL for their structural analysis. The software offers a comprehensive frame and finite element analysis solution for steel and concrete designs. Includes all the tools necessary to analyze a variety of structural engineering and finite element analysis problems.

Phone: 800-925-5099 Email: web@strongtie.com Web: www.strongtie.com Product: SC Bypass Framing Slide-Clip Connector Description: SC connectors are the optimal solution for slide-clip bypass framing, especially in high-seismic regions. They can be welded to the structure or attached with concrete screws or powder-actuated fasteners through anchorage holes in the clip. The clips are manufactured using heavy-duty steel to provide exceptional resistance to in-plane seismic load.

Phone: 847-966-4357 Email: info@structurepoint.org Web: www.structurepoint.org Product: spBundle Description: Recently upgraded to include ACI 318-14 and CSA A23.3-14, the remaining programs in our software bundle (spSlab, spBeam, spWall, spMats, spFrame) compliment spColumn for complete reinforced concrete building design including floor systems, beams, cast-in-place walls, tilt-up walls, commercial building foundations, pile caps, and slabs on grade.

Product: CFS Designer™ V2.0 Software Description: With the CFS Designer, you can design CFS beam-column members according to AISI specifications and analyze complex beam loading and span conditions. Intuitive design tools automate common CFS systems such as wall openings, shearwalls, floor joists, and, with the newest software update, up to eight stories of load-bearing studs.

Product: spColumn Description: Featuring a flexible graphical interface in the new spSection module for creating and modifying irregular sections, spColumn is used for design and investigation of columns, shear walls, bridge piers, and typical framing elements in buildings and other structures subject to combined axial and flexural loads.

Standards Design Group, Inc

Trimble

Phone: 800-366-5585 Email: info@standardsdesign.com Web: www.standardsdesign.com Product: Wind Loads on Structures 4 Description: Performs computations in ASCE 7-10, Chapters 26-31 and ASCE 7-98, 02, 05, Section 6. Computes wind loads by analytical method rather than the simplified method, provides basic wind speeds from a built-in version of the wind speed, allows the user to enter wind speed. WLS4 has numerous specialty calculators.

Phone: 770-426-5105 Email: kristine.plemmons@trimble.com Web: www.tekla.com Product: Tekla Structures Description: Move from design-oriented to construction-oriented engineering and enable structural engineers improved additional services. Through our open and collaborative software environment, you can work with other disciplines and reduce RFIs. From concept to completion, Tekla software gives you collaboration and control.

Product: Window Glass Design 5 Description: Performs all required calculations to design window glass according to ASTM E 1300-09. WGD5 also performs window glass design using ASTM E 1300 02/03/04, ASTM E 1300-98/00 and ASTM E 1300-94. GANA endorses WGD5 as best tool available in designing window glass to resist wind and long-term loadings.

Product: Tedds Description: A powerful software that will speed up daily structural and civil calculations, Tedds automates repetitive structural calculations. Perform 2D Frame analysis, utilize a large library of automated calculations to US codes, or write your own calculations while creating high quality and transparent documentation.

Strand7 Pty Ltd

WoodWorks® Software

Phone: 252-504-2282 Email: anne@beaufort-analysis.com Web: www.strand7.com Product: Strand7 Description: A general-purpose FEA system comprising integrated pre- and post-processing and solvers. Used for linear and nonlinear analysis of structures and components (static, dynamic and heat transfer). Strand7 has gained worldwide acceptance as a powerful tool for structural analysis, particularly nonlinear analysis.

Phone: 800-844-1275 Email: sales@woodworks-software.com Web: www.woodworks-software.com Product: WoodWorks® Design Office Description: Conforms to IBC 2015, ASCE710, NDS 2015, SDPWS 2015; SHEARWALLS: designs perforated and segmented shearwalls; generates loads; rigid and flexible diaphragm distribution methods. SIZER: designs beams, columns, studs, joists up to 6 stories; automatic load patterning. CONNECTIONS: Wood to: wood, steel, or concrete. Canadian version available.

PFS TECO Phone: 608-839-1013 Email: steve.winistorfer@pfsteco.com Web: www.pfsteco.com Product: Certification, inspection and Testing Services Description: Employee-owned, independent, third-party certification and testing agency for manufacturers of EWPs (including cross laminated timber), panels products, building components, and hearth products; and is an approved IPIA/DAPIA for HUD-code manufactured housing.

RISA Technologies Phone: 949-951-5815 Email: info@risa.com Web: www.risa.com Product: RISAFloor Description: RISAFloor and RISA-3D form an unrivaled building analysis and design package. Modeling has never been easier whether you’re doing a graphical layout, importing a BIM model (from Autodesk Revit Structure), or prefer spreadsheets. Full code checks and optimization for six different material types makes RISA your first choice in buildings.

S-FRAME Software Phone: 604-273-7737 Email: info@s-frame.com Web: www.s-frame.com Product: S-CONCRETE Description: For designing and detailing reinforcedconcrete column, beam, and wall sections. Optimize a single section or evaluate thousands of concrete sections at once. S-CONCRETE generates comprehensive reports that include clause references, equations employed, intermediate results, and diagrams. Latest release includes Eurocode 2 updates, ADAPT Builder Wall Design integration. Product: S-FOUNDATION Description: S-FOUNDATION Pro 2017 released! New pile features include soil definitions to calculate deflections, axial forces, compression, and tension along the depth of each pile. Quickly and easily define and detail pile supported foundations. Seamlessly transfer model data from S-FRAME Analysis or easily import support data from 3rd party analysis software.

All Resource Guide forms for the 2017 Editorial Calendar are now available at www.STRUCTUREmag.org. Listings are provided as a courtesy. STRUCTURE® magazine is not responsible for errors.

STRUCTURE magazine

June 2017

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Punching of Slabs A Discussion on the Development of the Delamination of Concrete Cover in the Soffit of the Slab By Aurelio Muttoni, Ph.D., Miguel Fernández Ruiz, Ph.D., and João T. Simões Aurelio Muttoni is a Professor at École Polytechnique Fédérale de Lausanne, Lausanne, Switzerland. He can be reached at aurelio. muttoni@epfl.ch. Miguel Fernández Ruiz is a Senior Lecturer at École Polytechnique Fédérale de Lausanne, Lausanne, Switzerland. He can be reached at miguel.fernandezruiz@epfl.ch. João T. Simões is a PhD student at École Polytechnique Fédérale de Lausanne, Lausanne, Switzerland. He can be reached at Joao.simoes@epfl.ch.

unching shear behavior is a topic that has attracted much attention from engineers in the last decades because of several collapses caused by punching shear failures. Introducing transverse reinforcement is the most common solution when the geometry of the slabcolumn connection has to be maintained and punching shear resistance has to be increased. To that aim, several transverse reinforcement systems may be used, not only to increase the punching shear resistance but also to significantly increase the deformation capacity and the residual strength after a local failure. If usual detailing rules are fulfilled, the design of slabs with shear reinforcement is governed by one of the following three potential failure modes: 1) crushing of the concrete struts in the column vicinity (maximum shear strength, see Figure 1a); 2) punching within the shear reinforcement (governing the dimensioning of the transverse reinforcement, Figure 1b); and 3) punching outside the shear-reinforced region (governing the dimensioning of the size of the zone with shear reinforcement), Figure 1c. Several experimental investigations of slabs with transverse reinforcement available in the literature have revealed the development of horizontal cracking at the height of the compression reinforcement, which can be seen as a delamination of the concrete cover in the soffit of the slab. Questions might arise whether its potential occurrence influences or even governs the punching shear behavior and whether this effect should be accounted for in the design. To investigate the influence of this phenomenon, the origin and the development of such cracking should be understood. According to the several test programs, the phenomena leading to cover delamination appear to be multiple. Several authors of experimental studies have mentioned that delamination of concrete cover in the soffit of the slab could be observed before failure, even in cases where failures were shown to be within the shear-reinforced region or by crushing of concrete struts. This has also been observed in

slabs with shear reinforcement recently tested by the authors and presenting a failure due to crushing of concrete struts (Figure 2). The origin of this delamination can be explained by the tangential strains developing at the column vicinity as a function of the rotation of the slab (Equation 1).

