STRUCTURE magazine | April 2022

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APRIL 2022

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APRIL 2022

Contents APRIL 2022

LEAVES OVER FOREST PARK

Cover Feature

By David Fields, P.E., S.E., and Ronald Klemencic, P.E., S.E., Hon. AIA

St. Louis’ latest high-rise building, the 36-story 100 Above The Park, consists of eight tiers of stacked and undercut floor plans, each four stories high and shaped like the leaves of a tree. This unconventional form became a reality by using novel framing solutions.

The primary arrangement of columns, slab reinforcement, and P.T. tendons is set at a 130-degree angle about the centerline of the building. Of course, wind turbulence and pressures were of great concern. Cover photo courtesy of Tom Harris Photography.

F E A T U R E S PAN AMERICAN UNITY

By Erik Kneer, S.E., et al.

Pan American Unity is the largest contiguous mural created by Diego Rivera. Rivera painted the mural for the Golden Gate International Exposition in 1940. Well after the fair ended, Rivera’s masterpiece was installed in the City College of San Francisco’s Diego Rivera Theatre.

Unfortunately, the building was seismically unsafe and the exhibit needed a new home. The safe relocation of this cultural treasure was complex and challenging.

ADAPTIVE REUSE OF THE HISTORIC WITHERSPOON BUILDING – PART 5

By D. Matthew Stuart, P.E., S.E., P.Eng

Pertinent drawings detailing the Witherspoon’s original construction were unavailable prior to the design process for the adaptive reuse of the building.

In this, the final article in the series, the author reports the finding of several construction photos and the original 1895 specifications from a source originally unknown to the Team. This article examines how well assumptions made during the adaptive reuse project match the original Specifications.

STRUCTUREmagazine

C O L U M N S a n d D E PA RT M E N TS Editorial

A Case Against Remote Work

By Kevin H. Chamberlain, P.E.

Building Blocks

Better Materials, Better Designs, Better Ways to Build with Concrete By Luke Pinkerton, P.E., and Jayendra “Jay” Patel, P.E.

InSights

Reducing Embodied Carbon in Structural Concrete

By David Diedrick and Cecile Roman

Structural Design

Special Steel-Reinforced Concrete

Structural Walls – Part 1

By David A. Fanella, Ph.D., S.E., P.E.

Structural Performance Wind Uplift

Wood Roof Detailing for

By John “Buddy” Showalter, P.E.

Structural Connections

Applications

Undercut Anchors for Structural

By Philipp Mahrenholtz, Ph.D., Mark Ziegler, P.E., and Derrick Watkins, Ph.D., S.E.

Structural Inspections

Jurisdictional Inspections, Structural Observation, and Special Inspections By John A. Dal Pino, S. E.

Just the FAQs

FAQs on ASCE Standards

By Laura Champion, P.E., and Jennifer Goupil, P.E.

Guest Column

The REACH at the Kennedy Center

By Yvonne Nelson, P.E.

Construction Issues

Adjacent Construction

Evaluation and Mitigation of Risks from

By Antonio De Luca, Ph.D., P.E., S.E., and Meeok Kim, P.E., Ph.D.

Engineer’s Notebook Simple Capacity Checks for Commonly Used Steel Sections

Guest Column

By Hee Yang Ng, MIStructE, C.Eng, P.E.

Cross-Laminated Timber

By Borjen Yeh, P.E.

Structural Influencers Career Reflections

By Carol Post, P.E., S.E.

Historic Structures

Silver Bridge Failure 1967 (aka Point Pleasant Bridge) By Frank Griggs, Jr., D.Eng, P.E.

In Every Issue

Advertiser Index Resource Guide – Engineered Wood Products NCSEA News SEI Update CASE in Point

Spotlight

Efficient Structural Elements Define Stunning Architectural Form

Spotlight

Investigation and Creative Structural Solutions

Structural Licensure

SECB Passing the Torch

By the SECB Board of Directors

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, the Publisher, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

STRUCTURE magazine is not a peer-reviewed publication. Readers are encouraged to do their due diligence through personal research on topics.

APRIL 2022

EDITORIAL A Case Against Remote Work

By Kevin H. Chamberlain, P.E.

Y

our business development, and sanity, can benefit greatly emergency, it seems to be here to stay. Why? Because employees of all from becoming an active member of CASE. I can speak ages and backgrounds want that flexibility in their lives, we are told. highly of the membership benefits, and I’m not just saying that Is this shift towards remote work good for our profession? We cannot as the incoming chair! possibly know for a while. That would Have you been to a national conferbe like trying to predict the course of ence? If you get the chance, go! I returned COVID. We are still in the early stages of How do you replicate recently from the ACEC Coalitions this industry-wide transition to remote Winter Meeting in sunny San Diego, work, with primarily existing, experithat one-on-one mentoring where CASE met over 2 days. Besides enced employees who developed under experience thru an internet taking a much-needed break from a cold the old-fashioned method of working connection? This may burst winter, it was an excellent opportunity in a physical office around other more to meet with structural engineers from experienced engineers and learning from some bubbles, but I do not all corners of the country. It was great them. think you can. to see familiar colleagues and make new So how are new hires supposed to learn? connections. We share stories of painHow do you replicate that one-on-one in-the-neck claims, demanding clients, mentoring experience thru an interhiring challenges, and lessons learned over net connection? This may burst some laughs and a beer. To me, personal conbubbles, but I do not think you can. nections are the most important part of Young engineers need to be in a setting what we do to advance our profession. It motivated me to join CASE with more experienced staff in close communication every day. They 5 years ago and it is why I remain engaged. were learning not just by direct instruction but by listening to all the Unfortunately, for 18 months, we could not connect as we did goings-on around them, which you can’t get in an office of one. Fine, because of an ugly 5-letter word. COVID sucked the joy out of many but how long does that need to last? To answer that question, ask aspects of our lives, and the structural engineering profession was yourself when you think you will finish learning how to be a sound no exception. It transcended business practice issues, employment, engineer. After a 6-month training period? After sitting for the PE hiring, retention, and professional development.

Exam? After becoming a junior principal? None of the above. An At CASE, we meet in person twice a year, winter and summer. Our excellent structural engineer stops learning when they retire and take showcase offering at our conferences is our roundtable. We all pack up golf. The back and forth of the mentoring process should have no into a room and talk about business practice topics of most interest end date or age limit. I have found that when you meet a structural to our firms. We make a list of them and pick a few to discuss by engineer who has “learned everything they need to know,” chances a show of hands. For me, hearing business issues faced by the rest are you will find that person is also a lousy engineer. of the room is like binge-watching Netflix; pass me the popcorn! I On the employee side, think carefully about what it is you’re asking always learn something from the roundtable, and I try to impart a tiny for. Do you have a home office with a door you can shut or one end nugget of wisdom from my small perch back home in New England. of the dining room table with chaos encroaching from all sides? How During COVID, we tried to do virtual conferences, but it is just not do you set boundaries: for yourself, for your family, for your employer, the same. For me, it was slightly more engaging than a manufacturer’s and your firm’s clients? Each has a different interest and goal. Will you lunch and learn, but without the lunch. So when we were able to be sidetracked by distractions that drop your productivity and mean meet together in the flesh last summer for the first time, it was like you have to put in extra hours when the house is quiet? How will a ray of sunshine. clients contact you? More and more, I see engineers listing their cell On a parallel track, the most prominent topic of discussion at our phone numbers on their email signatures. To me, that screams, “Hey, most recent roundtable was remote work. That is what I find the most call me on weekends and after-hours.” Giving out cell phone numbers fascinating because, at our firm, remote work has not been a “thing.” at our firm is a good way to get fired. How does your employer know The only remote work happens when one of the engineers has a sick if you are happy? How do they help you grow professionally? Never child at home or when I decide to get “caught up” when the weather forget that you are their most prized asset. Is a remote employment is bad on the weekend. We had a few weeks at the pandemic’s start marriage doomed to end in divorce? when we tried to minimize the number of people in our building Hopefully, COVID can be buried on the scrap heap of history very simultaneously, but we have moved on. soon. Let’s hope that the development of new structural I have learned that most firms offer at least some component for engineers and the health of our profession today and in the remote work. Some mega-firms have eliminated desk space and are future does not fall victim to its legacy.■ renegotiating office leases to reduce their footprint. Many firms Kevin H. Chamberlain is the CEO and Principal of DeStefano & Chamberlain, have a significant portion of employees working remotely. Although Inc. in Fairfield, CT, and the Chair-Elect of CASE (kevinc@dcstructural.com). the remote work trend incubated and grew under a public health STRUCTURE magazine

APRIL 2022

7

building BLOCKS Better Materials, Better Designs, Better Ways to Build with Concrete

By Luke Pinkerton, P.E., and Jayendra “Jay” Patel, P.E.

D

esign and construction profesCONSTRUCTION: DESIGNING AND BUILDING sionals face significant challenges driven by increasing customer Considering the specific needs of the construction project and using Optimize concrete mixes demands, rapidly changing material only the materials necessary, avoiding excess emissions. costs, strained supply chains, and Switching to solar, wind and other renewable sources of energy the loss of workers going into trades.

Use renewable fuels directly reduces emissions from other energy sources.

However, a more urgent issue is that Increase the use of climate change is a real and growing Diverting these materials from landfills. recycled materials concern. Given that concrete is the most widely used manufactured mateDesigning for the specific needs of the construction project reduces Avoid overdesign and rial in the world, the Portland Cement unnecessary overproduction and emissions; leverage construction Association (PCA) has challenged the incorporating just-in-time deliveries. technologies design and construction industries to Improve design and specifications to be more performance oriented address the issue by developing a roadwhich will permit innovation in cement and concrete manufacturing.

Educate design and map to carbon neutrality by 2050. construction community Encourage the use of advanced technologies to improve structural Optimization, avoiding overdesign, performance, energy efficiency, resiliency, and carbon sequestration. designing for performance, and leveraging technology are critical parts of Figure 1. PCA roadmap. Courtesy of Portland Cement Association. the path (Figure 1).

This article focuses on one opportunity to improve environmen- of slabs are common in Automated Storage and Retrieval Systems tal outcomes and the competitiveness of concrete as a structural (ASRS)with rack-supported roofs (Figure 2). system by using advanced analysis, materials, and methods to design plain concrete in structural ground-supported slabs. Since What is Plain Concrete? structural ground-supported slabs are foundations that transmit vertical loads or lateral forces from other portions of a structure The definition of plain concrete is structural concrete with no reinforceto the soil, the provisions of ACI 318, Building Code Requirements ment or with less reinforcement than the minimum amount specified for for Structural Concrete and Commentary, are applicable. These types reinforced concrete in the applicable building code.

Figure 2. ASRS warehouse – racks act as structural columns.

8 STRUCTURE magazine

Concrete strength in bending can be divided into two categories: a) Flexural Strength/Modulus of Rupture (MOR) – the stress that causes it to crack initially and, if it contains reinforcement, b) Residual Strength – the stress it can carry after a crack has developed.

Flexural Strength The flexural strength of concrete is generally inferred by the concrete’s compressive strength rather than being directly measured (ACI 318 and ACI 332, Residential Code Requirements for Structural Concrete).

It can, however, be specified directly and measured with the three-point load test (ASTM C78, Concrete Beam Bend Testing) (Figure 3) or center point load test (ASTM C293, Standard Test Method for Flexural Strength of Concrete).

Increasing compressive strength typically increases flexural strength, but flexural strength can also be optimized by careful design of the concrete mix and special additives. For example, a concrete mix with large, angular, well-graded blended coarse aggregates provides higher flexural strength than a mix using Figure 3. ASTM C78 flexural test. small, round, gap-graded aggregates.

ACI 322-72, Building Code Requirements for Structural Plain Concrete, Elastic Method was the last code that allowed for direct design of plain concrete with flexural testing. Currently, allowable flexural strength is determined In the 1920s, Westergaard developed equations that provide the using a fixed multiplier 5√f´c of the compressive strength in the bending stress at the point of initial cracking in a slab-on-ground building code (ACI 318). In contrast, a higher multiplier, 7.5√f´c , is based on the loading and soil subgrade modulus (Figure 4a, page 10). available in the residential code (ACI 332).

The Westergaard approach is outlined in ACI 360, Guide to Design ACI 380, a new committee formed to pick up where ACI 322 left of Slabs-on-Ground, Chapter 7. This method typically leads to the off in the 1970s, is investigating the feasibility of using a higher mul- most conservative design, given that it treats the initial cracking point tiplier or a performance-based design based on the flexural strength at the bottom of the slab under the load as the ultimate condition. for plain structural concrete. It would recommend modifying ACI Yield Line Methods 318-14 Equation 14.5.2.1a to use a higher multiplier of compressive strength or the actual specified flexural capacity (per ASTM C78) in In the 1960s, Meyerhof developed a method of design that nearly doubled place of the prescriptive 5√f´c for the allowable flexural strength of the efficiency of plain concrete pavement designs (Figure 4b, page 10). the concrete in ACI 318 as in Equation 1.

The solution, which considers the redistribution of loads after initial cracks form on the bottom slab surface, was conceived originally for φMn = φfr Sm (Eqn. 1) plain concrete pavements. The method has been adapted with lightly where: reinforced and fiber reinforced concrete. Walker and Holland, and fr = MOR estimated taken as a multiplier √f´c or specified with ACI 360, suggest this design approach is a valid option for slab-onrequired ASTM C78 testing ground design even with plain concrete.

Mn = Nominal Moment Capacity Allowing stresses to redistribute reduces the required thicknesses by a Sm = Section Modulus (bh2/6) factor of two or more. Resistance is computed using Equation 2, and φ = Resistance factor for bending (0.6) closed-form equations for demand are published in several references such as ACI 360-10 Chapter 11.

Residual Strength

Residual strength is the amount of force the concrete can take after a crack has formed. Plain concrete without any reinforcement has no residual strength. Steel rebar and wire mesh are the only forms of reinforcement recognized directly in ACI 318 for providing residual flexural capacity. While alternative reinforcement materials such as Fiber Reinforced Concrete (FRC) and Fiber Reinforced Plastic (FRP) rebar are proven in slab-on-ground applications, they are not currently recognized in ACI 318 as providing flexural reinforcement.

Design Three methods for slab design are considered: elastic design hand calculation methods (Westergaard), yield line methods (Meyerhof ), and linear finite element analysis.

Mn = (Mneg + Mpos)

(Eqn. 2)

where: Mn = Nominal Moment Capacity Mneg = Flexural strength (MOR) measured using ASTM C78 Mpos = Residual strength from bottom reinforcement, zero for plain concrete Testing has confirmed that even plain concrete ground-supported slabs have considerable capacity after the initial crack formation.

Finite Element Analysis Finite element analysis (FEA) is a method permitted by ACI 318 Section 6.9 that is able to consider the exact load distribution accurately. In this method, the concrete slab-on-ground, post loads/ reactions, and soil properties are incorporated into the computer APRIL 2022

9

Figure 4. a) Westergaard vs. b) Meyerhof Method. Courtesy of Walker and Holland.

model. SAFE, ADAPT, and RISAFoundation are commercially available software packages tailored to concrete foundation and slabon-ground design.

For heavy post loads such as for ASRS shown in Figure 2, slab thickness is almost always controlled by the flexural strength of concrete.

Worst-case bending moments for ACI 318 factored load combinations are obtained from computer models, and slab thickness is adjusted to keep bending stresses in the slab under allowable strength.

The Opportunity Attaining the ambitious goal of carbon neutrality put forth by the PCA is feasible if the industry focuses on developing materials that fully utilize the design methods indicated above.

Available now: • Design walls, footings, and structural slabs using linear elastic design with loads computed in accordance with ACI 318

Figure 5. Comparison of thickness and C02 footprint of various methods.

10 STRUCTURE magazine

Chapter 6 and resistance with ACI 318 Chapter 14 (5√f´c).

This includes the use of finite element analysis in ACI 318 Section 6.9.

• In residential concrete designed in accordance with ACI 332, using flexural strength equal to 7.5√f´c is permitted. This extends the plain concrete allowable capacity in residential concrete by 50% over the 5√f´c currently allowed in ACI 318 Chapter 14.