(a)

(b)

(c)

(d)

(Et=

Ψ ·c r

Equation 1

Where Ψ is the rotation of the slab, r the radial distance from the center of the column, and c the height of the compression zone. For large rotations, tangential strains in the vicinity of the column may largely exceed the peak uniaxial strain of concrete (more than 0.5% in many cases). The concrete cover, which is not confined by any reinforcement, enters a softening stage and a strain localization (horizontal cracking) occurs at the level of the compression reinforcement. This effect becomes more pronounced for large rotations, which are normally observed in slender slabs with shear reinforcement. The delamination of the concrete cover typically occurs in the critical shear region and reduces the lever arm in both radial and tangential directions. This phenomenon may constitute a limitation of the maximum punching shear strength. The impact of this effect on the efficiency of different transverse reinforcement systems deserves further research. Another common situation where delamination is observed occurs in a punching failure outside the shear-reinforced region. In this case, an inclined shear crack develops between the flexural reinforcement and the bottom of the last perimeter of transverse reinforcement (Figure 1c), together with horizontal cracking joining the bottom of the inclined shear crack and the column face. The Model Code for Concrete Structures 2010 (International Federation for Structural Concrete) already accounts for this phenomenon in the punching shear design for a failure outside the shear-reinforced area (considers a reduced effective depth). Delamination of the concrete cover under transverse reinforcement can also occur when

Figure 1. Different failure modes of flat slabs with shear reinforcement: a) crushing of concrete struts; b) failure within the shear-reinforced region; c) failure outside the shear-reinforced region; d) failure by delamination.

70 June 2017

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Figure 2. Specimen tested by the authors with failure due to crushing of concrete struts near the column; pronounced delamination observed in the soffit of the slab.

failure mode. In this case, the inclination of the failure surface in the areas between the transverse reinforcement rows is flatter than that developing in the regions of transverse reinforcement (Figure 3b). To deal with the development of non-uniform failure surface, codes normally limit both 1) the tangential distance between rows of transverse reinforcement at the location of the control perimeter, and 2) the maximal distance of straight lines of the outer control perimeter. Finally, when some important detailing rules are not respected, delamination may also occur (Figure 1d). This is the case of transverse reinforcement not embracing the flexural compression and tension reinforcement, which leads to a slab cover delamination and premature failure, although usually associated with large deformation capacities. In conclusion, delamination of the compression zone is mostly a matter of poor detailing: transverse reinforcement units too distant in the transversal direction, and cover of transverse reinforcement too large.▪

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Figure 3. Scheme of punching shear failures with non-uniform location of the failure surface: a) failure outside the shear-reinforced region; b) failure by concrete crushing.

STRUCTURE magazine

a large tangential distance between rows of transverse reinforcement is present in a radial and particularly in a cruciform arrangement. In these cases, inclined shear cracks might start developing first in between the rows of transverse reinforcement, followed by a tangential propagation until reaching the transverse reinforcement elements. If the amount of transverse reinforcement is low, a failure within the shear-reinforced area can occur. On the other hand, if a large amount of shear reinforcement is used, two situations can follow: If the last perimeter of shear reinforcement is not sufficiently distant from the column to avoid a punching failure outside the shearreinforced area, the inclined cracks already formed in between the rows of transverse reinforcement will tend to propagate tangentially through the formation of a separation crack (delamination) at the level of the compression reinforcement (without crossing the transverse reinforcement elements). This will be completed by inclined shear cracks around the rows of transverse of reinforcement (Figure 3a). If the last perimeter of shear reinforcement is sufficiently distant from the column, crushing of concrete struts will be the governing

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June 2017 CD17070_Clips_Structure_ThirdPg_VertAd_June17.indd 15/11/17 11:58 AM

Historic structures

significant structures of the past

Sciotoville Bridge

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng., P.E., P.L.S.

he Sciotoville Bridge, designed by Gustav Lindenthal (STRUCTURE, August 2010) over the Ohio River, is located about 90 miles upstream from Cincinnati. The Chesapeake and Ohio Railroad, in order to make Chicago accessible for its coal-carrying trains out of Kentucky, built a 30-mile line from its tracks in Edgington, Kentucky northerly across the Ohio River to Waverly where it connected with a line to Columbus, Ohio and then on to Chicago. The Ohio was about 1,500 feet across at the selected site, with bedrock only 10 feet or so below low water. The Ohio had been bridged for railroads several times by Jacob Hays Linville and others, starting with the Steubenville, Parkersburg, and Benville Bridges upstream and downstream at Cincinnati. The railroad was looking for a twin track structure that would meet demands of the War Department to provide for passage of steamboats. Lindenthal, as was his custom, looked at all types of bridges that had been used to carry heavy railroad loadings over long spans. He determined that a continuous truss with two spans of 775 feet best met the site conditions. The fact that rock was accessible at a shallow depth removed one of the factors that ruled against continuous spans in the past – settlements of the central piers could be eliminated. The other reason continuous spans were not adopted, with notable exceptions of C. Shaler Smith’s Lachine Rapids Bridge (STRUCTURE, April 2017) across the St. Lawrence and Lindenthal’s two smaller, shorter span continuous spans in the Pittsburgh area in 1883 and 1890, was the difficulty in making calculations for an indeterminate structure. He indicated that temperature effects were minimized in his design with a fixed support at the river pier and rollers at the abutments. With the help of O. H. Ammann and D. B. Steinman, Lindenthal designed the longest continuous span in the United States. The main advantages of the continuous structure were savings in material and its ability to be constructed by cantilever methods. He estimated savings in the material of nearly 25%, with an increased rigidity when compared to a cantilever bridge and an equal rigidity when compared to the simple span truss. He also stated, “From the esthetic point of view, the continuous bridge can well compete with the simple span or cantilever, if properly designed, but not with the more artistic arch or suspension bridge.”

Lindenthal incorporated a unique inverted two-hinged arch, building the floor beams into the verticals, with the thrust at the upper ends being taken by the upper lateral bracing This design allowed for smaller floor beams to span the distance between trusses, which were 38 feet 9 inches center-to-center. First, he designed trusses with 16-inch eye bars for all tension members that would not see a reversal of stress under live load. He then designed all members built up of riveted members and fully riveted connections and requested bids on both designs. The cost of steel came in lower with the eye bars, but he chose the fully riveted structure “in view of its superior rigidity, durability, and safety.” Moreover, finally, Lindenthal calculated the secondary stresses that would occur under normal erection procedures and found them very high. He wrote, “on account of the rigid truss connections; it was considered advisable to reduce the secondary stresses as far as possible, not only near the center support but throughout the truss, in the chords as well as in the web members.” This was done by cambering the trusses for full dead load plus one-half the live load, covering both spans, but assembling and erecting them so that the angles between the members and the bevels of the joints would correspond to the geometric form of truss. In other words, under the load (d+ ½ l), the trusses are calculated “to assume their true geometric form and the members to become straight and free of secondary stresses.” He determined his “secondary stresses under dead load only are equal, but of opposite sign, to those under full live loads covering both spans, and, in absolute value, equal to one-half of those which would be produced by full live load if the angles between the members would correspond to the cambered form of truss.” This cambering of the trusses required that members be preloaded to allow connections to be made that resulted in the cambering he needed. This was the first time that this expedient was utilized in American bridge building. Building on temporary falsework by cantilever methods also resulted in many different loading conditions of his members during erection and after completion. As would be expected, he used all of these various loading conditions to determine the maximum load that would fall on each member and designed the member to resist that load. For his erection process, he built the Ohio side of the bridge on false work but could not use as much falsework on the Kentucky side.