The following should be considered: • Revisions to ACI 318 Chapter 14 and ACI 332 Chapter 6 that allow a higher multiple of √f´c and/or actual modulus of rupture based on ASTM C78 for the design of plain structural concrete members.

• Revision to ACI 318 Chapter 14 and/or allowance in ACI 332 Chapter 6 that allow the use of moment redistribution methods for plain concrete ground-supported slabs, like that permitted for reinforced concrete foundations in ACI 318 13.2.6.2.

• Design considerations for steel reinforcement and/or alternative forms of reinforcement to provide post-crack capacity at the location of the first crack when the yield line method is employed.

In the meantime, ACI 318 Section 1.10 allows for alternative approaches. One example of an alternative approach is the use of the International Code Council Evaluation Service (ICC-ES) ESR 3949 for Helix 5-25 Micro-Rebar. This building product evaluation report contains a table of MOR values for concrete mixes derived from ASTM C78 testing. Design professionals may use these values as an alternative to those determined using the lower bound equation in ACI 318 Section 14.5.2 (5√f´c). The resulting higher nominal capacity available may be used to provide additional value in structural plain concrete applications such as slabs-on-ground, foundation walls, and

Table of example structural ground-supported slab designs.

Assumptions Analysis Method Resistance Factor, φ Allowable Bending Stress (psi), φ fr Residual Stress (psi)

Example 1

Example 2

Example 3

Example 4

Example 5

FEA

FEA

FEA

Meyerhof

Meyerhof

Plain Concrete

Plain Concrete

Plain Concrete

Plain Concrete

SFRC*

0.6

0.6

0.6

0.6

0.6

φ

5√fć

φ

7.5√fć

φ

9√fć

φ

9√fć

φ

9√fć

190

285

342

342

342

0

0

0

0

103

28.7

23.4

21.4

14.3

13.3

18%

25%

50%

54%

23.7

21.7

14.5

13.5

19%

25%

50%

54%

Results Thickness (in) % Change GWP* (kg CO2eq/ft2)

29.1

% Change

*GWP is conservatively estimated 3 kg/kg CO2 for ASTM A820 Type I SFRC and 338 kg/m3 CO2 for 4000 psi concrete. *SFRC: Steel Fiber Reinforced Concrete

Examples The design examples in the Table are based on the design of an ASRS slab with a rack-supported roof (Figure 2). Each example assumes the same loading configurations, concrete strength, and soil conditions. All designs include 50 lb/yd steel fiber. The steel fiber provides residual capacity in Example 5 and is used only for crack control in the other examples. Five approaches are considered with the required slab thickness and global warming potential (GWP) based on the ultimate limit state condition computed for each. The thickness and GWP are reduced as the assumptions change (Figure 5).

Conclusions

References are included in the PDF version of the online article at STRUCTUREmag.org.

Acknowledgments: Thank you to Wayne Walker, P.E., S.E., FACI (SSI), for his comments and contribution of figures, and Justin Idalski, P.E. (MI) (Helix Steel), for his support of example calculations.

Luke Pinkerton is the President, Chief Technology Officer, and Founder of Helix Steel. Luke is Chair of the direct tension task group of ASTM Committee C09, Secretary of ACI 380, and a consulting member of ACI 332 and ACI 551 (luke.pinkerton@helixsteel.com).

Jayendra “Jay” Patel is a Senior Structural Engineer for Haskell. Mr. Patel is a voting member of ACI 380, ACI 360, and ACI 551 and is an associate member of ACI 302 (jay.patel@haskell.com).

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• Actual measured flexural strength and higher multipliers of √f´c have been successfully used to design structural groundsupported slabs. Reductions in thickness and global warming potential (kg CO2 generated by materials used) of about 25% are possible with this approach.

• Applying the yield line method to plain concrete allows a reduction in slab thicknesses and global warming potential by about 50% • Applying the yield line method to concrete with bottom (positive moment) reinforcement (provided by steel reinforcement or alternative reinforcement such as FRC or FRP) allows for even thinner and lower carbon designs (more than 50% reduction).

• When allowed by code, alternative means and methods have been used successfully to satisfy code officials. Still, changes to the plain structural concrete chapter of ACI 318 would allow more widespread use of these field-proven methods.

Engineers who are already using methods like these need to communicate their experiences and support efforts (e.g., ACI 380) to revise codes. This can only help the industry answer the PCA

challenge. Meeting this challenge will reduce the environmental impact of concrete designs and help engineers better serve clients with more durable and efficient structures.■

www.dci-engineers.com

footings by allowing for thinner sections and/or reducing the need for traditional reinforcement.

APRIL 2022

11

INSIGHTS Reducing Embodied Carbon in Structural Concrete

By David Diedrick and Cecile Roman

A

nyone remotely involved in the building sector cannot help but notice the industry’s monumental shift in acknowledging the intensity of the climate crisis and in formulating strategies focused on meeting the urgent need to achieve net-zero carbon goals by midcentury. The heightened awareness and sustainability-driven activity will continue to gain momentum as highly influential players in the field, such as the Structural Engineering Institute of the American Society of Civil Engineers (SEI) and the American Institute of Architects (AIA) lead the charge.

For centuries, concrete has been used to build infrastructure. While well regarded for its durability and strength, concrete is also known as a CO2-intensive building material. Most of the embodied CO2 in concrete originates from the hydraulic cement used in the mix. As cement and concrete producers introduce new products and technologies, structural engineers need to be aware of the latest and emerging tools that impact the industry’s ability to tackle the challenges of embodied carbon. These tools will be critical to reducing the global warming potential (GWP) of projects and meeting SE 2050 and Architecture 2030 commitments.

SCMs and Blended Cements Supplementary Cementitious Materials (SCMs) impart a wide range of exceptional performance properties to concrete. SCMs can be natural pozzolans, or they can be byproducts of industrial processes. The most common are blast-furnace slag (byproduct of iron manufacturing), fly ash (byproduct of coal combustion in power plants), and silica fume (byproduct of silicon manufacturing).

SCMs are utilized in concrete as a separate component or as a constituent of blended cement. The amount used depends on performance needs and/or application. For example, a binary combination of Portland cement and slag can improve strength and reduce permeability. A ternary blend of Portland cement, silica fume, and slag can dramatically densify concrete and minimize permeability.

The partial replacement of Portland cement with SCMs results in stronger, more durable, and longer-lasting concrete, reduces greenhouse gas emissions, and diverts the disposal of industrial byproducts from landfills. The beneficial reuse of SCMs in concrete also contributes credits in a wide variety of LEED® categories.

ASTM C595 (Standard Specification for Blended Hydraulic Cements) specifies four types of blended cements. Type IL contains up to 15 percent limestone; Type IP contains up to 40 percent pozzolan; Type IS contains up to 95 percent slag cement; and Type IT contains either two different pozzolans, slag and a pozzolan, a pozzolan and a limestone, or a slag and a limestone.

PLC Portland Limestone Cement (PLC, Type IL) is an effective alternative to ordinary Portland cement for reducing CO2 emissions. As 12 STRUCTURE magazine

The high-strength concrete used in the One World Trade Center was designed with fly ash, silica fume, and slag cement SCMs.

allowed by ASTM C595, PLC can be manufactured with up to 15 percent of high-quality limestone. It can provide equivalent or better performance than Type I/II cements and has been rigorously tested to verify concrete strength development, durability, and other desired performance properties.

With Type IL cement, similar percentages of SCMs can be used in concrete mixes while also replacing up to 15 percent of the Portland cement with limestone. This results in the potential of an additional 10 percent reduction in greenhouse gas emissions associated with the production of Portland cement clinker. When combined with typical SCM replacement, the effective reduction in the CO2 footprint of concrete is highly significant.

Low-Carbon Concrete To meet growing demands for sustainable construction solutions, multiple concrete producers in the U.S. have developed low-carbon concretes to reduce embodied carbon. Available in various strength classes and compliant with industry standards, low-carbon concrete can be used in all types of structural applications, regardless of performance requirements.

One example, produced with low clinker content, is a low-carbon concrete called ECOPact, which was developed by Holcim to provide 30 to 100 percent fewer carbon emissions than standard concrete.

Up to 80 percent less carbon is achieved primarily with lower CO2intensive materials. The last 20 percent can be reached through offsets with certified carbon projects for fully carbon-neutral solutions.

Lastly, regarding contributions to a circular economy, the concrete can integrate upcycled construction and demolition materials where regulatory conditions allow.

Following sustainability targets, the engineer can either specify the desired percentage of GWP reduction compared to a regional baseline or indicate a maximum GWP value per class of concrete mix. It is recommended that the specifier works with local concrete producers to determine the most practical limits to apply to mixes since the availability of cements and SCMs varies by geographical location.

Relevant ASTM Specifications ASTM C 989

Standard Specification for Ground Granulated Blast-Furnace Slag for Use in Concrete and Mortars

ASTM C 618

Standard Specification for Coal Fly Ash and Raw or Calcined Natural Pozzolan for Use in Concrete

ASTM C 1240 Standard Specification for Silica Fume Used in Cementitious Mixtures

EPDs

ASTM C 150

Standard Specification for Portland Cement

ASTM C 595

Standard Specification for Blended Hydraulic Cements

ASTM C 1157 Standard Performance Specification for Hydraulic Cement

Life-cycle assessment (LCA) is a leading tool for assessing environmental performance, defined by ISO14040-14044 standards. Manufacturers support the LCA process with Environmental Product Declarations (EPD), an important tool for providing data and transparency on materials and supporting complex integrated design processes.

EPDs offer a substantive characterization of the environmental impacts for cement and concrete, and they provide a standard way to communicate the GWP of products. Based on the level of transparency, project teams can earn credits within the Materials and Resources category of LEED v4 for products that have verified EPDs.

The Path Ahead The building sector needs to act immediately if it is to decarbonize by midcentury. Reliance must shift to building materials that offer the lowest amount of embodied carbon possible without

sacrificing performance to curb global warming and meet the SE 2050 Challenge.

Various organizations, including the National Ready Mixed Concrete Association, the American Concrete Institute, and the Slag Cement Association, offer recommendations on specifying alternative cementitious materials to lower embodied carbon. Product manufacturers can also provide technical assistance to help develop specifications, and most offer detailed test results, quality-control records, and EPDs.■ David Diedrick is Director of Quality & Product Performance for U.S. Cement at Holcim (dave.diedrick@lafargeholcim.com).

Cecile Roman is a Civil Engineer and Commercial Innovation & Sustainable Solutions Manager at Holcim (cecile.roman@lafargeholcim.com).

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APRIL 2022

13

structural DESIGN Special Steel-Reinforced Concrete Structural Walls Part 1: Significant Changes to the Design and Detailing Requirements

By David A. Fanella, Ph.D., S.E., P.E., F.ACI, F.ASCE, F.SEI

S

ignificant changes were made to the design and detailing requirements for special steel-reinforced concrete structural walls in the 2019 edition of Building Code Requirements for Structural Concrete (ACI 318-19) (hereafter referred to as ACI 318). According to ASCE/SEI 7-16, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, special structural walls are required in buildings with bearing walls, building frames, and dual systems assigned to Seismic Design Category (SDC) D, E, or F. The following revisions to the design and detailing requirements can be found in Sects. 18.2.6 and 18.10 of ACI 318: 1) The introduction of Grade 80 and Grade 100 deformed bars to resist the effects from flexure, axial force, a combination of axial force and flexure, and shear.

2) New requirements for longitudinal bar termination and splice locations.

3) New requirements for minimum area of boundary longitudinal reinforcement for slender walls.

4) Increase of the design shear force for slender walls.

5) New requirements for expected wall deformation capacity and expected wall drift demand.

6) Revised details for transverse reinforcement within boundary elements and the wall web.

7) New maximum vertical spacing requirements of transverse reinforcement at wall boundaries.

The requirements for Items 1 through 3 are covered below; Items 4 through 7 will be covered in a future article.

High-Strength Reinforcement Recent advances, including substantial new research, have enabled reinforcing steels of higher strengths to be a viable option in a variety of applications in reinforced concrete structures, including buildings with special structural walls (see the CRSI Technical Notes Guide to the Use of Grade 80 Figure 1. Longitudinal bar requirements for special structural walls.

Reinforcing Bars in ACI 318-19 and Guide to the Use of Grade 100 Reinforcing Bars in ACI 318-19). Permissible applications of high- effects. All components of special structural walls, including coupling strength steel reinforcement (that is, reinforcement with a yield strength beams and wall piers, are permitted to use this reinforcement. of 80,000 psi or 100,000 psi) were significantly expanded in the 2019 According to ACI 318 Table 19.2.1.1, the minimum specified compresedition of ACI 318. For special structural walls, ASTM A706 Grade 80 sive strength of the concrete, f c´, for special structural walls with Grade and Grade 100 deformed reinforcing bars are permitted to be used to 60 or Grade 80 reinforcement is 3,000 psi; for walls with Grade 100 resist the effects of flexure, axial force, and a combination of flexure and reinforcement, minimum f c´is 5,000 psi. The higher minimum comaxial force. Similarly, ASTM A615, A706, A955, and A996 Grade 80 pressive strength for Grade 100 reinforcement enhances bar anchorage and Grade 100 deformed reinforcing bars are permitted to resist shear and reduces the neutral axis depth for improved overall performance.

14 STRUCTURE magazine

Utilizing high-strength steel reinforcement in special structural walls typically results in smaller bar sizes and/ or fewer bars than Grade 60 reinforcement. This translates to improved concrete placement and consolidation and a reduction in reinforcing bar congestion, especially in walls with coupling beams. It also results in lower placement costs because fewer bars need to be placed in the field.

Longitudinal Reinforcement Termination and Splice Locations New requirements for termination and lap splice locations of longitudinal reinforcement in special structural walls are given in ACI 318 Sect. 18.10.2.3.

Except at the top of a wall, longitudinal reinforcement must extend at least 12 feet above the point where it is no longer required to resist flexure but need not extend more than a tension development length, ld, above the next floor level, where ld is determined in accordance with ACI 318 Sect.

25.4.2.3 or 25.4.2.4. This requirement is illustrated for the termination of the “A” bars in Figure 1. The limit of 12 feet accounts for buildings with relatively large story heights.

At locations where yielding is likely to occur because of lateral displacements (like the base of a cantilever wall), ld must be multiplied by 1.25. This factor accounts for (1) the likelihood that the actual yield strength of the reinforcement exceeds the specified yield strength, fy and (2) the influence of strain hardening and cyclic load reversals. Extending longitudinal bars a distance of ld above the next floor level Figure 2. End longitudinal reinforcement requirements for special structural walls. is a more practical approach for bar development than the requirements in previous editions of ACI 318.

This requirement intends to promote the formation of well-distribLap splices of longitudinal reinforcement within defined boundary uted secondary flexural cracks in the wall plastic hinge region, thereby regions are not permitted over a story height, hsx, above a critical reducing the potential for reinforcement fracture at these locations. section and a distance of ld below a critical section where the value The minimum required distances above and below the critical section of hsx need not exceed 20 feet (see Figure 1 for the case of a single where the minimum longitudinal reinforcement must be provided critical section at the base of a wall where yielding of the longitudinal are the greater of lw and Mu/(3Vu ); these are the lengths over which reinforcement is likely to occur as a result of lateral displacements). yielding is expected. The terms Mu and Vu are the factored bending Boundary regions, in this case, include those within the lengths moment and shear force at the critical section, respectively, obtained specified in ACI 18.10.6.4(a) for the horizontal extent of special from analysis of the building using code-prescribed seismic forces. boundary elements and within a length equal to the wall thickness No more than 50 percent of the minimum longitudinal reinforcemeasured beyond the intersecting region(s) of connected walls (see ment is permitted to be terminated at any one section of the wall; the shaded areas indicated on the wall plan in Figure 1). Test results this requirement intends to avoid a weak section in the wall adjacent have shown that the inelastic deformation capacity of a structural to the anticipated plastic hinge region. wall is significantly reduced where lap splices are located at or near More in-depth information on the changes outlined in this article, the critical section. including design aids and worked-out examples, can be found in the CRSI Design Guide on the ACI 318 Building Code Requirements Minimum Area of Boundary Longitudinal Reinforcement for Structural Concrete. Also available is the CRSI Design Checklist A minimum area of longitudinal reinforcement must be provided at for Special Steel Reinforced Concrete Structural Walls, which contains the ends of structural walls or wall piers that are slender (that is, the an easy-to-use list of essential items that must be completed when overall height-to-length ratio, hw /lw is greater than or equal to 2), are designing and detailing special structural walls. Visit www.crsi.org essentially continuous from the base of the structure to the top of for more information on these and other CRSI resources.■ the wall, and are designed to have a single critical section for flexure and axial loads. According to ACI 318 Sect. 18.10.2.4, the minimum References included in the PDF version of the longitudinal reinforcement ratio within 0.15lw from the end of a vertionline article at STRUCTUREmag.org. cal wall segment and over a width equal to the thickness of the wall, b, must be 6√f c´/fy (see Figure 2, where the minimum longitudinal David A. Fanella is Senior Director of Engineering at the Concrete reinforcement ratio requirement is given in terms of the minimum Reinforcing Steel Institute (dfanella@crsi.org). area of reinforcing steel, Al(end)).

APRIL 2022

15

structural PERFORMANCE Wood Roof Detailing for Wind Uplift Fastener Schedules, Overhang Limits, and Uplift Connectors

By John “Buddy” Showalter, P.E.

C

omponent and cladding (C&C) wind pressures calculated using ASCE 7-16, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, increased over ASCE 7-10 C&C wind loads. In addition to larger corner and edge areas on roofs, ASCE 7-16 also includes increased roof pressures for low-rise (simplified) buildings with height (h) less than 60 feet and buildings taller than 60 feet with hip, gable, or flat roofs.

In the case of a flat roof on a low-rise building using the simplified method, pressures for corner, edge, and interior areas increased from 13 percent to 81 percent, with an average increase of over 40 percent.

For other roof slopes, Table 1 shows a comparison of C&C loads for ASCE 7-16 versus ASCE 7-10.

Between the increase of C&C roof areas assigned to corner and edge regions and the increase in C&C roof pressures, nail schedules for wood structural panels (WSP) changed, with nail spacing cut in half in some cases. For similar reasons, in the International Residential Code® (IRC), overhang detailing includes limits on gable endwall WSP cantilevers. Uplift connectors for gable endwall rake overhang outlookers to the endwall require engineering or can be sized based on Wood Frame Construction Manual (WFCM) prescriptive tables to account for increased C&C loads at roof edges.

Uplift connectors at rafter or truss bearings are based on main wind force resisting system (MWFRS) loads. MWFRS loads did not change in ASCE 7-16. The International Building Code® (IBC), IRC, and WFCM have uplift connection load tables that can be used to size roof-to-wall uplift connections.

The WFCM is a referenced alternative approach to the IRC based on IRC Section R301.1.1. Also, note that, for risk category I or II buildings, IBC Section 2309 permits the use of the WFCM and its load assumptions for buildings within the WFCM scoping limitations. Table 2 shows scoping provisions for the IRC and IBC Sections 2308 and 2309 relative to the roof and wind loads for light-frame wood construction.

Roof Sheathing Fastener Schedule Changes In the 2021 IBC and 2021 IRC, nailing patterns for wood structural panel roof systems have been updated. Reduced nail spacing is based on wind loads from ASCE 7-16 and is consistent with roof sheathing nailing requirements in the 2018 WFCM.

Figure 1. C&C wind loads at roof edges require tighter nailing schedules.

A previous article (STRUCTURE, Changes to the 2018 WFCM, June 2018) provided background information on increases to component and cladding wind loads in ASCE 7-16 which led to these changes (Table 1). Updates to the prescriptive fastener tables in the 2021 IBC and 2021 IRC now provide consistency with the building codes and their referenced standards.

IBC Table 2304.10.2 and IRC Table 602.3(1) contain similar prescriptive fastener schedules for wood construction. Only changes to IBC Table 2304.10.2 Items 30 and 31 are shown here (Table 3).

Similar changes were made to IRC Table 602.3(1).

Wind uplift nailing requirements for common species of roof framing with specific gravities (G) of 0.42, based on spruce-pine-fir (SPF), are the basis of the proposed nail spacing requirements in IBC Table 2304.10.2. This is to meet the wind uplift loading requirements of ASCE 7 without being overly complex in the specification of WSP roof sheathing nailing. The basic roof sheathing nailing schedule is 6 inches on-center at panel edges, and 6 inches on-center at intermediate supports in the field of the panel.

This nailing schedule applies to ⅜-inch through ¾-inch wood structural panels fastened to framing with 8d common or deformed

Table 1. Comparison of C&C wind loads for ASCE 7-16 versus ASCE 7-10. Courtesy, American Wood Council, Leesburg, VA.

Ratio of ASCE 7-16/ASCE 7-10 Roof GCp – GCpi Roof Slope

Roof Overhang GCp – GCpi

7 < θ < 20

3r

3e

2r

2n

2e

1

3r

3e

2r

2n

2e

1

1.36

1.14

1.69

1.69

1.16

2.02

1.27

1.11

1.59

1.59

1.14

–

27 < θ < 45

1.36

0.96

1.43

1.43

0.89

1.56

1.27

0.97

1.36

1.36

0.91

–

1.58

2.45

1.43

1.58

1.43

1.68

1.40

2.00

1.30

1.40

1.30

–

20 < θ < 27

16 STRUCTURE magazine

nails or roof sheathing ring shank (RSRS-01) nails with dimensions as shown in IBC Table 2304.10.2. Head diameters are specified and important due to new provisions in the 2018 National Design Specification® (NDS®) for Wood Construction accounting for head pullthrough in calculating nail capacities.

As shown in Table 4 for the common case of roof framing spaced at 24 inches on-center, nailing at intermediate supports in the interior portions of the roof is 6 inches on-center for wind speeds within the scope of IBC Section 2308 Conventional Light-frame Construction.

The 6 inches on-center spacing is also appropriate for edge zones (Figure 1) except where ultimate wind speeds equal or exceed 130

mph in Exposure B and 110 mph in Exposure C, where 4 inches on-center nailing is required. These special cases are addressed by the addition of IBC Table 2304.10.2 footnote “e” (Table 3).

To update the alternative fastening to uplift loading requirements of ASCE 7 without being overly complex in the specification of wood structural panel roof sheathing attachment schedules, IBC Table 2304.10.2 footnote “f ” was also added. The reference calculation leading to the use of 3 inches on-center spacing at all locations is based on a 0.113-inch diameter nail (e.g., 6d common) shank withdrawal from wood framing with specific gravity equal to 0.42 (SPF) and pre-calculated wind uplift loads in WFCM Table 3.10. continued on next page

Table 2. Scoping provisions for the IRC, IBC 2308, and IBC 2309 relative to the roof and wind loads for light-frame wood construction.

IRC

IBC 2308

IBC 2309 (WFCM)

Occupancy

One- and two-family dwellings, and townhouses as defined

All buildings except those within the scope of the IRC

All buildings, subject to the limitations of WFCM Section 1.1.3

Risk Category

N/A

I, II, III; IV if in SDC A

I and II

Maximum Number of Stories

3

3

3

Rafter or Roof Span

26’ lumber 36’ truss roof span

40’ roof span

26’ lumber/I-joists 60’ truss roof span

Slope

Flat – 12:12

Flat – 12:12

1.5:12 – 12:12

V < 130 mph; or V < 140 mph in non-hurricaneprone regions

V < 130 mph; or V < 140 mph Exp B in nonhurricane-prone regions

90 < V < 195 mph

Roof

Basic Wind Speed (3-sec gust) Exposures B, C and D

Table 3. Excerpt of 2021 IBC Table 2304.10.2 showing changes to fastening schedules for roof sheathing, including applicable footnotes (underline depicts new text; strikethrough indicates deleted text).

DESCRIPTION OF BUILDING ELEMENTS

g

NUMBER AND TYPE OF FASTENER

SPACING AND LOCATION

Wood structural panels (WSP), subfloor, roof and interior wall sheathing to framing and particleboard wall sheathing to framing

30. ³⁄8˝– ½˝

Edges (inches)

Intermediate supports (inches)

6d common or deformed (2˝ × 0.113˝) or 2 ³⁄⅜8˝ × 0.113˝ nail (subfloor and wall)

6

12

8d common or deformed (2½˝ × 0.131˝ × 0.281” head) (roof) or RSRS-01 (2³⁄8˝ × 0.113”) nail (roof)d

6e

12 6e

1¾˝16 gage staple, 7⁄16˝ crown (subfloor and wall)

4

8

4 3f

8 3f

1 ¾˝ 16 gage staple, 7⁄16˝ crown (roof)

3f

6 3f

8d common (2 ½˝ × 0.131˝); or 6d deformed (2˝ × 0.113˝) (subfloor and wall)

6

12

8d common or deformed (2½˝ × 0.131˝ × 0.281” head) (roof) or RSRS-01 (2³⁄8˝ × 0.113”) nail (roof)d

6e

12 6e

2³⁄8˝ × 0.113˝ × 0.266” head nail; or 2˝ 16 gage staple, 7⁄16˝ crown

4

8

2 ³⁄8˝ × 0.113˝x 0.266” head nail (roof)

31. 19⁄32˝″– ¾˝

(no changes to footnotes a through d) e. Tabulated fastener requirements apply where the ultimate design wind speed is less than 140 mph. For wood structural panel roof sheathing attached to gable end roof framing and to intermediate supports within 48 inches of roof edges and ridges, nails shall be spaced at 4 inches on center where the ultimate design wind speed is greater than 130 mph in Exposure B or greater than 110 mph in Exposure C. Spacing exceeding 6 inches on center at intermediate supports shall be permitted where the fastening is designed per the AWC NDS. f. Fastening is only permitted where the ultimate design wind speed is less than or equal to 110 mph. g. Nails and staples are carbon steel meeting the specifications of ASTM F1667. Connections using nails and staples of other materials, such as stainless steel, shall be designed by acceptable engineering practice or approved under Section 104.11.

APRIL 2022

17

Figure 2. Rake overhang outlooker uplift connection loads (WFCM Table 2.2C). Courtesy of American Wood Council, Leesburg, VA.

The use of a single 3-inch spacing at all supports was extended to staples based on the assumption that the ASCE 7 load increase would similarly require reduced spacing. This assumption was applied to staples because a withdrawal value is not available for staples in the NDS. Stainless steel nails have lower withdrawal strength when compared to carbon steel wire nails of the same diameter due to the reduced

surface friction of stainless steel. The differences in withdrawal strength vary with the specific gravity of wood (STRUCTURE, Changes to the 2018 NDS, February 2018). When stainless steel nails are specified as an alternative to reference smooth shank carbon steel wire (bright or galvanized) nails in wood construction, these differences in nail withdrawal strengths must be considered. For example, where smooth shank stainless steel nails are used for roof sheathing attachments, more nails or nails of greater length or diamTable 4. Excerpt of 2018 WFCM Table 3.10 roof sheathing attachment requirements for Exposure B wind loads. eter may be required to provide Courtesy of American Wood Council, Leesburg, VA. equivalent withdrawal strength Fastener Uplift Capacity (lbs) performance for wind uplift. IBC Table 2304.10.2 footnote “g” was 110 115 120 130 140 Wind Speed 3-second gust (mph) (See Figure 1.1) added to address this issue.

Fastener Spacing Sheathing Location

Perimeter Edge Zone

Gable Endwall Rake or Rake Truss with up to 9˝ Rake Overhang

Rafter/Truss Spacing (in.)

Panel Edges (in.)

Interm.

Supports (in.)

6

12

94

102

111

131

152

6

6

47

51

56

66

76

4

4

32

34

37

44

51

3

3

24

26

28

33

38

6

42

46

50

59

68

4

28

31

33

39

45

3

21

23

25

30

34

24

Uplift Load per Nail (lbs)

Fastener Spacing (in.)

Sheathing Thickness (in.)

18 STRUCTURE magazine

7/16

15/32

Framing Member G

0.42

0.49

0.42

0.49

8d common

68

100

67

98

10d box

82

118

81

120

RSRS-03

99

106

99

114

Gable Endwall Overhang Detailing While gable endwall overhangs can be engineered per the NDS, the WFCM contains helpful tables and details for common rake overhang conditions. Per WFCM Section 2.1.3.4(c), rake overhang length is not to exceed the lesser of one-half of the outlooker length or 2 feet.

WFCM Section 2.2.6.8 indicates that rake overhang outlookers shall be connected to the gable endwall in accordance with the wind uplift loads specified in Figure 2.

Tabulated outlooker uplift connection loads in Figure 2 are based on Zone 3 C&C roof wind loads

Table 5. Excerpt of 2021 IBC Table 2304.10.2 showing rafter or roof truss to top plate fastening schedule, and assume a building located in including applicable footnote.

Exposure B with a mean roof height (MRH) of 33 feet. For buildings 3-10 common (3˝ × 0.148˝); or 6. Rafter or roof truss to 2 toenails on one side and 1 toenail located in Exposure B with mean 3-16d box (3½˝ × 0.135˝); or top plate (See Section on opposite side of rafter or truss roof heights less than 33 feet or 4-10d box (3˝ × 0.128˝); or 2308.7.5, Table Exposures C or D, tabulated values 4-3˝ × 0.131˝ nails; or 2308.7.5) are increased with appropriate adjust4-3˝ 14 gage staples, 7⁄16˝ crown ments. Tabulated outlooker uplift c. Where a rafter is fastened to an adjacent parallel ceiling joist in accordance with this schedule and the connection loads are based on 2-foot ceiling joist is fastened to the top plate in accordance with this schedule, the number of toenails in the rafter overhangs, 2x4 gable endwall framshall be permitted to be reduced by one nail. ing, and uplift connectors location on the inside face of 2x4 gable endwall framing. For overhangs less than 2 feet, tabulated uplift connector load values can be decreased linearly. For overhangs located in Zone 2 per Figure 1, tabulated uplift connector loads are permitted to be multiplied by 0.74. Tabulated outlooker uplift connection loads are calculated assuming a roof pitch range greater than 1.5:12 and less than or equal to 6:12. For roof pitches greater than 6:12, tabulated values are permitted to be multiplied by 0.85. Tabulated Figure 3. Rake overhang lookout block limits (excerpt of WFCM Figure 2.1h). Courtesy of American Wood Council, uplift connector loads are specified in Leesburg, VA. pounds per linear foot along the gable endwall. Tabulated uplift connector loads are adjusted based on the connector spacing to determine 3.4C shows both Exposure B and Exposure C values and is expanded connection requirements (pounds). to show values based on overhang span and outlooker spacing.

In IRC Section R803.2.3, WSP roof sheathing cantilevers are limited to no more than 9 inches beyond the gable endwall unless supported Rafter/Truss Overhang Limits by gable overhang framing. This is consistent with WFCM prescriptive limits for rake overhangs. Per WFCM Sections 3.1.3.4(c) and There is no specific limit for rafter/truss overhangs in the IBC. Per 3.5.1.1.3, rake overhangs must be continuous 2x4 members, and the IBC Section 2308.8.2, structural elements not described in IBC overhang length is not to exceed the lesser of one-half of the outlooker Section 2308 require design. length or 2 feet (Figure 2). An exception permits rake overhangs using IRC Section R802.7.1.1 indicates that notches on cantilevered porlookout blocks of no more than 9 inches (Figure 3). tions of rafters are permitted provided the dimension of the remaining WFCM Table 3.4C provides rake overhang outlooker uplift connec- portion of the rafter is not less than 3½ inches and the length of the tion loads similar to those shown in Figure 2. However, WFCM Table cantilever does not exceed 24 inches per IRC Figure R802.7.1.1.

continued on next page Table 6. Excerpt of 2021 IBC Table 2308.7.5 showing required rating of approved uplift connectors and adjustments for MRH and Exposure category.