STRUCTURE magazine

June 2017

He, in association with McClintic Marshall, the contractor, built the Ohio side on falsework and then cantilevered out to the eighth panel point from the Kentucky side and placed a steel bent in the river. He then jacked up Rigid frame tying floor the span a desired beams into truss verticals. (calculated) amount and continued cantilevering out to the fourth panel point from Kentucky and jacked up the span once again. He completed the span by cantilevering out to the Kentucky abutment. When the span reached panel point, Lo, the truss was 16¼ inches low at which time the end was jacked up and set on the rocker bearings. The superstructure was finished on August 17, 1917. The Engineering Record published an article on the bridge and in an editorial wrote: “In the final selection of type of structure, the particular nature of the profile at the crossing, the clear opening required during erection and the nature of the foundations at possible locations of piers or abutments are all determining factors. The advantages in erection and the fact that solid rock foundations were found at small depths so that the settlement at the piers should be practically negligible, were the most important considerations in the selection of the continuous [structure] type in this case. Economy for this type can generally be easily shown…The indeterminate character of the stress analysis is also generally considered a disadvantage. The experience with the Queensborough Bridge may be cited in this connection. The continuous span on many supports was shown by the report of the investigation engineers to be designed with astonishing errors in the distribution of the metal…It is notable, however, that methods of treating indeterminate structures are now becoming well standardized and should soon be simplified into such form as to make the cost of designing low.” Lindenthal did not describe the bridge in the Transactions of ASCE until 1922 when he wrote a 43-page article entitled, “The Continuous Truss Bridge over the Ohio River, at Sciotoville, Ohio of the Chesapeake and Ohio Northern Railroad.” Engineering and Contracting ran a

Sciotoville truss work and falsework supports.

long article based on the ASCE piece in its April 22, 1922, issue. After giving a brief history of continuous bridges in Europe and the United States, Lindenthal challenged the reluctance of his fellow engineers to utilize the continuous truss writing, “The continuous truss type has nowhere met with more indiscriminate and unqualified condemnation than by engineers in the United States who have alleged three principal objections against it, none of which is novel or decisive: 1st – That it is statically indeterminate, that is, its reactions and stresses are dependent on the elasticity of its members; 2nd – That it is subject to stresses from unequal settlements of its supports; and, 3rd – That it is affected by temperature changes.” He then made his case as to why none of these objections apply to his Sciotoville Bridge and proceeded to describe his structure as follows. General Proportions – The steel superstructure, as built, is a double-track, two-span, continuous bridge, with a total length of 1,550 feet between centers of end piers, or two equal spans of 775 feet and two clear openings of 750 feet at lowwater level. To obtain ample lateral rigidity, the width between centers of trusses was [a] somewhat smaller width than was considered sufficient, in view of the continuity of the lateral truss. The height at the middle of each span is 103 feet 4 inches between centers of chords. To fix this height, the portion of the span between the end pier and the point of contra flexure was considered as a simple span. This portion varies from three-fourths of the span length for uniform load on both spans to seven-eighths of the span length for uniform load on one span only and averages 630 feet in length. The height

of the truss was chosen approximately one-sixth of this length. The height over the center pier is 129 feet 2 inches or one-sixth of the span length; this is the proper height for a simple span of the same length, which has the same maximum moment at the center as the two-span continuous bridge over the middle support. The height at the end was made 77 feet 6 inches, or equal to a double panel, in order to give the end posts an inclination of not less than 45 degrees. These heights also secured a pleasing outline for the top chord. The web system is of the Warren type, with subdivided panels of a uniform length of 38 feet 9 inches, which was found to be the most economical. The discussion of the article ran 21-pages. C. A. P. Turner wrote a seven-page response whose main point was that two simple spans would have been cheaper and concluded that Lindenthal’s “preference for the continuous bridge of moderate span differs from the majority opinion of American bridge engineers, because of lack of demonstrated economy on a scientific, mathematical or design basis.” J. E. Greiner wrote praising Lindenthal, “It may be said also that, in every structure of importance designed by Mr. Lindenthal, there is evidence of this genius which originates. Each of his structures is practically a new creation as compared with the ordinary and stereotyped bridges throughout the United States. The Sciotoville Bridge is no exception. It is a daring and handsome structure, decidedly, ‘Lindenthalic’ in all its features…” Charles Fowler wrote, “The Sciotoville Bridge is a striking example as to what may be accomplished in the use of the continuous bridge. It is the longest of that type ever constructed and now gives America the proud distinction of having the longest spans for every type of bridge construction.” Lindenthal responded to the individual commenters, taking exception to Turner’s claim that two simple spans would have been less expensive. As to J. E. Greiner’s discussion in which he argued that Lindenthal had not designed the bridge to Cooper’s E60 loading, Lindenthal wrote that Greiner, “was unjustified and made

Sciotoville Bridge (color postcard author’s collection).

STRUCTURE magazine

June 2017

without examination of the requirements of the writer’s specifications other than the basic unit stress (20,000), which may mislead superficial readers.” He commended Steinman’s contribution but disagreed with Steinman’s remark that reducing the height of the trusses over the middle pier and increasing the height at midspan “would not be on the whole, advantageous or economical.” Steinman had written, “The Sciotoville Bridge is a striking example of scientific design. It represents an unusually intensive application of engineering theory and resourcefulness to the determination of the most efficient disposition of the materials of a structure, and to the rigorous advance planning of every step in the operations of fabrication and erection.” Steinman had also written an article for the Engineering Record on August 28, 1915, entitled “The Elastic Curve Applied to the Design of the Sciotoville Bridge.” It is evident from the record that most bridge builders in the United States were still not ready to adopt continuous bridges and preferred long simple truss spans or cantilevers. Shortly after the bridge was opened, the Bessemer and Lake Erie Railroad built a continuous bridge over the Allegheny River near Pittsburgh with spans of 272 to 520 feet. The Hudson Bay Railway built a bridge over the Nelson River in 1918 with spans of 300 to 400 feet. When opened, the Sciotoville Bridge’s 775foot riveted spans were the longest continuous truss spans in the world until the opening of the Duisburg-Rheinhausen Bridge in Germany in 1945 with its 835-foot spans. With its rigid construction, Lindenthal’s continuous truss bridge serves the Chesapeake and Ohio Railroad to this day.▪ Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He was formerly Director of Historic Bridge Programs for Clough, Harbour & Associates LLP in Albany, NY, and is now an Independent Consulting Engineer. Dr. Griggs can be reached at fgriggsjr@verizon.net.