TABLE 2308.7.5 REQUIRED RATING OF APPROVED UPLIFT CONNECTORS (pounds) NOMINAL DESIGN WIND SPEED, Vasd

ROOF SPAN (feet) 12

20

24

28

32

36

40

OVERHANGS (pounds/feet)

85

-72

-120

-145

-169

-193

-217

-241

-38.55

90

-91

-151

-181

-212

-242

-272

-302

-43.22

100

-131

-281

-262

-305

-349

-393

-436

-53.36

a. The uplift connection requirements are based on a 30-foot mean roof height located in Exposure B. For Exposure C or D and for other mean roof heights, multiply the loads by the following adjustment coefficients:

Mean Roof Height (feet) EXPOSURE

15

20

25

30

35

40

45

50

55

60

B

1.00

1.00

1.00

1.00

1.05

1.09

1.12

1.16

1.19

1.22

C

1.21

1.29

1.35

1.40

1.45

1.49

1.53

1.56

1.59

1.62

D

1.47

1.55

1.61

1.66

1.70

1.74

1.78

1.81

1.84

1.87 APRIL 2022

19

Table 7. Excerpt of 2018 WFCM Table 2.2A uplift connection loads from wind. Courtesy of American Wood Council, Leesburg, VA.

Wind Speed 3-second gust (mph) (See Figure 1.1) Roof/Ceiling Assembly Design Dead Load

0 psf

10 psf

90

95

100

105

Roof Span (ft)

115

120

130

140

150

160

170

180

195

Unit Connection Loads (plf)

12

79

88

97

107

118

128

140

164

190

219

249

281

315

369

24

130

145

161

177

195

213

232

272

315

362

412

465

521

612

36

182

203

225

248

272

298

324

380

441

506

576

650

729

856

48

234

261

289

319

350

383

417

489

567

651

741

836

938

1100

60

287

319

354

390

428

468

509

598

693

796

906

1022 1146 1345

12

31

40

49

59

70

80

92

116

142

171

201

233

267

321

24

46

61

77

93

111

129

148

188

231

278

328

381

437

528

36

62

83

105

128

152

178

204

260

321

386

456

530

609

736

48

78

105

133

163

194

227

261

333

411

495

585

680

782

944

60

95

127

162

198

236

276

317

406

501

604

714

830

954

1153

WFCM Sections 2.5.1.1.2 and 3.5.1.1.2 specify that rafter overhangs shall not exceed the lesser of one-third of the rafter span or 2 feet.

Rafter/Truss Uplift Connections Uplift connectors at rafter or truss bearings are based on main wind force resisting system (MWFRS) loads. The IBC, IRC, and WFCM have uplift connection load tables that can be used to size roof-towall uplift connections.

IBC Section 2308.7.5 requires rafter and truss ties to the wall below with the resultant uplift loads transferred to the foundation using a continuous load path. The rafter or truss-to-wall connection has to comply with Tables 2304.10.2 and 2308.7.5. The former includes prescriptive rafter or roof truss to top plate uplift fastener schedules, as shown in Table 5 (page 19).

IBC Table 2308.7.5 (Table 6, page 19) includes tabulated uplift connection values for uplift connectors and overhangs. The uplift connection requirements are based on framing spacing of 24 inches on-center. Smaller spacings can be adjusted linearly. The uplift connection requirements include an allowance for 10 pounds of dead load.

For the effects of overhangs, the magnitude of the loads is increased by adding the overhang loads found in IBC Table 2308.7.5. The overhang loads are also based on framing spaced 24 inches on-center.

The overhang loads given are multiplied by the overhang projection and added to the roof uplift value in the table.

The uplift connection requirements are based on wind loading on end zones as defined in ASCE 7 Figure 28.5-1. Loads for connections located a distance of 20 percent of the least horizontal dimension of the building from the corner of the building are permitted to be reduced by multiplying the table connection value by 0.7 and multiplying the overhang load by 0.8.

Interpolation is permitted for intermediate values of Vasd and roof spans. The rated capacity of approved tie-down devices is permitted to include up to a 60-percent increase for wind effects where allowed by material specifications such as the NDS. IRC Table R802.11 contains ASD wind uplift loads based on ultimate design wind speeds (VULT) of 110 – 140 mph for both Exposure B and C. Rafter or truss spacings range from 12 to 24 inches on-center,

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roof spans range from 12 to 48 feet, and roof pitches are shown for less than 5:12 and for 5:12 and greater.

WFCM Table 2.2A includes wind uplift connection loads, as shown in Table 7. Tabulated uplift loads equal total uplift minus 0.6 of the roof/ceiling assembly design dead load. Tabulated unit uplift connection loads shall be permitted to be multiplied by 0.75 for framing not located within 6 feet of corners for buildings less than 30 feet in width (W) or W/5 for buildings greater than 30 feet in width.

Tabulated uplift loads are based on MWFRS wind loads and assume a building located in Exposure B with an MRH of 33 feet. For buildings located in Exposures B with MRHs less than 33 feet, or in Exposures C or D, tabulated values for 0 psf roof dead load can be adjusted then reduced by the appropriate design dead load.

Tabulated uplift connector loads are specified in pounds per linear foot along the wall. Tabulated uplift connector loads are adjusted based on the connector spacing to determine connection requirements (pounds).

Conclusion C&C wind pressures calculated using ASCE 7-16 increased over ASCE 7-10 C&C wind loads. Between the increase of C&C roof areas assigned to corner and edge regions and the increase in C&C roof pressures, nail spacings for wood structural panels (WSP) are significantly reduced in some cases. For similar reasons, in the IRC, overhang detailing includes 9-inch limits on gable endwall WSP cantilevers. Uplift connectors for gable endwall rake overhang outlookers to the endwall require engineering or can be sized based on WFCM prescriptive tables to account for increased C&C loads at roof edges.

Uplift connectors at rafter or truss bearings are based on MWFRS loads. MWFRS loads did not change in ASCE 7-16. The IBC, IRC, and WFCM have uplift connection load tables that can be used to size roof-to-wall uplift connections.■ John “Buddy” Showalter is a Senior Staff Engineer with ICC’s Product Development Group (bshowalter@iccsafe.org).

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structural CONNECTIONS Undercut Anchors for Structural Applications Ductility is Critical

By Philipp Mahrenholtz, Ph.D., Mark Ziegler, P.E., and Derrick Watkins, Ph.D., S.E.

U

ndercut anchors are the ultimate post-installed anchor category as their load transfer mechanism is bearing, similar to that of headed studs cast into concrete. Tensile loads are transferred into the concrete employing an expansion sleeve driven over a cone into a cavity formed at the back of the drill hole. This mechanical interlock prevents the anchor from pulling out and results in high load capacities and small displacements. Undercut anchors typically show low sensitivity to extreme conditions like large crack widths. If they are made of steel with sufficient material ductility, undercut anchors show a high resistance against cyclic loading. When Figure 1. Undercut anchor (left) and typical applications (right; example shows jet fans anchored to tunnel lining). designed properly and with sufficient embedment depth, undercut anchors have high strength, stiffThe Undercut Anchor System ness, and ductility levels. The design of undercut anchors can also qualify them for ductile connections in seismic design. Their general Undercut anchors outperform other post-installed anchor types robustness makes undercut anchors the preferred choice for critical in several ways. High-performing undercut anchors do not fail in connections where high load demands are to be transferred into the pullout failure mode because the load is transferred deep into the concrete safely (Figure 1). concrete by their solid bearing at the end of the anchor. The anchor load capacity is controlled by the concrete breakout capacity or the steel capacity of the rod, depending on the embedment depth and steel grade chosen.

For this reason, undercut anchors are very stiff and develop small displacements when loaded in service.

Their ultimate load capacity is very high, and their unique design facilitates the yielding of the rod along its entire length, resulting in a pronounced ductile behavior and deformation capacity.

In Figure 2, load-displacement plots give the typical behavior of Adhesive Anchors (AA), Screw Anchors (SA), Undercut Anchors (UA), Expansion Anchors, sleeve type (EA-s), and Expansion Anchors, bolt type (EA-b). As can be seen, the stiffest and strongest anchors are the adhesive and undercut; however, the undercut anchor has greater ductility in tension than the expansion anchor due to the rod of the undercut being debonded over its length (greater stretch length).

Figure 2. Typical load-displacement curves measured for anchors d = 3/8 -inch tested in tension, made of B7 Undercut anchors are used for critical applications steel or other grades of similar strength, and installed in ~3500 psi concrete with an effective embedment of ~4-inch. The undercut anchor shows high load capacity and small displacement at serviceability load where their superior performance outweighs the level (factored design strength of the rod is at ~7000 lbf) but large ductility reserves at higher loads. higher relative costs when compared to other anchor

22 STRUCTURE magazine

High-performing undercut anchors do not fail in pullout failure mode because the load is transferred deep into the concrete by their solid bearing at the end of the anchor.

Figure 3. The complete undercut anchor system: anchor, primary drill bit, undercut bit, setting tool, hammer drill, dust extractor.

types like expansion anchors. Specifiers, design professionals, and installers should thoroughly understand the complete undercut anchor system and its benefits. The correct selection of anchors and accessories is vital for proper and easy installation. The complete system typically consists of: • Undercut anchor • Drill bit for the primary hole • Undercut bit for creating the undercut • Setting tool to set the anchor • Hammer drill for drilling, powered undercutting, and setting • Optional dust extractor to vacuum the dust created during drilling

Installation

(HEPA) vacuum is the easiest way to meet the Occupational Safety and Health Administration’s (OSHA) Table 1 requirements for reduced dust exposure. Hollow bits remove the drill dust directly via a vacuum connected to the bit during drilling. Then, the undercut is customarily formed utilizing a special undercut bit, which may be hollow to suction drill dust, easing the undercutting operation and making an additional cleaning step unnecessary.

Finally, the anchor is inserted in the hole, and the expansion sleeve is driven over its conical end with a hammer drill. The sleeve allows the internal anchor rod to stretch freely when loaded in tension while it supports and prevents the rod from buckling when loaded in compression. This is a critical requirement for ductile design, as discussed later in this article.

Selection

Undercut anchors are not an everyday construction product, but they can provide specialized solutions. Undercut anchors do not require There are various versions of undercut anchors available and in different specialized installer training. However, since they are utilized less materials. Selecting the optimum version requires careful consideration frequently than typical expansion anchors, the installer should be famil- by the planner. The design professional defines the anchor diameter iar with their installation (Figure 3). and embedment depth to meet the No power tool other than a hammer load demand. Other parameters like drill is required for the installation. The minimum thickness and edge distance primary hole is drilled straight into the of the concrete member also influence concrete member using a drill bit that the selection of the appropriate anchor. meets the American National Standards The standard version of the anchor is Institute (ANSI) requirements. The hole pre-set prior to the installation of the has to be drilled to the required depth. fixture. There is also a thru-bolt version Stop drill bits are also available, which where the sleeve protrudes through control specific dimensions meeting the fixture for increased shear load the anchor requirements and can be a capacities. foolproof choice.

Selecting steel grades that are ducUndercut anchors are insensitive to tile and develop sufficient resistance uncleaned holes; however, to faciliagainst cyclic loading for high seismic tate the following undercutting and load capacities is critical (Figure 4). installation procedure, the hole must Ductile steel is also critical in applicabe free of drill dust which is removed tions where the anchor may be subject using a vacuum or compressed air for to fatigue loading, like the anchordownward and horizontal installa- Figure 4. Due to the extreme load capacities, high-performing age of sign structures or rotating tions. Using hollow bits in conjunction undercut anchors can be designed for ductile steel capacity, provided equipment. If the anchor is exposed with a high-efficiency particulate air they are embedded deep enough to prevent concrete breakout. to a corrosive environment during

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Table of ACI requirements to qualify the anchor for ductile design in seismic design applications and connections.

ACI 318-14 Design Reference*

ACI 318-19 Design Reference*

Clause 2.3 Clause 17.2.3.4.3 (a)

Property

Requirement

Elongation of tensile tested rod

≥ 14%

Reduction in cross section area

≥ 30%

Steel failure guaranteed with margin

1.2 x nominal steel strength < concrete governed strength

Clause 17.10.5.3 (a) (i)+(ii) (iii)

Free stretch length

≥ 8 times the rod diameter

(iv)

Protection of ductile element against buckling

Sleeve encasing rod

Shape regularity of rod

Entirely threaded or ratio of tensile to yield strength ≥ 1.3

(v) *See Reference code for details.

construction or service life, stainless steel anchors should be strongly considered. Unlike adhesive anchors, undercut anchors are not affected by weather conditions like wetness or extreme temperatures, neither during installation nor in service, and can be immediately loaded after installation.

On top of these technical requirements, there might be specific project and/or legal requirements for parts being made in the United States. For example, this is often the case for military and governmentfunded infrastructure projects.

Product Qualification The importance of product qualification for critical connections should not be taken lightly or for granted. Only undercut anchors independently tested and evaluated to recognized standards should be considered for use. For post-installed mechanical concrete anchors, including undercut anchors, ACI 355.2, Qualification of Post-Installed Mechanical Anchors in Concrete, provides testing conditions and assessment criteria to qualify anchors for their intended use. This is the basis for product evaluation reports issued from recognized approval bodies. The International Code Council

Evaluation Service (ICC ES) is one such approval body that issues Evaluation Service Reports (e.g., ESR-4810).

The extensive test program for an ACI 355.2 qualification assesses the behavior and performance of the anchor regarding its reliability, suitability, and serviceability. The outcome is the design data certified in the product evaluation report, which are product-dependent load capacities. Most tests are static pullout tests on installed anchors under various conditions, e.g., uncracked and cracked concrete and low strength and high strength concrete. More sophisticated tests include cyclic tests assessing the anchor performance under seismic conditions. The anchor is loaded by a recurring and repeating standard seismic protocol to defined load levels simulating earthquake loading. Figure 5 shows the load-displacement diagram of an example shear test. After completion of all prescribed lateral loading cycles, the anchor must still be able to develop sufficient residual load capacity and show proper behavior to achieve qualification for seismic connections.

Design and Software

Post-installed concrete anchors are regulated in the design code ACI 318, Building Code Requirements for Structural Concrete and Commentary, and ACI 349, Code Requirements for Nuclear Safety-Related Concrete Structures, for structural concrete. The resistance of the anchor connection is calculated as the minimum of various failure modes, namely concrete breakout and steel rupture. High-performing undercut anchors do not fail in pullout failure mode, in which case this mode can be neglected. The design strength capacities of the anchors are calculated in tension and shear, which are checked against the applied demand loads derived from static and seismic analysis. Specialized design software (for example, DDA™ from DEWALT: www.DEWALTdesignassist.com) can quickly model and accurately carry out the complex anchor design.

There are several options for the seismic design of anchors in tension. Generally, the most overall economic design is the ductile design considerFigure 5. Example test data: load-displacement hysteresis diagram for ACI 355.2 qualification ation. The ductile design option does not require testing (simulated seismic shear Test Series 19) on an undercut anchor made of ductile 316 SS the increase of the earthquake design load by the material with ¾-inch rod and 1¼-inch outer sleeve diameter, cyclically loaded to 30,000 lbf overstrength factor (W0), which is often as high without fatigue failure.

24 STRUCTURE magazine

The importance of product qualification for critical connections should not be taken lightly or for granted. Only undercut anchors independently tested and evaluated to recognized standards should be considered for use. as 2.5 for the elastic design option. However, additional provisions must be met for a ductile anchor design.

Here is a breakdown of the ductility provisions within ACI 318, as shown in the Table.

• When tested in tensile tests, the rod as the load transferring steel element must show minimum elongation and reduction on the cross-sectional area. If the steel can be classified as ductile, beneficial strength reduction factors may be used to calculate the design strength of the anchor.

• The nominal steel strength of the anchor has to be smaller than any concrete-governed strength by a margin of 20% to safely preclude brittle concrete failure. In the case of undercut anchors, concrete-governed strength is the nominal strength considering concrete breakout and, only if applicable, pullout strength.

• The rod must provide a free stretch length of 8 times the rod diameter or more to ensure sufficient absolute deformation can be generated.

• The rod must be protected by a sleeve against buckling occurring when the rod is loaded in compression during cyclic load reversals.

• The rod must be either entirely threaded, or the tensile strength must be 30% higher than the yield strength to ensure that yielding occurs within the stretch length before failure in the threads.

ACI 318 illustrates two possible details to facilitate sufficient free stretch length, both of which require special measures and additional anchor lengths. Undercut anchors, however, can enable an economical solution for ductility with no extra effort (Figure 6).

Conclusion The robustness of heavy-duty undercut anchors makes them an ideal solution for structural applications in transportation, energy, industrial, port, and waterway projects. Undercut anchors can meet the challenging demands of extreme design situations with a high level of safety by providing high levels of strength, stiffness, and ductility. Ductility plays a vital role in various aspects of both product qualification and design.

Anchors made of ductile steel resist high loads during simulated seismic tests according to the ACI 355.2 standard to qualify for high seismic design strengths documented in Evaluation Service Reports (ESRs).

In addition, undercut anchors meeting stringent geometric and material requirements can qualify for the additional and highly beneficial ductile design option according to the ACI 318 design code.

Specialized design software is available to support the structural engineer in conducting these complex anchor design ductility checks and associated calculations.■ Philipp Mahrenholtz is a Senior Engineering Manager for DEWALT anchoring and fastening systems, responsible for product approvals such as ICC-ES ESRs (philipp.mahrenholtz@sbdinc.com).

Mark Ziegler is Technical Director for DEWALT anchoring and fastening systems. He is actively involved with several working groups of the Concrete and Masonry Anchor Manufacturer’s Association, which address connections and fastening systems in construction (mark.ziegler@sbdinc.com).

Derrick Watkins is Vice President for Structures at SC Solutions. He participates in technical committees of the American Society of Civil Engineers (dwatkins@scsolutions.com).

Figure 6. Details of connections facilitating the anchor stretch length requirement for ductile design according to ACI 318: anchors with rods extending up-air and sitting on an elevated chair (left), anchors with rods extending deep into the concrete to enable the near-surface debonding of the rod from the concrete (middle), and undercut anchors with rods encased in a sleeve (right).

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structural INSPECTIONS Jurisdictional Inspections, Structural Observation, and Special Inspections

By John A. Dal Pino, S.E.

J

urisdictional Inspections, Structural Observation, and Special Inspections are intended to ensure that projects are completed according to the approved construction documents and the relevant building code and standards to protect public safety. But since there are three or perhaps more entities involved in this process with some overlap (picture a Venn diagram), teamwork and clear communication are required. Even then, there are chances that some deficient items might slip through undetected. The contractor is ultimately responsible for the construction in accordance with the approved plans and specifications. Still, if this does not occur, the structural engineer will almost certainly be involved in the fallout. Therefore, structural engineers have a vested interest in understanding the process and what they need to do as the professionals of record to protect themselves and the public.

There are obviously state, regional, and local differences regarding practices and enforcement. While this article attempts to provide a uniform treatment of the subject, it might not apply in all circumstances, places, and times. This article focuses on buildings in California. A future article is planned to address bridges.

Jurisdictional Inspections The Authority Having Jurisdiction (AHJ), the federal government, a state government, a county, a city, or some special government agency, issues the permits for a specific project. For the sake of this article, let’s call this entity the Building Department. Several permits may be issued on any given project covering various items such as stormwater discharge, air quality, the building itself, etc. However, let’s focus on the building permit and assume that this permit covers all aspects of the building construction.

In theory, the Building Department inspects the construction of all buildings within its jurisdiction. Typically, the responsible building inspector stops at a project to ensure the approved plans are on site and the work appears to be going according to those documents. But saying that the building inspector is truly inspecting, as most people understand the term, is inaccurate. There is really no practical way in a given workday that the building inspector has enough time to actually inspect all of the work on projects being performed in their geographic area of responsibility.

On larger projects with more experienced engineers and contractors, the building inspector is more apt to leave the majority of the observing to the structural engineer of record (SEOR) and inspecting to the Special Inspectors (more on them later). Even when the building inspector is called in for an inspection, say before a concrete pour is to be made, the intent of the jurisdiction’s inspection is to observe whether the work looks more or less correct rather than to spot every minor deviation from the drawings. In fairness, how could the inspector ever be expected to know a project as well as the structural engineer who designed it and the special inspectors who are on-site every day?

26 STRUCTURE magazine

Inspection of concrete flat work.

The jurisdiction’s inspectors often take a more active role on smaller commercial projects and are often quite involved in residential projects.

There seems to be a presumption that the contractors involved in those projects are less experienced and that there is a higher probability that the work may not be correct. One would assume that this is based on the jurisdiction’s past project experience and is a prudent course to take. In addition to the review based on the approved construction documents at the job site, the inspector typically reviews any changes to the approved design documented in RFIs issued by the structural.

Some inspectors want to see the RFIs before the work is undertaken, while others settle for seeing the RFIs at some point, so long as the SEOR eventually stamps the documents.

In some areas of the country and on certain types of projects (for example, California hospitals), the Building Department has a far larger staff and budget for detailed inspections. Under those instances, they exert far more influence on the quality and completeness of the completed project. The SEOR must understand and expect this higher level of scrutiny. There is likely to be additional paperwork involved in processing design changes resulting from RFIs and getting details approved that correct deficiencies in the actual construction. On smaller projects discussed above, the Building Department may accept a change documented with nothing more than a sketch stamped by the SEOR. Still, the Building Department is more likely to require a detailed submittal package containing revised drawings and calculations that undergo more than a cursory review on these special projects.

This additional review can result in delays in construction and extra costs if the structural engineering firm is not staffed to provide an immediate response.

Typically, the responsible building inspector stops at a project to ensure the approved plans are on site and the work appears to be going according to those documents.

Regardless of whether the project is large or small, or of a special nature, the building department inspector may have sovereign immunity (derived from British common law doctrine and based on the idea that the King could do no wrong) and is not legally responsible even if the construction is not in compliance with the building code.

That responsibility falls mainly on the contractor and, to some degree, on the owner and the SEOR.

Structural Observation The Building Department requires that construction is in accordance with the provisions of the relevant codes and standards.

The building code (for this discussion, assume the California Building Code, Chapter 17) describes what must be observed during construction by the SEOR or the designated observer (if it is not the SEOR). Before adopting ASCE 7, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, as a national standard document, structural observation requirements were contained in the model building codes (the Uniform

Building Code in California) and then adopted by the individual states. Today, the requirements in each state may still be different.

In addition, cities, counties, and other special agencies may have additional requirements.

Structural observation is typically required for the main elements of the wind or seismic force-resisting systems for certain buildings posing a greater risk to the public. In California, the SEOR can also require that structural observation be performed. However, it is up to the building owner to decide whether to employ the SEOR or hire a qualified third party.

Structural observation should not be confused with inspection.

Structural observation is a serious professional responsibility for the SEOR (or the designated observer). However, the effort expended is based on the concept that it is the contractor’s responsibility to construct the project according to the drawings and the SEOR is only acting in an oversight role. The structural observer visits the site to observe the construction progress and informs the contractor about deviations from the drawings. Inspection, by definition, implies a more detailed review. This is discussed later in the article.

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by the checking of shop drawings (concrete reinforcement, posttensioned concrete reinforcement, structural steel, steel joists and trusses, wood trusses, suspended MEP loads, etc.) and the review of submittals (concrete mix designs, material certifications, product submittals, etc.). Reviewing these documents before a site visit is an essential first step in the structural observation process and ensures that the right pieces and parts arrive at the site. The actual structural observation can then be focused on whether the elements are assembled properly.

Special Inspections

Inspection of shear wall reinforcing.

In practice, the structural observer typically visits the site at important stages of construction to observe the work and note issues that need to be corrected. They should walk the site with the contractor to discuss the issues that need to be addressed and develop a resolution together.

They usually document the site visit in a report issued to the owner and the contractor. Best practices dictate that the issues are tracked in a log until all items have been corrected. The importance of the site visit log becomes more apparent when preparing the close-out letter, as discussed below.

While not the purpose of the site visit, if any unsafe construction practices are observed, the observer should notify the contractor immediately since the contractor is solely responsible for site safety.

If the SEOR is the structural observer, the site visit also offers the SEOR the final opportunity to observe design aspects that may not have been addressed adequately in the project drawings and make (ideally only modest) design changes. It is rare for drawings to be perfectly clear and to contain no errors or omissions. Making design changes during construction may embarrass the SEOR and have cost implications for the owner. Still, the changes are likely to head off larger problems and more costly corrections later when the building does not perform as the owner expected. Some structural engineers believe that the less they see, the better off they are in terms of liability, but most engineers would argue differently.

Structural observation would be a nearly impossible effort if the major elements of the construction were not reviewed in advance

...on certain types of projects ... the Building Department has a far larger staff and budget for detailed inspections. Under those instances, they exert far more influence on the quality and completeness of the completed projects...

28 STRUCTURE magazine

The building code also lists special inspections to be made on particular aspects of the work. In the 1960s and 1970s, several structural failures occurred that were linked to specific inadequacies in the construction.

It was believed that a more detailed review of certain procedures, e.g., welding of structural steel, would better ensure public safety.

Special inspections were born, and the requirements for certain special inspections were added to the building code. The list of inspections has expanded over the years.

As with structural observation, the building code describes what items require special inspection (again assuming the California Building Code, Chapter 17). In addition, individual cities, counties, and other special agencies may have additional requirements.

The term special is a bit of a misnomer. Many engineers would argue that there is nothing special about them and that special was an unfortunate choice of words. Perhaps the word detailed or focused would have been a better choice. And the use of the word inspection could also be questioned since it implies a level of review in excess of observation for something that is ultimately the contractor’s responsibility. Would observation have been a better choice of words since this is more akin to what is actually done in the field? Would it be more accurate and perhaps more understandable to those involved in the construction industry to call what is actually done in the field as focused observation?

Short of a revolution, the industry is probably stuck with the term special inspection. Regardless of the name, the process is an integral part of the overall review process on the owner/SEOR side of the ledger, as opposed to the jurisdictional side. Third-party firms, with the requisite experience and training to do the work properly, usually perform special inspections. Some SEORs have the skills to perform certain special inspections, such as wood shear wall nailing, concrete reinforcement, or concrete placement. Some may even be able to perform the special inspection of welding, high-strength bolting, pile construction, etc. Therefore, there may be a mix of special inspectors, including the SEOR, reviewing the work on any given project.

The owner retains the third-party special inspectors. This benefits the SEOR on a risk management basis as it puts the responsibility for the proper review of certain aspects of the work squarely on the owner’s side of the ledger. On some smaller projects, special inspection is not required, and for projects when the contractor is also the owner, the contractor may legally employ the special inspectors. The regulations preventing the contractor (in most cases) from retaining the special inspector are intended to prevent a serious conflict of interest for the special inspector, who is expected to notify their employer (the contractor) that corrections to the work are required. In some locales, contractors have been known to form their own special inspection firms and have hired them on their projects, creating the condition for self-review and a bad situation.

Suppose the SEOR chooses to perform some of the special inspections. In that case, the SEOR must accept that they are expected to

perform those services that most in the profession consider being at a higher level than mere observation, whether that is, in fact, true or not. The SEOR should also consider whether they are elevating the standard of practice for themselves. If they are competent to perform special inspections on the same site visit during which they perform structural observation, are they also at least partially responsible for errors by the contractor that are not observed and corrected?

The special inspection requirements are usually mandated to be placed on the construction drawings and in a separate letter to the Building Department. It is critical that the SEOR meet with the contractor and the special inspectors at the start of construction to review what inspections need to be performed and when. An extremely difficult situation is created for the owner and the SEOR at the time of project closeout if some required special inspections have not been performed and the work is covered up and impossible to observe.

The Close-Out Letter

John A. Dal Pino is a Principal with FTF Engineering located in San Francisco, California. In addition, he serves as the Chair of the STRUCTURE Editorial Board (jdalpino@ftfengineering.com).

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At the completion of the project, regardless of what the Building Department has done (remember time available and sovereign immunity), most jurisdictions require the SEOR to prepare a closeout letter stating that 1) the project has been constructed in conformance with the construction documents and that 2) it complies with the building code, including changes made during construction. The basis for this statement is the structural observation site visits (this is when the running log plays an important role) and the SEOR’s review of the special inspection reports.

The language in the close-out letter must be chosen carefully to avoid overstating what the SEOR has actually done. For example, it is best to add the term “general” before conformance in the first point noted in the above paragraph since some aspects of the project may not comply with the drawings. In the second point above, it is best to rephrase the statement as it complies with the intent of the building code, including changes made during construction. This is because the SEOR expects, but cannot guarantee, the building will perform as the building code envisions it should in terms of protecting life safety. If the SEOR designs to the minimum requirements of the building code (the legal standard), there is little room for error in the statements, and they may want to avoid an overly strong statement concerning code compliance. If a less experienced staff engineer prepares the close-out letter, the SEOR should provide coaching on proper wording.

As noted above, it is the contractor’s responsibility to construct the building

according to the approved drawings and specifications. The close-out letter has the effect of shifting the legal burden (or at least some of it) back toward the SEOR’s side of the ledger. Therefore, the SEOR’s best risk mitigation actions are to understand what the SEOR is expected to do, do it well and competently, and insist that everyone involved in the construction process does likewise. And to not rely on the Building Department.■

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Pan American Unity

By Erik Kneer, S.E., LEED AP BD+C, Alejandro C. Ramírez-Reivich, Ph.D., Miguel Adrián Michel-Juárez, and Suraj Patel

Securing Historic Artwork through Structural Ingenuity

P

an American Unity is the largest contiguous mural created by Diego Rivera and his last work in the United States. Timothy Pflueger invited Rivera to paint the mural for the Golden Gate International Exposition, which took place on Treasure Island, San Francisco, in 1940. He participated in a special exhibit called Art in Action, where the spectators could experience the artistic process firsthand as muralists, painters, sculptors, and other artists performed their work live.

Figure 1. Geometric layout of mural panels.

30 STRUCTURE magazine

Diego Rivera, The Marriage of the Artistic Expression of the North and of the South on this Continent (Pan American Unity), 1940; © Banco de México Diego Rivera and Frida Kahlo Museums Trust, Mexico D.F. / Artist Rights Society (ARS), New York; Courtesy of City College of San Francisco.

Pflueger and Rivera designed the mural to make Art in Action portable. Unlike the original fresco painting technique, where a building’s wall supports the multiple plaster layers, Pan American Unity is formed by ten panels with metal frames that hold and protect the fresco layers (Figure 1).

Twenty years after the fair ended, in 1961, Rivera’s masterpiece was installed in the City College of San Francisco’s (CCSF) Diego Rivera Theatre. Unfortunately, however, the building was seismically unsafe, and plans to relocate the mural were needed once more. Additional historical documentation can be found on the Diego Rivera Mural Project’s website (riveramural.org).

In 2017, San Francisco Museum of Modern Art (SFMOMA) announced the plans to move Rivera’s masterpiece to the Roberts Family Gallery for a temporary exhibition. The safe relocation of this cultural treasure was complex and challenging. Therefore, CCSF and SFMOMA assembled a team of international experts in conservation,

engineering, science, and art handling to determine how to remove, transport, and reinstall the mural without damaging this priceless work of art. Holmes provided the structural engineering consulting through the design of the supporting structures. A team of experts from The National Autonomous University of Mexico (UNAM) was engaged to research the fresco’s materials and perform large-scale mock-ups to inform the preservation team of relocation methodologies. Atthowe Fine Arts, the art handler and fabricator, advised on removal, handling, and transportation logistics.

Mock-up and Testing Although CCSF’s Diego Rivera Mural Project had done an outstanding job collecting the historical documentation about Pan American Unity, the technical details of the mural and original installation were largely unknown. continued on next page

Figure 2. Extraction of the travel cart from the mural’s original location (left). Travel frame/cart on the truck bed at SFMOMA (middle). Panel to be hoisted vertically into place into the display structure (right).

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...the travel frame could extend no further than 15 feet 9 inches above the flatbed due to the truck’s requirements. Therefore, it was coordinated with the City of San Francisco to remove all overhead electrical lines along the truck’s path of travel.