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discussion of legal issues of interest to structural engineers

LegaL PersPectives

An Overview of Consent to Assignment Agreements By Gail S. Kelley, P.E., Esq.

esign agreements often stipulate and _____ as agent for the Lenders occurred under the Assignment, it will perthat neither the owner nor the (together with its successors in such capacform all of its obligations under the Design engineer can assign the agreeity, the “Agent”). Agreement for the benefit of Lender. ment without the consent of Alternatively, it may be structured as an Regardless of how the consent is structhe other party. As a result, the engi- agreement between the owner and the tured, the intent is to ensure that if the neer may be asked to sign a “Consent to engineer, with the lender as a third-party owner defaults on its loan and the lender Assignment” (sometimes referred to as an beneficiary. takes over, the engineer will continue to “Acknowledgement and Consent”) from the This CONSENT TO ASSIGNMENT of provide the services called for under the bank providing the construction loan. A Engineer’s Contract (this “Agreement”) is design agreement if requested to do so. The typical consent form requires the engineer made as of _____ by and between _____ agreement typically also gives the lender to agree that the design agreement can be (“Borrower”) and _____ (“Engineer”) for the right to use the engineer’s Instruments assigned to the lender. The assignment will the benefit of _____ (“Lender”). of Services to complete the project with actually occur at the time the loan is closed; It may also be structured as an agreement another engineer. however, the assignment is conditional in between the engineer and the lender: the sense that the lender can only assume This CONSENT TO ASSIGNMENT the design agreement if the owner defaults (this “Consent”) is dated as of _____ by Payment of Outstanding on the loan. _____ (“Engineer”) to _____ (“Lender”). Obligations If the owner defaults and the lender takes Finally, although less commonly, it may over, the lender’s chances of finding a buyer simply be written as what it is, which is a If the owner has defaulted on its loan, it is for the project are significantly better if the one-sided agreement under which the engi- likely to be behind in its payments to the buyer has the option of assuming all of the neer agrees to do, or not do, certain things: engineer. Thus, a key issue is the lender’s key contracts for the project. Thus the The undersigned, as Engineer under the obligation with respect to outstanding lender may ask the contractor, the engineer, Design Agreement dated as of _____ (the amounts due to the engineer. and the other key design consultants to “Design Agreement”) between _____ It is not uncommon to see consents with consent to an assignment of their contracts. (“Borrower”) and the undersigned, which the following wording: While Consent to Assignment requests is one of the contracts referred to in the Upon a Default under the Loan Agreement, are common, particularly on large projects, Assignment of Agreements (“Assignment”) Engineer, at Lender’s request, shall conthere is no standard form. The consent between Borrower and _____(the tinue performance on Lender’s behalf in should be read carefully and preferably “Lender”), agrees that upon receipt of accordance with the terms of Engineer’s reviewed by the engineer’s legal counsel. notice from Lender that a Default has Contract, and shall be reimbursed in Although the lender cannot assume the design agreement unless the owner defaults, the forms used by many banks contain provisions completely unrelated to the design agreement. Engineers who are not careful may find • Full line of high-strength, corrosion-resistant fasteners that they have given up valuable rights • Ideal for secondary steel connections and in-plant equipment or agreed to unreasonable obligations • Easy to install or adjust on site even if there is no default by the owner.

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There are several different ways to structure a consent agreement. Sometimes it is structured as an agreement between the owner (the borrower), the engineer, and the lender, with all three parties signing. For example: This CONSENT TO ASSIGNMENT (this “Agreement”), dated as of _____, by and among _____, (the “Borrower”), _____ (the “Engineer”),

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General Structure of a Consent to Assignment

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accordance with the Contract for all work, labor, and materials rendered on Lender’s behalf subsequent to Lender’s request. Note that the words in bold print will almost certainly not be in bold in the actual consent. What they are saying is that if the lender exercises its rights under the Assignment and requires the engineer to continue performance, the lender has no obligation to pay outstanding amounts due to the engineer. The lender’s obligation is limited to compensation for services provided after the lender notifies the engineer that it wants the engineer to continue performance. The engineer would have to file a mechanic’s lien against the property for the outstanding amounts. Depending on the priority of its lien, it might receive only a fraction of these amounts, if anything. The engineer should disagree with these terms and change the provision to read: ...provided that Engineer shall be reimbursed in accordance with the Contract for all work, labor, and materials including all outstanding amounts due. The lender may object to these terms on the grounds that if it has already advanced

funds to the owner for the engineer’s services, it should not have to pay the engineer for those same services. The lender may propose the following as an alternate: ...provided that Engineer shall be reimbursed in accordance with the Contract for all work, labor, and materials including all outstanding amounts due unless Lender has already advanced such funds to the Borrower. However, the engineer has no control over the lender’s disbursements; it is the lender’s obligation to monitor the loan. If the lender does not agree to pay all outstanding amounts due, the engineer should, at a minimum, require the following language: ...including all outstanding amounts due unless Lender had already advanced such funds to the Borrower prior to receipt of Engineer’s notice of Borrower’s default under Engineer’s contract. Lender shall not advance any funds to Borrower for Engineer’s services subsequent to receiving such notice. This puts the burden on the engineer to pay close attention to its payments. If a payment is late and the owner does not provide adequate assurance that the payment will be

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made, the engineer should advise the owner that it will notify the lender of the default.

Conclusion Typically, the request to sign a Consent to Assignment comes when the owner is arranging the construction financing, which may be months after the design agreement was executed. More often than not, it will indicate that the consent needs to be returned immediately because the owner is trying to close on its loan. This can put the engineer in a difficult position if there is objectionable language, particularly if the lender is not willing to negotiate. While the engineer generally has no obligation to sign a consent, refusing to do so can affect its relationship with the owner and jeopardize the prospect of future work. If the owner is already behind on payments to the engineer, it may say that it will not be able to pay until it closes on the loan. If the lender does not negotiate the terms, the engineer must make a business decision with respect to signing the consent. As a practical matter, unless the project is fasttrack, if the owner runs into financial trouble during construction and defaults on its loan, the engineer will likely have finished the plans and specifications and received payment for them. If the engineer is only doing limited construction administration, it may not have a significant risk with respect to payment. However, unless the consent indemnifies the engineer for the lender or subsequent buyer’s use of the plans and specifications, the engineer may still be at risk for claims.▪ Disclaimer: The information in this article is for educational purposes only and is not legal advice. Readers should not act or refrain from acting based on this article without seeking appropriate legal or other professional advice as to their particular circumstances. Gail S. Kelley is a LEED AP as well as a professional engineer and licensed attorney in Maryland and the District of Columbia. Her practice focuses on reviewing and negotiating design agreements for architects and engineers. She is the author of “Construction Law: An Introduction for Engineers, Architects, and Contractors,” published by Wiley & Sons. Ms. Kelley can be reached at Gail.Kelley.Esq@gmail.com.

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June 2017

2017 CASE RISK MANAGEMENT SEMINAR Time-Tested Techniques for Managing Your Firm’s Risk Scheduled for August 3-4, 2017 in Chicago, Time-Tested Techniques for Managing Your Firm’s Risk will help your firm reduce its rate of claims against structural engineering projects, as well as raise the level of quality services provided by all project participants. Who should attend? Principals, Owners, Project Managers, Risk Managers!

Session Topics • Strategies Forensic Engineers Use to Unravel Construction Disputes • Information Security in Contracting • Case Studies for Professional Liability A special dinner presentation on “The Future of Structural Engineering” by Ashraf Habibullah, President and CEO of Computers & Structures Inc. and structural engineering roundtable! Register at: www.acec.org/coalitions/upcoming-coalition-events/ >> $399/Coalition Members >> $499/ACEC Members >> $599/Non Members LOCATION All sessions will take place right off Chicago’s infamous Magnificent Mile at the DoubleTree Magnificent Mile, located at 300 East Ohio Street Chicago, Illinois.