Figure 3. Existing panel frame (orange), panel mount (gray), travel frame and cart (yellow), and wire rope isolators (gray) on the truck bed.

In June 2019, the UNAM and Holmes’ teams met to plan the removal and relocation. One of the activities involved core-drilling the outer wall of the theatre to uncover, for the first time in almost 60 years, what was behind the iconic arrangement of colors of the pictorial layer. The team took photographs and measured vibrations and the mural dimensions to document its composition and supporting structure.

With this information, the team built precise 3-D models and technical drawings that culminated in creating a full-scale replica of each type of panel. Those models were used to create simulations that allowed quick iterations between handling approaches. Furthermore, the mock-up panels aimed to understand their behavior in the

Figure 4. Top of panel mount connection to display structure (panel mount seat).

32 STRUCTURE magazine

presence of vibrations and impacts. They were also used as test specimens to validate the final de-installation approaches and detect cracks in brittle materials.

The experiments were carried out in the UNAM School of Engineering’s Center for Mechanical Design and Technological Innovation (CDMIT). These experiments included a logarithmic decay test to identify the dynamic parameters of the panels, the comparison of Nelson Stud cutting methods to select the tool that induced the least vibrations to the fresco, hoisting tests to validate the panel manipulation approaches, and resistance tests in which the panels were bent and hit to understand how cracks were made. continued on page 34

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Figure 5. Bottom of panel mount connection to display structure.

After the cart had been removed from the travel frame, it was rolled in front of the display structure. Lifting outriggers were integrated into the top of the display structure to hoist the panels into place...

Structural Sequence The structural design for the relocation and final housing of the murals required: 1) a new panel mount to support each mural panel, 2) a vibration-isolated travel cart and frame to protect the panel during transportation, and 3) a display structure to support the panel mounts in their temporary location at SFMOMA. Each frame was designed and detailed with the intent of a seamless installation to each other. First, each de-installed existing mural panel frame was welded to the panel mount. The panel mount was then connected to the travel cart, which was rigged to the travel frame (described later). Finally, the assembly was picked entirely onto a truck bed. Upon arrival at SFMOMA, the travel cart was removed from the travel frame, rolled into the gallery, and lastly connected to the display structure. See Figure 2 (page 31) for images of the structural sequence from beginning to end. Due to the integration of the frames, it was critical to evaluate the loading at each stage to ensure the controlling demands on the frames and connections were identified.

Panel Mount and Travel Frame The panel mounts were framed out of HSS tube steel and internally braced by steel rods. Spaced around the perimeter of each frame were 34 STRUCTURE magazine

L-angle brackets that connected to the existing mural frame through welded studs.

The Nelson Stud welding technique was considered safe after analyzing its effects on UNAM’s mock-up panels. Heat flow was recorded using a thermographic camera, which allowed the researchers to understand the temperature gradients in the panel. The heat was contained to the surface of the beam that was in direct contact with the welded stud and a small portion of the beam’s web. Furthermore, it was found that the temperatures decreased quickly after the weld was complete.

The panel mounts were modeled using a linear analysis with support conditions at each stage represented by the connections to the travel frame or display structure. Finally, the model extracted the reactions to design the travel frame, display structure, and connections.

The travel frame, strapped onto the truck bed (Figure 3, page 32), navigated the hilly and uneven streets of San Francisco. Additional challenges that needed to be accounted for in protecting the mural during its transportation included overhead mass transit electrical lines, impacts to the panels, dangerous vibrations, and a large wind event.

To add further complication, the travel frame could extend no further than 15 feet 9 inches above the flatbed due to the truck’s requirements.

Therefore, it was coordinated with the City of San Francisco to remove all overhead electrical lines along the truck’s path of travel. However, as an added layer of assurance, the top of the travel frame was built 1 inch above the top of the mural and inset with a steel ladder-like framing for a non-conductive structural cover to be laid over. Due to the maximum height limitation of the travel frame, the design of the travel cart involved very narrow tolerances. The 15-foot-tall existing mural was only 2⅞ inches above the truck bed.

The travel cart and frame were connected through wire rope isolators to mitigate vibration and motion induced by the truck during transportation. The first stage in designing the isolator’s parameters was a vibration test conducted by the UNAM, Atthowe, and SFMOMA teams. Next, the same trucks used for the mural were instrumented to log the accelerations that occurred while traveling on a representative road of San Francisco. Finally, the no-load and full-load scenarios in the trucks were tested. The results were the NLTH forcing functions which were used to tune the stiffness of the isolators and predict their performance during the move. The wire rope isolators effectively reduced the acceleration amplitudes by about 40% to 50% in the fresco compared to the truck’s base.

Upon arrival at SFMOMA, the travel cart was extracted from the frame, and 4-foot-long outriggers (4 total) were slotted into each end of the travel cart for stability. The purpose of the outriggers was an assurance against the event of a powerful gust of wind overturning the entire frame as it is rolled into SFMOMA.

Display Structure Design The temporary display structure at SFMOMA was unable to be anchored through the gallery’s terrazzo floor tiles. Hence, the columns are braced back to the existing structure only through two horizontal truss diaphragms connecting to the building columns instead of anchoring into the slab below.

After the cart had been removed from the travel frame, it was rolled in front of the display structure. Lifting outriggers were integrated into the top of the display structure to hoist the panels into place (Figure 2).

Attached to the display structure at the corresponding top corners of the panel mounts are steel seats for the panel mount HSS member to bear on as the frame is hoisted into place (Figure 4 , page 32). The panel mount was able to slide horizontally on the seat so adjacent panels could run up against each other with minimal gaps in between. The bottom connection between the panel mount and display structure contained a 1-inch vertical slot to provide the vertical adjustability for the upper and lower

panels to be flush (Figure 5 , page 34). Careful attention was paid to the panel mount and display structure connections. Appropriate adjustability and tolerances were provided to allow the fresco panels to achieve a perfect alignment.

Conclusion The Diego Rivera Pan American Unity Mural was successfully installed in SFMOMA’s Roberts Family Gallery on June 28th, 2021 (Figure 6 ).

After years of planning, meetings, site visits, testing, and design iterations, the engineering teams from UNAM and Holmes helped preserve and protect Rivera’s masterpiece for all to enjoy for generations to come.■ Erik Kneer is a Principal at Holmes in San Francisco, CA (erik.kneer@holmes.us).

Alejandro C. Ramírez-Reivich is a Professor in engineering design at UNAM School of Engineering (areivich@unam.mx).

Miguel Michel is a Project Leader at UNAM Center for Mechanical Design and Technological Innovation in Mexico City (migmichel27@gmail.com).

Suraj Patel is a Project Engineer at Holmes in Los Angeles, CA (suraj.patel@holmes.us).

The Diego Rivera Pan American Unity Mural was successfully installed in SFMOMA’s Roberts Family Gallery on June 28th, 2021.

Figure 6. Diego Rivera’s Pan American Unity fresco installed at SFMOMA. Credit Katherine Du Tiel, courtesy SFMOMA. APRIL 2022

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Adaptive Reuse of the Historic Witherspoon Building Part 5: Wrap-Up

By D. Matthew Stuart, P.E., S.E., P.Eng, F.ASCE, F.SEI, A.NAFE

A

four-part series of articles in late 2021 (Part 1, STRUCTURE, September, Part 2 – October, Part 3 – November, Part 4 – December) discussed the adaptive reuse of the Witherspoon Building in Philadelphia, PA. This article wraps up the discussion with information obtained from the Presbyterian Historical Society, including the original 1895 specifications.

Epilogue Because any structural investigation of an existing building should begin with a search for drawings, initial efforts to track down information for the adaptive reuse of the Witherspoon Building, which was initially constructed for the Presbyterian Board and Sabbath School and other various Presbyterian Church groups, began with Princeton University. This was because of the connection of the building’s name and Architect, which were documented in the available National Register of Historic Places, and their relationship to Princeton University. As a result, no building drawings were found, except for some renderings of the façade.

However, even though the author lived near and had dealings with Princeton University and interns from the Princeton Theological Seminary, he was unaware that the Princeton Theological Seminary is not a part of Princeton University until after Parts 1 through 4 of the series of articles had been published. The only current connection between the Seminary and the University is the physical proximity of the adjacent campuses. As a result of this revelation, and because the Presbyterian Church operates the Princeton Theological Seminary, an inquiry to the Presbyterian Historical Society (PHS) resulted in the procurement of several construction photos and the original 1895 specifications. Unfortunately, no drawings other than the façade were available.

The photos included with this epilogue include several construction photos (Figures 23-28) and the cover sheet for the 1895 specifications (Figure 29). (All of the images were provided courtesy of the Presbyterian Historical Society, 425 Lombard Street, Philadelphia, PA 19147-1516.)

As a result of a review of the 1895 specifications, including structural properties for rivets, steel, and clay tile, Pennoni’s analysis was confirmed to be accurate, as shown in the Table.

Table Note 1 Figure 23. South elevation of the building with partially completed façade prior to the construction of what is now the Wells Fargo Building to the west.

Figure 24. Initial erection of the first-floor framing.

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As indicated in the New Floor Openings section of Part 4 of this series, it was determined that the allowable uniform loading of the

Figure 25. Ongoing erection of the first- and second-floor framing.

Table of assumed structural properties confirmed by original specifications.

Structural Component

Original Specs

Assumed by Pennoni’s Analysis

Notes

Rivets

Fy = 30 ksi Maximum Shear Capacity = 9.0 kips

Fy = 30 ksi Maximum Shear Capacity = 7.5 kips

Pennoni’s assumptions based on available historic information from multiple sources

Carnegie Steel

Fy = 28 to 32 ksi

Fy = 32 ksi

Based on coupon sample test of Fy = 32.9 ksi

12-inch Hollow Clay Flat Arch Tile

500 PSF Minimum Load Carrying Capacity

280 PSF Minimum Load Carrying Capacity

See Note 1

arches in the Witherspoon Building was 280 psf. This resulted from lab tests of an exposed tension rod from a demolished portion of the clay tile flat arch floor in conjunction with the calculation of the load-carrying capacity of the tiles based on the methods outlined in the Principals of Tile Engineering Handbook of Design from the early 20th Century. Deducting the existing topping weight, the self-weight of the tile and plaster ceiling resulted in a reserve load carrying capacity of approximately 170 psf, which was almost twice the reserve load-carrying capacity of 100 psf determined for the floor beams. Therefore, based on the author’s experience, and as described below, it was not surprising to determine that the arches would have more capacity than the beams, which is why the load capacity for the adaptive reuse project was based on the beams and not the tile.

Hollow clay tile arch framing systems were initially developed for improved fire resistance and not for reasons associated with structural innovations. That is why it is not unusual to find that the capacity of the supporting steel beams is less than that of the flat arch floor framing.

The primary factor behind the desire to improve fire-resistant construction was the prevalence of devastating city-wide fires during the 19th Century in many different major cities. Therefore, the intent of the original designers of the Witherspoon Building, as well as other similar buildings from the same era, was to ensure that the tiles served two primary purposes: transferring loads to the supporting beams and providing fire protection of Figure 29. Cover sheet for 1895 specifications. the structural steel.■ D. Matthew Stuart is Senior Structural Engineer at Pennoni Associates Inc. in Philadelphia, PA, and Adjunct Professor at the School of Engineering, Widener University in Chester, PA (mstuart@pennoni.com).

Figure 26. A portion of the fourth-floor transfer truss prior to erection.

Figure 27. Erected fourth-floor transfer truss with additional erected floors above.

Figure 28. Partially erected tower and lower façade.

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Leaves Over

Forest Park

By David Fields, P.E., S.E., LEED AP, and Ronald Klemencic, P.E., S.E., Hon. AIA

S

t. Louis’ latest high-rise building, the 36-story 100 Above The Park, brings a new flavor of architecture to the Gateway City (Figure 1). Eight tiers of stacked and undercut floor plans,

each four stories high and shaped like the leaves of a tree, create a form at once organic yet modern. Magnusson Klemencic Associates (MKA) teamed with Studio Gang Architects (SGA) to make this unconventional form a reality by using novel framing solutions following this leaf-like building form. The result embraces Forest Park immediately to the west and provides a direct line-of-sight to the Gateway Arch to the east.

A Form Reflecting the Eye of the Beholder 100 Above the Park has been likened to a corona, icy shards of glass, or glowing canyon walls.

These seemingly contradictory aesthetics all present themselves depending on the viewer’s location and the light in the sky. As imagined by the architect, the core vision for the building form was to create living spaces that were each unique, private, and generous with views. Typical floors have 11 apartments, each of which is a “corner” unit. The extended prow of primary living spaces creates the overall leaf shape of the floorplan. continued on page 40

38 STRUCTURE magazine

Figure 1. 100 Above the Park. Courtesy of Tom Harris Photography.

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Rather than resist this form using a conventional layout of orthogonal columns and walls, the structure works with this geometry, following the concept of veins in a leaf. The primary arrangement of columns, slab reinforcement, and P.T. tendons is set at a 130-degree angle about the centerline of the building.

This creates primary lines of support and load paths that follow the layout of units, leveraging the unique floor plan to stiffen key points around the slab’s perimeter. These locations, found at the narrow points around the slab edge, allow each glazed tip to be column-free while also maximizing slab cantilever distances.

Following this same geometry, the south shear wall demises the 130-degree-oriented units and straightens briefly where east and west segments are coupled above the central corridor (Figure 2).

The unique geometry of the building also brought into question the validity of using ASCE 7, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, prescriptive wind loads, which were developed using regular and prismatic archetypes. As a result, Rowan Williams Davies & Irwin Inc. (RWDI) was retained as the wind consultant for 100 Above the Park, conducting studies on primary building loads, cladding pressures, and “pedestrian” effects (i.e., the wind pressures and turbulence that would be felt regularly on the roof terraces).

Wind tunnel testing was conducted using a 1:300 scale model of the subject building and all buildings of significance within 1,200 feet. Such proximity modeling captures the effects of buffering or buffeting that may result from the turbulent airflow around upwind structures. In the case of 100 Above the Park, adjacent building effects were far less critical than the shape of the building itself.

It is common for wind tunnel testing to reveal crosswind responses for tall towers. This is a phenomenon in which wind blowing in one direction causes rhythmic sway in the perpendicular direction. Cross-wind response can define the governing design loads and create accelerations that are unacceptable to building occupants. When such a response is revealed during

testing, the many remedies typically discussed include reshaping the tower to create roughness along the façade to break the slipstream effect along the sides. 100 Above the Park was several steps ahead of this conversation. The leaf points in plan and the undercut terracing in elevation are precisely the perimeter irregularities that the wind consultant would have recommended for a problematic building. Although the wind tunnel revealed overall wind loads up to 10% greater than ASCE 7 prescriptive loads, cross-wind response was virtually non-existent, with dominant loading clearly parallel to the wind direction (drag loads).

Rather than resist this form using a conventional layout of orthogonal columns and walls, the structure works with this geometry, following the concept of veins in a leaf.

Figure 2. Primary slab load path in green. Shear walls in blue.

40 STRUCTURE magazine

Overhanging Façade – Sawtooth Columns The architectural concept also incorporates extensive private roof terraces (Figure 3). Unlike typical apartment towers where balconies are stacked repetitively, each shading the one below, the terraces at 100 Above the Park are all open to the sky. This is a luxury typically provided only to penthouse units. Adding to the actual 14,500-square-foot roof are an additional 23,000 square feet of occupiable roof terraces created at the top of the four-story tiered stacks.

This form is created where three successive floors extend farther from the building centerline than the floor below.

The fourth floor then returns to the original and smaller leaf shape, and the cycle is repeated. This setback at the top of each four-story stack becomes additional rooftop space around the entire perimeter of the building, yielding each unit approximately 175 square feet of private terrace.

To accommodate the undercut tiers form, the support points around the building perimeter must extend outward along with the building façade. The relationship of column location to terrace cantilever tip is held consistent by the use of sawtooth concrete columns (Figure 4, page 42).