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award winners and outstanding projects

Spotlight

Madison Square Park Tower 45 East 22nd Street By Joseph Savalli, P.E., Matthieu Peuler, P.E., and Leslie Morris, P.E. DeSimone Consulting Engineers was an Award Winner for its 45 East 22nd Street project in the 2016 NCSEA Annual Excellence in Structural Engineering Awards Program in the Category – New Buildings $30M to $100M.

n one of the few Manhattan neighborhoods not entirely punctuated by mirror-clad high-rises, Madison Square Park Tower blends turn of the century design with modern composition and accomplishes the tallest development between Midtown, Manhattan, and the Financial District. Located at 45 East 22nd Street and designed by architects Kohn Pedersen Fox, the tower topped out at 777 feet comprised of 65 residential and amenity levels. However, unlike any other development in the area, the tower’s floor plate progressively expands as it rises to maximize saleable area on the building’s most premium levels. The design strategy introduced complex geometric constraints that required close collaboration among all design consultants and construction team members. Overall, the design and structural components from ground-floor to pinnacle were impacted by the historical significance of the Flatiron neighborhood and tight constraints of the build site. At ground level, the tower rises from a 75-foot wide site wedged between two historic structures on each side. The design team opted for a granite-clad podium, rather than a topto-bottom glass façade, to complement the neighborhood’s Chicago School architecture dating back to the early 1900s. The tower is sculpted so that the floor plate is as small as 62 feet wide by 52 feet deep near the base, producing a maximum slenderness ratio of about 13 to 1. Above the 5th floor, the structure transitions into a more modern, glass façade and then cantilevers westward above its low-rise neighbor as it rises to a maximum floor plate of 94 feet wide by 52 feet deep. The cantilever allows the tower to expand to a maximum width of 125 feet at the top creating a flared, champagne flute silhouette. The cantilever also ensures that every floor above the sixth level is a unique shape. The gravity system is comprised of flat slabs spanning from the interior shear wall core to perimeter columns with varying cantilevers, a result of the tower’s complex geometry. Floor plates vary from 10 to 12 inches thick at residential levels as the spans lengthen with increasing

height. Thicker slabs are used at mechanical areas, including a 20-inch slab at the west side of the roof to support a 1.2 million-pound tuned mass damper system. The lateral system is comprised of one full-height shear wall core centered on the south side of the building, with wall thicknesses ranging from 42 inches at the base to 24 inches at the roof. The core is connected to perimeter columns at the 33rd floor mechanical level with a 1-story outrigger/belt wall. In addition to the unique shape of each plate above the sixth floor, column-free interiors were required. A two-way flat plate system was implemented to accommodate these specifications. The system provides flexibility in locating columns, including sloped and walked columns, while providing maximum ceiling height for floor-to-ceiling glass walls. The design of the lateral system was the greatest challenge of the development. The system required a robust design to resist required forces, including strict drift and acceleration limits and the inherent gravity overturn resulting from the top-heavy design. The selected lateral system is comprised of high-strength concrete shear walls, measuring 14,000 psi at the base down to 8,000 psi at the roof, which are coupled to perimeter columns at mid-height of the building with the belt wall. The design of this system was accompanied by wind tunnel shaping studies to determine the optimum configurations. While the concrete system alone meets the drift requirements, a 600-ton tuned mass damper at the roof level was implemented to reduce building accelerations to acceptable levels. The building is supported on a 50-foot wide by 80-foot long mat slab that is 8 feet thick. The mat bears on 20 tsf bedrock and includes 32,270-ton rock anchor tie-downs to resist overturning. The mat is reinforced with four layers of rebar top and bottom and local shear reinforcement as required. Construction below the 11th floor included numerous structural transfer elements as the tower reduces to its minimum footprint at the 7th floor. From the 7th to 11th floor, the west side of the tower cantilevers 16.5 feet out over an adjacent building. The system is designed as

STRUCTURE magazine

June 2017

a multi-story bracket braced to the core with the base of the bracket (compression) reacting through 8-foot wide by 2-foot deep bracing beams, and the top of the bracket reacting (tension) through groups of #14 GR97 SAS threaded bars. Above the 11th floor, the building was constructed on a fast-track, two-day pour cycle, alternating the pour of the vertical concrete elements and the horizontal concrete elements. This allowed for a shortened construction schedule that wouldn’t be possible with other structural systems. Throughout the project, GR75 vertical reinforcement was used in the shear core and columns to reduce reinforcement congestion. The belt wall at the 33rd-floor mechanical level was constructed with 2-foot thick perimeter walls that couple to the core of the perimeter columns. Large openings for mechanical louvers were accommodated with careful analysis and added reinforcement. The tuned mass damper is located on the west side of the roof level with its weight bearing half on the core and half on the slab. A system of a 20-inch slab with 48-inch wide by 40-inch deep beams was utilized at the slab support to provide adequate strength for the 1.2 million-pound steel damper assembly.▪ Joseph Savalli is a Principal in DeSimone’s New York office. Matt Peuler is a Senior Project Manager at DeSimone’s New York office. Leslie Morris is a Senior Project Engineer at DeSimone’s New York office.

Committee Update

NCSEA News

News form the National Council of Structural Engineers Associations

Basic Education Committee Works to Better Curriculum & Profession The NCSEA Basic Education Committee (BEC) has pursued activities which will help define the recommended curriculum for the Structural Engineering profession; these activities exist independently, but are intertwined in our mission to better educate the future of the profession. The committee is continually assessing the fundamental education recommended for entry into the structural engineering profession. In late 2016, the committee completed the Structural Engineering Curriculum Practitioner’s Survey of the NCSEA membership to determine which topics are relevant to the profession. The survey asked respondents to rate the importance of material design classes, analysis coursework, plus allied professional skills such as: communication, sustainability, architecture, and construction management. The committee is interpreting the data and evaluating trends related to wood and masonry design, computer software use and its relationship to structural analysis, similarities or differences between newly hired and experienced engineers, regional hiring preferences, and office size. It is anticipated that the findings, which will be distributed later this year, will help inform the committee about the recommended NCSEA basic education curriculum. The Structural Engineering Curriculum Survey is conducted to better understand the course requirements and class availability within accredited civil engineering, architectural engineering, engineering technology, and structural engineering programs. The results are compiled triennially from universities around the country and then published in STRUCTURE magazine. The last survey results may be accessed following this link: http://bit.ly/2rmPxkl. In past years, the committee has observed trends when identifying which institutions fulfilled the recommended curriculum and those which only fulfilled part. Those that met part of the curriculum typically lacked the timber or masonry design component; the significance of this has been debated at the committee level and led to NCSEA involvement in the Wood Education Symposium sponsored by the American Wood Council (AWC)/American Society of Civil Engineers (ASCE) – Structural Engineering Institute Wood Education Committee. The next education survey will be conducted in 2019 and will reflect results of the practitioner’s survey as well as take into consideration the ASCE-SEI/AWC Wood Education Symposium when forming the recommended curriculum and survey. The Wood Education Symposium was held on April 5th, 2017 in Denver, CO. The event was held to discuss the importance of wood education at the university level; what should educators,

design professionals, and industry manufacturers provide to meet the needs of stakeholders, and how can wood education be expanded to a wider audience. The symposium attendees included practitioners from engineering and architecture, academics, industry consultants, and grant advisors from government entities. NCSEA and the Basic Education Committee sent five attendees on their behalf: two representatives from SEAs and three BEC members. The perspectives were broad, but the group identified changes which are common across the industry, not just those pertaining to designing and building with wood. The key elements identified and which align with BEC objectives are: teaching building systems in context versus component or elemental design – change the way in which wood or material design is taught; exposure to connections and load path at the detail level; use real world problems so students understand the context; the need to produce thinkers and problem solvers, not just graduates with analytical skills; and material behavior – wood is not isotropic, but teaching material behavior should be a requirement for all design materials. Interestingly, about 40 percent of the attendees in the design professional subgroup had no formal timber education, yet, their career paths and company interests required some wood design knowledge so they were self-taught. Thus, the critical question arose, “should timber design be taught in the classroom?” Additionally, there was no clear consensus when the subgroup was asked about the importance of wood design experience as a factor in hiring. It was evident that wood education is not critical to companies that do not design wood structures and, conversely, those that work with wood structures consider wood design a requirement. So another question was posed, “is the demand for structural engineers with wood design experience being met?” And if the answer is “yes,” then does it need to be taught at every university with a civil or structural engineering program? Questions such as these and many more will be evaluated over the next year. As the BEC interprets the results from the practitioner survey and assesses the implications of the Wood Education Symposium, it will determine if the coursework recommendations should be modified and proposed for the 2019 survey. This information was presented during an NCSEA MO Communication webinar on May 26, 2017 to showcase this material in more depth and to engage the NCSEA community in additional dialog. A recording of this webinar can be found in the NCSEA Member Portal on the MO Communication Meeting Resources page. Professor Kevin Dong, California State Polytechnic University, NCSEA Basic Education Committee Co-Chair.