While three faces of the sawtooth columns remain vertical and create consistent demising locations within the apartments, the outer face slopes upward and outward on the same four-story slope as the building perimeter. The vertical sawtooth geometry is present in all 15 tower columns, allowing the tips of the leaf veins to stiffen and support the leaves as they grow and shrink.

Transferring Half a Tower The relatively narrow project site, averaging 130 feet from east to west, necessitates parking and vehicle circulation to reside partially under the tower. A ground floor filled with retail, lobby, and back-of-house space is topped by five stories of modest-slope, on-ramp parking for 250 vehicles.

Parking is most efficient when laid out in an orthogonal pattern. A framing grid of 28- by 60-feet allows three stalls to be arranged off each side of a central drive aisle. However, the leaf-and-vein layout of 32 levels of residential framing above certainly does not conform to this parking ideal.

Zoning constraints limited the wider podium of the building to no higher than six stories. This meant that optimum parking efficiency needed to be established to satisfy the project parking requirements. The structural solution was for every tower column on the eastern side of the building to be transferred over the podium drive aisle (Figure 5, page 42).

Seven concrete transfer beams carry seven tower columns, the largest of which is five feet wide and seven feet deep on spans up to 34 feet. This web of beams delivers the tower column vertical forces, as large as 4,900 kips ultimate, to nine podium columns, the central core, and the south shear wall. The resulting space below is free and clear for optimized parking.

Quality Rock… But Where?

Figure 3. Rooftop terrace at each setback tier. Courtesy of Tom Harris Photography.

The project site is underlain by karstic limestone, which has the characteristic of being highly variable in geometry, APRIL 2022

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including abrupt elevation changes and the potential for voids. The project geotechnical engineer, Geotechnology Inc., based recommendations on auger borings that reached refusal at a depth varying between 55 and 73 feet below ground surface. Rock samples indicated Rock Quality Density (RQD) values as high as 97% for high-quality limestone, generally known as soft rock.

Due to the uncertainty of the bedrock profile across the full site, as well as the common occurrence of a weathered lower-capacity top layer, preliminary recommendations called for the entire building to be supported on drilled piers with tip elevations between 65 and 76 feet below ground surface. These recommendations intended to have the best estimate of the actual subgrade conditions, with verification to occur before placement of any deep foundation elements. At each of the 57 drilled piers, probe holes were drilled as near as possible to the centerline of the pier locations, extending two pier diameters below the tip elevation to verify the absence of voids. Testing the material extracted during this probe program verified that end-bearing values of 80 ksf and side friction values of approximately five ksf could be reliably used.

The final foundation design could not be completed until the probe and testing program was performed. In most instances, the preliminary recommendations adequately covered the variability of the limestone.

However, in select cases, nine-foot-diameter concrete piers needed to be socketed as far as 15 feet into the limestone to achieve the intended capacity.

Summary Figure 4. Building section and sawtooth column.

100 Above the Park embodies exterior geometries and interior residential spaces that have never before been created. The concrete frame, hidden within and below this new gem for the city of St. Louis, challenges how a residential building can be shaped, designed, and built.

Without straying from the standard palette of structural elements of a high-rise concrete building – flat slabs, shear walls, (partially) rectangular columns, and drilled piers – 100 Above the Park delivers on its goal of immersing its residents in the canopy of Forest Park while creating an iconic westward expansion of the St. Louis skyline.■ David Fields is a Senior Principal at Magnusson Klemencic Associates and leader of the firm’s Residential Specialist Group. (dfields@mka.com).

Figure 5. Tower column transfers over the garage.

Ron Klemencic is the Chairman and C.E.O. of Magnusson Klemencic Associates and a past 5-year Chairman of the Council on Tall Buildings and Urban Habitat (rklemencic@mka.com).

Project Team Owner: Mac Properties (Chicago) SER: Magnusson Klemencic Associates (Seattle/Chicago) Architect: Studio Gang Architects (Chicago) General Contractor: Clayco (St. Louis)

42 STRUCTURE magazine

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just the FAQs FAQs on ASCE Standards What You Always Wanted to Ask

By Laura Champion, P.E., F.SEI, F.ASCE, and Jennifer Goupil, P.E., F.SEI, M.ASCE

T

hank you for all your comments in response to the launch of this new quarterly column in the January 2022 issue. These articles address some of the questions received about structural standards developed by the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE), including the recently released ASCE 7-22. In addition, questions received from engineers, building officials, and other design professionals are often considered to develop future editions. Following are some questions received by SEI and responses to clarify the provisions.

ASCE/SEI 7: Minimum Design Loads and Associated Criteria for Buildings and Other Structures When Can You Use the 0.7 Multiplier in Seismic Category A? Q: Is the resultant value in Equation 1.4-1, referenced in ASCE/SEI 7-16 Section 1.4.2, an allowable stress or strength level force?

A: The force defined by Equation 1.4-1, Fx = 0.01Wx, (where Fx is the design lateral force applied at story x) and Wx is the portion of the total dead load of the structures, D, located or assigned to level x is a strength level force.

Because Equation 1.4-1 is required to be used in the lateral force design of structures assigned to Seismic Design Category A, it is permitted to be multiplied by 0.7 when used with allowable stress design (ASD) load combinations of Section 2.4. Because many nonbuilding structures designed under Chapter 15 are designed using ASD procedures, it would be common to see these nonbuilding structures use the force from Equation 1.4-1 multiplied by 0.7 as the minimum base shear. While this question was specific to SDC A, this applies to all Seismic Design Categories. The 0.01 factor found in Equation 1.4-1 is the same value as the 0.01 minimum limit for the Cs value determined from Equation 12.8-5. For nonbuilding structures falling in Seismic Design Categories B through F, additional minimum base shear values are specified in Section 15.4.1 and may exceed the 0.01 value.

What are the Requirements for Adhesives in the Seismic Design of Access Floors?

Q: What is the intent in Section 13.5.7 of ASCE/SEI 7-16 with Rp = 1.5 for access floors and the fact that adhesive is explicitly ruled out

44 STRUCTURE magazine

for special access floors where Rp = 2.5? Does Section 13.5.7.1 allow the use of a qualified adhesive, whereas 13.5.7.2 prohibits it because a larger response modification factor is allowed? Therefore, can access floor base plates be attached with a qualified adhesive when using an Rp = 1.5 per Section 13.5.7.1?

A: ASCE/SEI 7-16 Section 13.5.7.1 states that adhesives (bonding or gluing of the pedestal to the supporting structure) can be used for the attachment of ordinary access floors (Rp = 1.5) to the supporting structure. Per ASCE/SEI 7-16 Section 13.5.7.2 Item 2, adhesives are prohibited for special access floors (Rp = 2.5). However, per Section 13.5.7.2 Item 1, adhesive anchors that comply with Chapter 17 of ACI 318 are permitted for special access floors. Note that there is a difference between adhesives and adhesive anchors.

What is the Definition of a Rooftop Structure when Determining Wind Loads on Buildings?

Q: Are penthouse structures that are habitable spaces included in the definition of rooftop structures as defined in ASCE/SEI 7-16 Section 29.5? Is an elevated platform supporting an air handling unit (AHU) included in the definition of rooftop structures?

A: Section 29.5 is used to determine the wind pressures on parapets. The provisions of Section 29.4 should be used to determine the pressures on rooftop structures (penthouses) and rooftop mechanical units.

What Building Height Should be Used for Determining Wind Loads when Designing a Roof Canopy?

Q: In equation 30.11-1, is qh defined at the top of the canopy or the roof structure?

A: Per Section 30.11, the velocity pressure qh is determined from Section 26.10 and is evaluated at the mean height of the roof or the eave height for a flat roof structure. This does not change when determining the design pressures on a canopy.

Where is Vehicle Barrier Loading Applied?

Q: What is the application area regarding vehicle barrier loads referenced in ASCE/SEI 7-16 Section 4.5.3? Where is the 6000-pound impact load applied?

A: Section 4.5.3 of ASCE/SEI 7-16 does not specify the required height of the vehicle barrier system itself. However, this section specifies a range of heights for which the horizontal load must be applied.

In addition, a maximum area is specified over which the load can be applied. A smaller area can be used. For whichever size area is assumed, up to the maximum area of 12 by 12 inches, it would be

rational to center the load area at both the minimum and maximum height to determine the maximum load effects. Note, be sure to confirm if your project needs to meet additional minimum vehicle barrier heights as required in the International Building Code.

ASCE/SEI 24: Flood Resistant Design and Construction Is ASCE/SEI 24-14 Applicable for Non-Building Structures?

Q: What is the intent of Section 1.5.5 regarding wood-constructed boardwalk ramp connections? Does this section only allow bolted connections and preclude the use of screw connections and/or mechanical connections, such as pre-engineered manufacturer connections?

A: ASCE/SEI 24-14 standard requirements are intended principally for buildings and not for docks, piers, boardwalks, etc. These other structures may still be subject to floodplain management or local jurisdiction requirements (since the structures are considered “development” under floodplain regulations). Section 1.5.5 was written with buildings in mind, specifically, the attachment of the elevated portion of the building to the foundation with bolted connections. Other attachment methods should be acceptable for nonbuilding construction if the design objectives are satisfied to resist all loads. This includes, but is not limited to, lateral and vertical load and requirements from the authority having jurisdiction for flood plan regulations to prevent flotation, collapse, and displacement.

Is Parking Allowed Below Grade Where Flood-Resistant Design and Construction are Required?

If you have a question you want to be considered in a future issue, send it to sei@asce.org with FAQ in the subject line.

Visit asce.org/sei to learn more about ASCE/SEI Standards.

This article’s information is provided for general informational purposes only and is not intended in any fashion to be a substitute for professional consultation. Information provided does not constitute a formal interpretation of the standard. Under no circumstances does ASCE/SEI, its affiliates, officers, directors, employees, or volunteers warrant the completeness, accuracy, or relevancy of any information or advice provided herein or its usefulness for any particular purpose.

ASCE/SEI, its affiliates, officers, directors, employees, and volunteers expressly disclaim any and all responsibility for any liability, loss, or damage that you may cause or incur in reliance on any information or advice provided herein.

Laura Champion is a Managing Director of the Structural Engineering Institute and Global Partnerships at the American Society of Civil Engineers.

Jennifer Goupil is Senior Manager of Codes and Standards and Technical Activities at the Structural Engineering Institute of the American Society of Civil Engineers.

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Q: Does Chapter 9 Section 9.4.1 of ASCE/SEI 24-14 allow belowgrade parking? Is ASCE 24’s definition of below-grade parking in conflict with the National Flood Insurance Program (NFIP) regulations?

A: ASCE/SEI 24-14 provisions related to below-grade areas are patterned after NFIP regulations in 44CFR 60.3 for Flood Plain management. This has been the case since the first (1998) edition of ASCE 24. Below-grade parking requirements in ASCE/SEI 24-14 are referenced in several additional chapters. Chapter 1 applies to all flood hazard zones, Chapters 2 and 4 provide the basic requirements, depending on the flood hazard zone, and Chapters 5-9 provide additional requirements.

Section 9.4.1 states, “Attached garages and accessory storage structures are permitted below the elevations specified in Table 2-1, provided the walls meet the requirements of Section 2.7” and, “Attached garages, carports, and accessory storage structures are permitted below the elevation specified in Table 4-1, provided the walls comply with the requirements of Section 4.6.” The elevations referenced are for the Design Flood Elevation (DFE), and the intent is to allow construction between the DFE and grade. Sections 2.7 and 4.6 define flood openings in walls or breakaway walls between the DFE and grade, respectively. An exception to the flood opening requirement of Section 2.7 is mentioned in Section 2.3.

Section 9.4 states, “The floors of garages, carports, and accessory storage structures shall be at or above grade on at least one side.”

However, a “basement” is defined as “the portion of a structure having its lowest floor below ground level on all sides.” And, basements are not permitted in residential structures (explained in Section C.2.3 and C.9.4.1) or in any structure (residential or non-residential) in a Coastal High Hazard Area, Coastal A Zone, or other high-risk flood hazard area. Basements are permitted below non-residential structures (and non-residential portions of mixed-use structures), but the enclosed area must be dry-floodproofed to the DFE. Flood insurance credit for floodproofing may require dryfloodproofing at a higher elevation.■

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guest COLUMN

The REACH at the Kennedy Center

By Yvonne Nelson, P.E.

T

he REACH, a 72,000-square-foot expansion of the John F. Kennedy Center for the Performing Arts in Washington, DC, received the Overall Excellence Award in the 2020 American Concrete Institute’s (ACI) Excellence in Concrete Construction Awards program. The REACH is comprised of three prominent pavilions connected by studios, classrooms, and a parking structure below an interwoven green roof. This article provides insights into the execution of some of the project’s outstanding features.

Overview At first glance, the three pavilions of the REACH – the Welcome, Skylight, and River Pavilions – appear to be three independent structures planted within a terraced landscape covering about 5 acres.

However, upon further inspection, the terrace is a green roof above two levels of interconnected corridors, performance spaces, practice halls, and a parking structure below.

The green roof is supported by a post-tensioned The conical slab comprises a gray structural concrete layer, a roofing membrane, and a white slab incorporating Cobiax void formers to minimize exterior layer. The form was achieved using both formed concrete and shotcrete placements. dead load. Void formers significantly reduce the nonstructural concrete from the middle of the slab, effectively horizontal and vertical forces. Additionally, the distinctive shape decreasing the dead load and maximizing the overall slab span. required a dampproof membrane between two layers of concrete During the construction phase, some voids had to be eliminated and an architectural concrete finish. to ensure enough anchorage capacity for formwork loads without impacting the original design parameters for span and deflection.

The Skylight Pavilion The post-tensioning was also reviewed to ensure adequacy for construction loads.

The Skylight Pavilion is the project’s crowning achievement, featuring Throughout, the REACH complex features a variety of exposed an expansive atrium and a 145-foot-long, 42-foot-tall, asymmetrically concrete finishes and structural and architectural components. A curved wall. Overlooking the Potomac River, natural light from 2-foot-thick mat supports the base of the structure on existing cais- the slotted roof and large, curved windows in the south-facing wall sons and new H-piles, whereas the level above comprises solid slabs create an elegant gathering space. On the north face of the pavilion, (both one-way and two-way), two-way voided slabs, and two-way a vertical exterior wall serves as a large outdoor projection screen for post-tensioned voided slabs. The green roof structure primarily audiences to enjoy movies and simulcasts. contains one-way post-tensioned voided slabs except for the roof The Skylight Pavilion is a very complex structure and the largest of structure over Studio K, which is a post-tensioned one-way slab the three pavilions. The “wave-shaped” wall is curved and warped, with a sawtooth soffit. with each spline unique. The base includes the large opening for the curved window; at the top, it is loaded and laterally supported by a reinforced concrete roof that is slotted for skylights at each end.

The Welcome Pavilion The base of the curved wall was complicated even further because it The Welcome Pavilion provides a spacious main entry and visitor extended below the historic flood elevation and required a membrane lobby for the REACH. Prominent features include 40-foot-tall, sandwiched between two placements of concrete to prevent flooding board-formed concrete walls, cantilevered concrete stairs, a conical of the enclosed space. The high mass, complex geometry, and key roof over the uppermost entryway, and a vaulted lower ceiling near design features of the curved wall provided tremendous challenges the lower entryway. for the construction team.