Designed by Structural Engineers for Practicing Structural Engineers NCSEA 2017 Structural Engineering Summit October 11–October 14, 2017 Washington Hilton, Washington, DC Register now for NCSEA’s early bird rate and save $100 on your full conference registration! Visit www.ncsea.com for more information, schedules and hotel reservations. STRUCTURE magazine

June 2017

June 15, 2017 Seismic Design of Large Wood Roof Diaphragms in Heavy Wall Buildings John Lawson, S.E. June 27, 2017 Special Inspections for Existing Buildings Chris Kimball, S.E., P.E. July 11, 2017 Repair of Construction Defects David Flax July 25, 2017 Nonstructural Components Chris Kimball, SE, P.E.

EXCELLENCE IN STRUCTURAL ENGINEERING AWARDS

The NCSEA Excellence in Structural Engineering Awards annually highlights some of the best examples of structural engineering ingenuity throughout the world. Structural engineers and structural engineering firms are encouraged to enter this year’s program. Projects will be judged on innovative design, engineering achievement and creativity. Awards can be entered in one of seven categories: • New Buildings under $20M • New Buildings $20 Million to $100M • New Buildings over $100M • New Bridge & Transportation Structures • Forensic/Renovation/Retrofit/Rehabilitation Structures up to $20M • Forensic/Renovation/Retrofit/Rehabilitation Structures over $20M • Other Structures Awards will be presented during an honorary banquet at the NCSEA Structural Engineering Summit in Washington, D.C. on October 13th. This awards banquet is the premiere event of the Summit and illustrates the importance of artistry and inspiration our structural engineers and structural engineering firms provide to the association and to the world. Eligible projects must be substantially complete between January 1, 2014 and June 30, 2017. Entries are due Tuesday, July 18, 2017. For award program rules, project eligibility and entry forms, visit www.ncsea.com.

STRUCTURE magazine

June 2017

Two exclusive annual plans are available to NCSEA corporate members & SEA members only. The Live & Recorded Webinar Subscription Plan with access to all live webinars and the entire recorded webinar library, hosting over 180 webinars, or the Live Webinar Subscription Plan. Visit www.ncsea.com to purchase your subscription today!

Start Studying Now for the Fall Exam! Over 29 hours of recorded SE Review & Refresher sessions are available through NCSEA’s online store. Conveniently accessible through the Recorded Webinar Library, you can view these recordings all day everyday, 24/7 until the next exam!

Visit www.ncsea.com to get started now! Dates for the next live course will be announced shortly.

News from the National Council of Structural Engineers Associations

20 17

Visit www.ncsea.com to register and read the full description of each webinar. 1.5 hours of continuing education. Approved for CE credit in all 50 states.

NCSEA News

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The Newsletter of the Structural Engineering Institute of ASCE

Structural Columns

Electrical Transmission & Substation Structures Conference 2018

Top 5 Reasons to Attend 1) Expand your knowledge at technical sessions on transmission line and substation structures and foundations. 2) Earn professional development hours (PDH’s) by attending technical sessions and workshops. 3) Network with global leaders and colleagues working with high-voltage transmission structures around the world. 4) Connect with exhibitors showcasing state-of-the-art products, services, and solutions for your transmission line and substation projects. 5) Discover Southern hospitality and enjoy over 100+ live entertainment venues.

Call for Abstracts and Sessions

The State-of-the-Industry Forum for Transmission and Substation Engineers: • Discover Technical Knowledge • Hear Project Case Studies • Find Real-World Solutions • Visit Vendors and Learn about their Products and Services The SEI/ASCE Electrical Transmission & Substation Structures Conference is recognized as the must-attend conference that focuses specifically on transmission line and substation structure and foundation construction issues. This event – for utilities, suppliers, contractors, and consultants – offers an ideal setting for learning and networking.

Exhibits & Sponsorships Increase your company’s visibility and reach hundreds of industry professionals at this important specialty conference. Contact Bob Nickerson at renicker@flash.net or 817-319-8779, or Sean Scully at sscully@asce.org or 703-295-6154, for exhibiting and sponsorship opportunities. Questions? Contact Debbie Smith dsmith@asce.org or 703-295-6095. Submit your sessions at www.etsconference.org.

ELECTRICAL TRANSMISSION & SUBSTATION STRUCTURES CONFERENCE 2018

Dedicated to Strengthening

Atlanta, Georgia November 4–8

Abstracts & Session Proposals

our Critical Infrastructure due September 12, 2017

Dedicated to Strengthening our Critical Infrastructure

2017 ASCE Annual Convention

The American Society of Civil Engineers invites you to attend the ASCE 2017 Convention, October 8 – 11, in New Orleans. Among the broad range of activities planned, the Society’s flagship gathering will take advantage of the location to explore how the city and its infrastructure have rebounded since 2005. Three reasons to attend the ASCE 2017 Convention in New Orleans: • Learn how New Orleans and other major cities are using resiliency and recovery to strengthen social justice and community life. • Hear first-hand accounts from experts on natural disaster response and recovery. • Gumbo. It’s fun to say and great to eat. Check the Convention website at www.asceconvention.org frequently, as more information is available. Don’t miss your chance to make valuable connections with civil engineering professionals globally. We look forward to seeing you there.

Structures Congress 2017 Best of the Best Congratulations to the winner of the Best Presentation prize at the 2017 Structures Congress. Congress attendees selected Sharing the Story of a Real Claim: The Condo, presented by Roger Heeringa, P.E., S.E., for this annual honor. The Best Presentation winner will receive a complimentary full registration to Structures Congress 2018. Attendee Jon Stricker, S.M.ASCE, won the door prize drawing of a Fitbit.

STRUCTURE magazine

June 2017

SEI Local Activities New University of Notre Dame Graduate Student Chapter Welcome to the newly established SEI Graduate Student Chapter at the University of Notre Dame, chaired by Andrew Bartolini (contact him at sei.colorado.chapter@ gmail.com) with Faculty Advisor Dr. Tracy Kijewksi-Correa. The chapter’s goals are to: • Promote the mission and purpose of SEI/ASCE at the University of Notre Dame. • Enable professional development opportunities for graduate students within the Department of Civil & Environmental Engineering & Earth Sciences. • Facilitate social events for graduate students within the Department of Civil & Environmental Engineering & Earth Sciences to build a community atmosphere amongst graduate students. • Collaborate with the Earthquake Engineering Research Institute Notre Dame Chapter (EERI@UND) in planning outreach activities throughout the Michigan area and professional development events for graduate students. • Collaborate with the Notre Dame Student Chapter of ASCE in planning outreach and professional development events.

STRUCTURE magazine

Get Involved in Local SEI Activities Join your local SEI Chapter, Graduate Student Chapter (GSC), or Structural Technical Groups (STG) to connect with colleagues, take advantage of local opportunities for lifelong learning, and advance structural engineering in your area. If there is not an SEI Chapter, GSC, or STG in your area, review the simple steps to form an SEI Chapter at www.asce.org/structural-engineering/sei-local-groups. Local Chapters serve member technical and professional needs. SEI GSCs prepare students for a successful career transition. SEI supports Chapters with opportunities to learn about new initiatives and best practices, and network with other leaders – including annual funded SEI Local Leader Conference, technical tour, and training. SEI Chapters receive Chapter logo/branding, complimentary webinar, and more.

Errata SEI posts up-to-date errata information for our publications at www.asce.org/SEI. Click on “Publications” on our menu, and select “Errata.” If you have any errata that you would like to submit, please email it to Jon Esslinger at jesslinger@asce.org.

June 2017

The Newsletter of the Structural Engineering Institute of ASCE

ASCE 7 Online provides digital access to Standard ASCE 7-10 and 7-16. The interactive tools and feature-rich functionality, such as side-by-side display of the Provisions and Commentary, redlining comparison, and annotation tools, is a must-have for any structural engineer. ASCE 7 Hazard Tool easy mapping feature will simplify your workflow and save you time. Visit our website at http://asce7.online to learn more. Structures Congress attendees had the opportunity to see a demo of the new ASCE 7 Online. Daniel Sack, Principal Engineer of The Babcock & Wilcox Company, won a free six-month subscription to ASCE 7 Online at our Exhibit Hall drawing. The print edition of ASCE 7-16 will be available in June and provides the most up-to-date and coordinated loading standard for general structural design. ASCE 7-16 describes the means for determining design loads including dead, live, soil, flood, tsunami, snow, rain, atmospheric ice, earthquake, wind, and fire, as well as how to assess load combinations. The 2016 edition of ASCE 7, which supersedes ASCE/SEI 7-10, coordinates with the most recent material standards, including the ACI, AISC, AISI, AWC, and TMS standards. Significant changes in ASCE 7-16 include the following: • New seismic maps reflecting the updated National Seismic Hazard Maps; • New wind speed maps, including new Hawaii maps, which result in reduced wind speeds for much of the United States, clarified special wind study zones, and separate Risk Category IV from Category III; • New snow load maps incorporating regional snow data for areas that previously required site-specific case study zones; • Updated rain duration provisions that align design requirements with International Plumbing Code provisions for drainage; • Entirely new chapter covering tsunami design provisions, which are important to Alaska, Hawaii, California, Oregon, and Washington; and • New appendix provisions for fire design. Standard provisions are accompanied by a detailed commentary with explanatory and supplementary information developed to assist users of the standard, including design practitioners, building code committees, and regulatory authorities. Standard ASCE/SEI 7 is an integral part of building codes in the United States and is adopted by reference into the International Building Code, the International Existing Building Code, the International Residential Code, and the NFPA 5000 Building Construction and Safety Code. Structural engineers, architects, and those engaged in preparing and administering local building codes will find the structural load requirements essential to their practice. Pre-order your copy today at www.asce.org/asce-7.

Structural Columns

ASCE 7-16 Print and Online Products

JUST RELEASED - Updated National Practice Guidelines for

CASE in Point

The Newsletter of the Council of American Structural Engineers

Specialty Structural Engineers

This document was prepared to supplement CASE’s National Practice Guidelines for the Structural Engineer of Record by defining the concept of a specialty structural engineer and the interrelation between the specialty structural engineer and the Structural Engineer of Record. CASE encourages the concept of one Structural Engineer of Record for an entire project. However, for many, if not most projects, there may be portions of the project that will be designed by different specialty structural engineers. The primary purpose of this document is to better define the relationships between the SER and the SSE, and to outline the usual duties and responsibilities related to specific trades. This is done for the benefit of the owners, the PDP, the SER, the SSE, and the other members of the construction team. The goal is to help create positive coordination and cooperation among the various parties. The committee did an all-inclusive update to this document and brought it to current industry standards. To view the updated practice guideline, go to www.acec.org/case/getting-involved/guidelines-committee.

CASE Risk Management Tools Available Foundation 7: Compensation – Prepare and Negotiate Fees that Allow for Quality and Profit Develop fees based on work effort (task hour) and value to be delivered • Make allowances for unknown conditions • Share the backup for your fee with the client when appropriate • Negotiate based on scope of work • Be willing to walk away • Don’t continue to work for losing clients Tool 7-1: Client Evaluation Do you know who your best clients are? Do you know where you should be focusing your marketing and sales efforts to maximize the financial performance of your firm? You may be surprised. This tool will help you answer those questions by analyzing the amount of work and profit for each client. Tool 7-2: Fee Development This tool is intended to be used within a consulting firm to stimulate thought and consideration in the development of fees. Engineers in firms that may be experiencing new responsibilities as project engineers and project managers often ask the question

– “How do we decide on fees?” This tool may be a useful primer for these employees and lead to a further discussion with firm management on the firm’s fee development strategies. Foundation 8: Contracts – Identify Onerous Contract Language Negotiate Clear & Fair Agreements • Understand ever changing contract language and demands • Use a CASE, AIA, or another accepted base contract • Modify the contract for each project, as needed • Use a contract that you can understand • Use the contract to reasonably share project risks • Get a signed contract • Utilized legal review when appropriate Tool 8-1: Contract Review Do you (or your legal counsel) review every contract to find onerous clauses? Do you know what they are? Do you always find them? This tool will help you find these clauses or words throughout the document. You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

Looking for Innovative Ideas! CASE Summer Planning Meeting Does your firm have an innovative idea or method of practice? Looking to get more involved in short duration projects? We are inviting you to “share the wealth” and submit a proposal for a web seminar topic, publication, or education session you would like to see CASE present at an upcoming conference. Our forms are easy to use, and you may submit your information via email. Go to www.acec.org/coalitions and click on the icon for Idea Sharing to get started. Questions? Contact us at 202-682-4332 or email Katie Goodman at kgoodman@acec.org. We look forward to helping you put your best ideas in front of eager new faces!

August 2 – 3, 2017; Chicago, IL

The CASE Summer Planning Meeting will again be scheduled for August 2nd to the 3rd in Chicago, IL. A popular feature of the planning meeting is a roundtable discussion on topics relating to the business of Structural Engineering, facilitated by the CASE Executive Committee members. Topics have included the Business of BIM, using social media within your firm, Peer Review, and Special Inspections. Attendees to this session will earn 2.0 PDHs. Please contact CASE Executive Director Heather Talbert (htalbert@acec.org) if you are interested in attending or have any suggested topics for the roundtable.

Follow ACEC Coalitions on Twitter – @ACECCoalitions. STRUCTURE magazine

June 2017

August 3 – 4, 2017; Chicago, IL

How important is Risk Management to your firm? A proper program can reduce your chances of being sued and allows you to take on more risky projects which can generate a substantial profit when handled right. Developed by the Council of American Structural Engineers (CASE), Time-Tested Techniques for Managing Your Firm’s Risk will help your firm reduce its rate of claims against structural engineering projects, as well as raise the level of quality services provided by all project participants. Who should attend? Principals, Owners, Project Managers, Risk Managers SEMINAR AGENDA

ACEC’s 2017 Annual Convention and Legislative Summit On April 22 – 26, a record 1,500 ACEC members attended the ACEC Annual Convention in Washington, D.C., meeting with Senators, Congressmen, and Capitol Hill staffers to advocate for major infrastructure legislation in 2017 that both incentivizes private investment and includes substantially increased direct investment in core federal programs. Attendees highlighted the need to address critical transportation, water/wastewater, and other infrastructure needs. In addition, ACEC members engaged lawmakers on key industry priorities for reforming the nation’s tax code. 600-plus celebrated the 50th Anniversary of the Engineering Excellence Awards Gala, which recognized 162 preeminent

engineering achievements from throughout the world. Hosted by SNL-Alum, Kevin Nealon, SR 520 Floating Bridge Replacement and HOV Program in Seattle, WA was honored with the 2017 Grand Conceptor Award on April 25th. The engineering work for the project was done by HDR. ACEC’s Annual Convention also marks the induction of a new ACEC Executive Committee. Sergio “Satch” Pecori, Chairman and CEO of Hanson Professional Services, succeeded Peter Strub as ACEC Chairman for 2017-2018 at the spring meeting of the ACEC Board of Directors.

Pathways to Executive Leadership – Class Two Registration Now Open! A practical, focused program for new leaders facing the challenges of a continuously evolving business environment. New practice-builders need specific and relevant training in the intricacies of leading an A/E firm in ever-changing, always uncertain economic times to be successful at taking on higher levels of leadership responsibility and prepare for the demands of being owners. Pathways to Executive Leadership is an intensive leadership program for early-career elites and promising mid-career professionals with 8-12 years of experience who are just beginning to lead and think strategically about their practices and careers. The reality-based curriculum focuses on the core skills necessary to think strategically in their markets, build effective teams, and deliver great service for their most valued clients. TARGET AUDIENCE: Pathways to Executive Leadership fills a vital gap and creates a strong connection between ACEC’s Business of Design Consulting curriculum and the Senior Executive’s STRUCTURE magazine

Institute capstone program. It targets those who are making the transition between managing one team (e.g., project managers) to those managing managers and multiple teams. This program is designed to establish habits for long-term high-performance and to create a trusted, national network of colleagues with which to make the journey. FLOW OF LEARNING: New skills to manage people and the uncertainty of a continuously evolving business environment are required to confront that challenges that budding practice builders face. Pathways to Executive Leadership will lead participants through a practical curriculum focused on becoming more balanced in their personal and professional life, more influential in team development, coaching, and client relationships, and more strategic in their business relationships to build a strong client portfolio. For more information, visit http://programs.acec.org/ 2017-pathways or contact Katie Goodman, 202-682-4332, or kgoodman@acec.org.

June 2017

CASE is a part of the American Council of Engineering Companies

Thursday, August 3 6:00 pm Dinner – The Future of Structural Engineering Speaker: Ashraf Habibullah, President, Computers and Structures, Inc.

Friday, August 4 7:30 am Welcome – Corey Matsuoka, SSFM International 7:45 am Four Strategies Forensic Engineers Use to Unravel Construction Disputes Benjamin Cornelius, S.E., P.E., Leslie E. Robertson Associates Consulting Structural Engineers 9:15 am – Break 9:45 am Information Security in Contracting Nicholas Merker, Ice Miller LLP 11:30 am Roundtable Lunch Moderator – Corey Matsuoka, SSFM International 1:45 pm Professional Liability Case Study Marathon Karen Erger, Lockton Companies Eric Singer, Ice Miller LLP Brian Stewart, Collins, Collins, Muir + Stewart 3:15 pm Wrap-up & Adjourn Corey Matsuoka, SSFM International To register for this event, go to www.acec.org/calendar/ calendar-seminar/case-risk-management-seminar. For more information about this seminar, contact Heather Talbert (htalbert@acec.org) or Katie Goodman (kgoodman@acec.org).

CASE in Point

Annual CASE Risk Management Seminar

Business Practices

business issues

Techniques to Successfully Navigate Networking By Jennifer Anderson

or some people, networking can feel awkward, seemingly self-serving, discouraging, and embarrassing. Many people consider the term “networking” a dirty word and don’t like the uncomfortable feelings that come from attending a networking event. Uncomfortable or not, if you are not networking it is going to be much harder to grow and develop as a professional. Networking has the power to help you in very significant ways in your professional and personal life. Here are four techniques to help you navigate the world of networking more effectively:

1) Know the difference between types of networking. Networking is not limited to attending a luncheon to hear a guest speaker share thoughts about an engineering-related topic. Networking can look like many different things, such as volunteering for a cause that you’re interested in, participating in an industry conference, taking additional college courses and getting to know your classmates through group projects, joining a running club, etc. Networking should never just be limited to a bunch of people in a room trying to get other people’s business cards! It can be valuable to attend luncheons and learn from other professionals, but take some time to evaluate how you want to make a difference in the world. For example, the author has a colleague who is interested in STEM initiatives (Science Technology Engineering Math) to help adolescents learn about different ways they can engage with the world of science. She believes it is important to help children and teens learn about how science and math help better our world. As those children come to realize the importance of STEM, their eyes are open to more possibilities and they find themselves more interested in math and science courses. The ripple effect is quite substantial in their education and career. Along the way, the author’s colleague has met numerous individuals that have helped her stay connected in her field. As you find groups of people that are interested in similar hobbies, community causes, professional development, and so on, you will come to meet some interesting, passionate,

and thoughtful people. That is when networking becomes fun and worthwhile!

2) Make time for networking. This might seem like a very basic technique, but if it is so easy to do, why is it that people do not make time for networking? Typically, people say that they are “too busy” with work to network. Later, when they are ready to make a job change, or ready to hire a new person for the firm, they find themselves with weak connections to people in their network. It is imperative to invest in people within your professional circles so that you have time to get to know them, help them get to know you, and build trust and respect. In the future, you will be able to comfortably contact them when you’re looking to make a job change or add to your team. If you find yourself feeling awkward at networking events, it’s likely you are not participating in something that is interesting and thought provoking for you. Action item: Look at your calendar and mark out two lunch periods in the next month to meet up with people from your network. These networking lunches may well be some of the best “work time” you will spend each month because you are likely to connect on a deeper level, gain valuable professional and technical insights, and also make an impact in someone’s life.

3) Follow-up on social media. After meeting new people, follow-up and connect through social media. LinkedIn is a good place for connecting with other professionals. LinkedIn is where most people think to connect professionally. However, just because you are connected on LinkedIn or any social media platform, does not mean you are instantly best friends. You will still need to nurture and cultivate relationships. Look for ways to stay in touch. Find articles to share that you think will be interesting to specific people, then copy the URL and share it in a message. Periodically check-in with your connections. Even a quick “hello” message goes a long way. One major benefit of being active on social media is that you will be exposed to other people who are likely outside your immediate circles. Join groups on LinkedIn to interact

STRUCTURE magazine

June 2017

with other interesting people who live and work in different marketplaces. Gaining a different perspective is a great way to grow and develop – and you may be an interesting voice to someone else on the other side of the country.

4) “So, what do you do?” Be ready to answer that all-too-common question. In answering, you have a chance to differentiate yourself from other people. Stop and think about it. That question can be a conversation killer. After you answer the question, you typically ask them back, “And what do you do?” Then where do you go with the conversation? It’s a dead-end question. The best advice – do not ask the dreaded “So, what do you do?” Instead, ask other thoughtful questions that will help you to engage in better conversations. Some thought provoking questions might include: “What brings you to this conference?” “What have you learned at this trade show that is the most intriguing to you so far?” “What does your firm specialize in?” Conversations get much more interesting and far less awkward when you are discussing meaningful information beyond telling each other your job titles. In the end, networking can lead to great conversations and professional opportunities. Do not hold yourself back. Plan to attend. Go. Bring your business cards. Have a goal of meeting 2 to 3 people at the event. Then, wash, rinse, and repeat. You will find, over time, that networking is truly a wonderful way to meet and keep in touch with other interesting and thought-provoking professionals. This will, in turn, provide you the opportunity to build your professional network, enhance the probability for future opportunities in your career, and add interesting people to your firm.▪ Born into a family of engineers but focusing on the people side of engineering, Jennifer Anderson (www.CareerCoachJen.com) has nearly 20 years helping companies hire and retain the right talent. She may be reached at jen@careercoachjen.com.