The upper roof slab and beams of the Welcome Pavilion were After being awarded the formwork contract, PERI performed a cast on elevated forms supported by shoring up to 40 feet tall. complete analysis of a 3-D model of the wall formwork to assess The conical section of the upper roof created many unique design constructability. While the design team had specified that the wall and construction challenges because the shoring had to resist both be pumped from the bottom and placed monolithically, PERI was 46 STRUCTURE magazine

concerned about the high pressures from both the self-consolidating would have allowed the back of each formwork panel to be fabriconcrete (SCC) and the corresponding resultant forces in the two cated perpendicular to the ties, thus reducing the need for custom stop-end systems at each end of the wall. In lieu of a monolithic detailing of the stiffbacks. However, the design team rejected this placement, PERI first considered a two-lift placement and calculated concept, as the desired aesthetic instead required each form tie to the construction loads to be transferred to the voided slab of the be perpendicular to the face of the finished concrete. This condipermanent structure below. tion added to the engineering costs for the formwork because it The engineer of record (EOR) analyzed these anticipated formwork required that each stiffback was chorded between ties rather than reactions, which showed that the permanent structure below could simply running vertically for the length of the panel. The articunot support the construction loads resulting from a full-length, two- lated stiffbacks required more intricate bracing for each panel, so lift placement. To avoid overloading the supporting structure, PERI the complexity of the bracing calculations and associated details recommended a three-lift placement. This increase in the number increased significantly. continued on next page of placements significantly impacted the material and labor costs. Although the outside formwork could be left in place for all placements regardless of the number, the change from two-lift construction to three-lift construction added about 50% to the quantity of forms and attachments needed for both the outside and inside formwork assemblies. This modification had the greatest impact on the inside face of the formwork, as each lift of interior forming and falsework had to be stripped and the wall braced before the subsequent lift of forming and falsework could be set. Because each spline of the wall is unique, each panel of formwork also had to be unique, and each connection and bracing condition had to be designed and detailed independently.

Early on, formwork designers anticipated that the curved wall would need Komponent delivers in design, construction, and in-service to be tied into and placed with end wall with quality, efficiency, and cost savings! returns to facilitate forming the bulkheads at each end and prevent cracking ADVANTAGES in the curved wall. Because it could not  Improves structural performance  Up to 60% greater abrasion resistance be reasonably assumed that the wall  Maximizes design versatility  Increases dimensional stability and durability was self-supporting until the roof slab  Maximizes joint spacing  Enhances compressive and flexural strengths was tied into the curved wall, it was  Speeds time to completion  Minimizes creep and moment imperative that much of the formwork  Reduces mobilization & formwork  Increases density and lowers permeability and shoring remain in place and that  Reduces project costs  Prevents curling and drying shrinkage cracking bracing be provided for the stripped inside face of the wall until the roof slab had been placed and gained strength.

This requirement added substantially to the rental costs for the formwork and shoring for the outside face, plus the formwork, shoring, and labor for the inside face. Ultimately, leaving the inside face of the formwork and shoring in place also complicated the shoring design for the roof slab and sequencing for the entire operation, including transfer beams bisecting the roof at the Shrinkage-Compensating Concrete & Grout Solutions top of the wall and supporting the slab edge at the skylights.

As the formwork design was develCTScement.com  800.929.3030 oped, PERI initially assumed that the form ties would be horizontal. This

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APRIL 2022

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These channels extended down one level and were anchored into the foundation wall below.

Studio K and the Justice Forum

The River Pavilion Located at the southernmost end of the Kennedy Center campus, the River Pavilion houses a café with views of the Potomac River and a new pedestrian bridge. The smallest of the three aboveground structures in the REACH, the River Pavilion continues the themes of the other structures with a conical roof slab and an asymmetrical curved wall with board-formed finishes.

As with the curved wall of the Skylight Pavilion, the curved wall of the River Pavilion was constructed with white SCC and tongue-and-groove board form lining. While this shorter wall was constructed in two lifts rather than three, the wall’s proximity to the cut for the Rock Creek Parkway made it necessary to anchor the formwork bracing only on the interior side. Because there was no slab to support the exterior forms, heavy channel strongbacks were used to resist the loads of the asymmetrical placement.

Over 11,500 square feet of the interior wall surfaces of the REACH were constructed with a crinkle concrete texture featuring 3-inch-deep random folds and angles designed to break up sound waves and avoid reverberation between parallel walls. Studio K is the largest space with this feature, followed by the Justice Forum auditorium.

Steven Holl, the REACH’s architect, came up with the crinkle concrete’s unique form pattern by bending metal sheets.

Subsequently, he worked with Fitzgerald Formliners and Form Services, Inc., to transfer the irregular pleated textures to 4- × 10-foot elastomeric form liners. During the liner fabrication process, special care was taken to ensure that the folds in the pattern would not inhibit stripping of the formwork.

The crinkle concrete walls are not only decorative and sound-enhancing but also serve as primary structure supports for the portion of the buildings where they reside. The exteriors of the Studio K and Justice Forum auditorium walls were constructed with board-formed surfaces. The forms included proprietary crack inducers to minimize the visual impact of contraction joints. The 22-foot-tall walls were placed full height with a gray, high-slump concrete mixture.

Challenges The REACH’s unique features demanded extra attention from the planning phase through execution. A few of these features are included below.

Much of the project comprises exposed concrete walls that define the aesthetics of the REACH. The exterior walls were constructed

ACI Excellence in Concrete Construction Awards The American Concrete Institute (ACI) Excellence in Concrete Construction Awards program offers a new self-nomination option, making it easier for companies and organizations to achieve recognition for their work.

ACI’s Excellence in Concrete Construction Awards recognize projects in seven categories that exhibit innovation and complexity and showcase the value provided by concrete as a material.

For companies or organizations that do not have an ACI chapter or partner award program available in their area, self-nomination is a solution. “The option for self-nomination is a marvelous enhancement of the awards program, ensuring recognition for the best of what’s out there,” said Michael J. Paul, former Chair of the International Project Awards Committee (IPAC).

Entries for the 2022 ACI Excellence in Concrete Construction Awards are due Friday, April 29, 2022. To learn more about the Awards Program or self-nominate a project, visit https://bit.ly/36G93SD.

48 STRUCTURE magazine

First place winners for the decorative concrete (Qatar National Library) and low-rise buildings (Hamad Port Project- Design and Build Visitors Centre) category for the 2019 ACI Excellence in Concrete Construction Awards.

using self-consolidating white concrete with tongue-and-groove board form on the outside face. The finish on the inside face of the exterior walls varies. Many walls were placed with board form on one side of the wall and a “crinkle” liner on the other. These varied finishes created challenges with sequencing and stripping the formwork.

Further, tie holes had to be laid out in a regular pattern – aligning both vertically and horizontally. This was difficult to achieve on irregular shapes and at corners, as they resulted in high reaction loads.

Construction joint locations were limited to wall corners, resulting in the need for large quantities of formwork and form liner materials, particularly boards used to create the board-formed finishes.

The length, height, and asymmetrical shape of the curved wall on the Skylight Pavilion required careful analysis of its effects on the void slab below. Placement height and formwork pressure affected construction loads, labor and formwork costs, and schedule. The position of the curved and asymmetrical wall in the River Pavilion created another set of challenges, as a steep embankment made it impossible to brace the formwork on the concave side of the wall.

The shapes of mockup panel assemblies were verified at the fabrication plant and job site using laser scanners. When formwork was placed for the feature wall of the Skylight Pavilion, a laser scan was also performed to ensure proper geometry. A final scan was performed on the cast concrete to verify that the formwork had not moved during placement.

Demonstrated Excellence The REACH at the Kennedy Center excels at many levels. Outside, its brilliant and sculptural forms create striking icons both day and night. Inside, its textured walls provide aesthetic features as well as acoustic functions. It truly is a demonstration of excellence in concrete construction.■ Portions of this article were previously published in Concrete International, March 2021. It is reprinted with permission.

Yvonne Nelson is a member of ACI Committees 301, Specifications for Concrete Construction, and 347, Formwork for Concrete, as well as Chair of ACI Subcommittee 301-B, Formwork and Formwork Accessories – Section 2. Nelson was the Formwork Engineering Manager for the concrete subcontractor, The Lane Construction Corporation, during the construction of the REACH.

Project Team Architect: Steven Holl Architects Structural Engineer: Silman Façade Consultant: Thornton Tomasetti General Contractor: Whiting-Turner Concrete Contractor: The Lane Construction Corp.

We are excited to see a lessening of the pandemic’s impact on our authors and advertisers and expect to have more normal page counts in the future.

As a result, we have taken the opportunity to showcase the work that other organizations do in supporting SEs by reinvigorating our Guest Column program. It is a pleasure to have these organizations add to STRUCTURE’s knowledge base. If your organization would like to submit an article proposal, please contact Chair@STRUCTUREmag.org.

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APRIL 2022

49

construction ISSUES Evaluation and Mitigation of Risks from Adjacent Construction

By Antonio De Luca, Ph.D., P.E., S.E., and Meeok Kim, P.E., Ph.D.

B

uilding density in urban settings is growing across the United States. Owners, developers, contractors, and engineers are becoming more sensitive to the detrimental effects of construction activities on neighboring structures. Construction activities disturb the soil in the construction zone and its immediate surroundings. The new construction team should identify all potential risks associated with the planned construction activities before the commencement of the site work.

Activities such as excavations, dewatering, pile driving, and/or drilling always cause, to some extent, permanent deformations in the surrounding soils. The magnitude of such deformations is a function of several variables. Characterizing the heterogeneity of the soil conditions and evaluating the effects of changing the groundwater table are essential to quantify all risks associated with planned construction activities.

For a successful project, the new construction team should establish a mitigation strategy against identified risks and establish effective construction and site monitoring plans. When the risk of damaging adjacent structures is found to be moderate or high, a detailed evaluation of the effects of construction activities, such as through soil-structure interaction analyses, is critical.

Besides the potential for damaging adjacent buildings, construction activities also generate noise and ground vibrations that are often felt as disturbing by the adjacent buildings’ occupants. While it is understood that construction-induced vibrations are generally unlikely to cause severe structural damage, it is best practice for the new construction team to predict what levels of ground vibrations can be experienced by the neighbors and identify what structural components and architectural features are most at risk prior to the start of the project.

Construction Activities and Associated Risks Excavations are the removal of soil from a site to reach a lower working surface to construct a new foundation. The soil removal causes the loss of lateral restraint for the adjacent soils, inducing ground settlements. The edges of the excavation require a support-of-excavation (or earth retention) system to limit the severity of these settlements.

Typical support-of-excavation (SOE) systems include, for example, sheet piling, soldier piles and lagging, secant or tangent piles, soil mix walls, and slurry walls. Anchoring or bracing is often used to provide

Figure 1. Soil-structure interaction allows for realistic estimation and accurate assessment of soil deformations and below-grade structure behavior. The plot shows the comparison between the predicted lateral displacements of a support for an excavation system and the in-field measured values.

50 STRUCTURE magazine

additional lateral support to the SOE and increase lateral restraint to the soil retained. Underpinning is a) often required during excavation to support shallow foundations of adjacent buildings.

The extent and amount of ground deformation should be evaluated before the start of construction to determine what buildings fall in the area of influence of the excavation and to what degree. The most common factors to take into account for the selection of the means and methods of excavation and the design of the support of excavation measures include, but are not limited to: • Heterogeneity of the soil conditions • Engineering characteristics of the soil • Presence of the groundwater • Rigidity of the SOE systems • Stability of the excavation • Potential for groundwater seepage through or around the SOE barrier b) • Presence of a cavity or artisan water • Presence of buried structures Analyzing the amount of deformation and stress changes caused by excavations in the surrounding soils involves complex mechanics. Several empirical or semi-empirical models are available in the literature. These models are based on selected case studies and generally lead to conservative estimates of ground deformations. They are most useful for a preliminary assessment when a rough order of magnitude of the excavation-induced ground movements is acceptable.

Once the area of influence of the excavation is determined and the anticipated ground displacement is evaluated, the excavations’ effects on neighboring Figure 2. Numerical model a) and color-coded contour plot b) of the calculated soil deformations surrounding an existing utility pipe due to the construction of a new deep buildings should be assessed. This is typically done foundation and lateral loadings. by assigning the predicted ground displacement to the adjacent buildings’ foundations and calculating the induced strain levels. Finite element analysis is often used to dewatering activity to occur, as much as possible, within the limit assess the building’s response to these induced settlements. A more of the excavation footprint. The permeability of the soils, the approximate approach may involve using empirical damage criteria type of SOE system, the duration of dewatering, etc., are critical based on available field data that account for the estimated differential variables to consider in evaluating the influence zone and degree settlements and the type of construction. of dewatering.

Construction dewatering is typically required when the bottom of Pile driving advances prefabricated piles in the foundation soil the excavation is at a lower elevation than the groundwater table. It by displacing the soil near the tip of the pile. The source of energy temporarily pumps out groundwater from the bottom of the excava- for the pile advancement may be from the impact of a deadweight tion and may be required to keep the excavation dry and/or to limit onto the top of the pile or from a vibratory hammer. Given the buoyant forces on the new structure being built. Dewatering can be dynamic nature of this construction activity, ground vibrations controlled within the excavation limit if an SOE can be installed as are the inevitable effect of pile driving and are often the only side a groundwater cut-off barrier in conjunction with less permeable soil effect. In some sandy soils, however, pile driving may cause vibrastrata. However, dewatering may be needed to lower the groundwater tory settlements. For example, studies have shown that “narrowly table to below the bottom of excavation when no clear groundwater graded, single-sized, clean sands with relative densities (corrected for cut-off mechanism can be established. In this case, it affects a much depth) of 50 to 55% (or less)” are prone to vibratory densification, wider area than the excavation footprint. as reported by Dowding (2001).

Dewatering also affects the design and construction of the SOE system. When dewatering is planned, and the SOE acts as a groundConstruction Vibrations water cut-off barrier, the SOE system is to resist the hydrostatic pressure from the groundwater on the outside of the excavation Construction vibrations are elastic waves generated from conand is to be engineered to prevent excessive seepage. As a side struction activities and propagate through the ground. They are effect, dewatering may cause significant soil settlements outside the most common and most noticeable side effects of constructhe project area. For this reason, it is important to design the tion activities. Their dominant frequencies are in the range of

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Numerous models are available in the literature that propose safe, not-to-exceed vibration thresholds during construction. In the United States, the first criteria were developed in the late 1970s and early 1980s and were based on research studies focused on the effects on adjacent low-rise buildings caused by mine blasting. They were derived from a statistical analysis of damage data from residential buildings subjected to ground vibrations due to blast activities. The buildings were characterized by a natural frequency ranging between 4 and 11 Hz and were finished with drywall or plaster over wood lath support. The Figure 3. Numerical model and color-coded contour plot of the calculated soil deformations underneath study results indicated that a peak particle a pile-supported embankment for rising water level. velocity of 0.5 inches/second or lesser had a low probability of causing threshold damage, approximately 10 to 30 hertz (Hz). For this reason, some low-rise which was defined as the “loosening of paint; small plaster cracks buildings, building components, and architectural finishes may be at joints between construction elements; lengthening of old cracks” prone to resonant vibrations. The amplitude of the construction as reported by Siskind et al. (1980). vibrations depends on the energy imparted by the construction Available data-driven criteria are often misinterpreted. Exceedance machinery. Impact pile drivers generally cause the largest vibration of such thresholds does not necessarily correlate to the appearance of amplitudes, followed by vibratory pile drivers, vibratory rollers, structural damage. Building occupants tend to pay closer attention hoe rams, and jackhammers. to the condition of their buildings after they experience construction Construction vibrations propagate as three-dimensional waves. Most vibrations; that is the time they often notice pre-existing cracks they of their energy is transmitted through surface waves. The dominant were not aware of. surface waves are the Rayleigh or R-waves. Their amplitude decays as they propagate away from the source. Another contributor to the Soil-Structure Interaction attenuation of the vibration amplitudes is the material damping of the soils the vibrations propagate through.

While adjacent buildings undergo deformations due to the excaThe effects of construction vibrations on structures depend on vation-induced movements, the buildings themselves affect the their source, the distance from the source, the vibration frequency, soil displacements directly beneath them. It is essential to consider amplitude and duration, the type and condition of both structure how soil and structure interact to better estimate permanent (settleand foundation soil, etc. Generally, the problem is simplified by ments due to loss of lateral support) and transient (vibrations) adopting the peak particle velocity (PPV) of the vibrating ground soil deformations due to adjacent construction. Advanced finite as the sole parameter to assess or monitor for potential vibration element analyses are required to do so effectively. In the majority damage to adjacent buildings. The peak velocity values in damage of the cases, simplified analyses in conjunction with empirical esticriteria can be defined as the maximum resultant of the three mations lead to results that are good enough to meet the project’s velocity components or the resultant of the maximum velocities requirements. Instead, super-tall buildings, high-profile structures, in the three directions. settlement-sensitive structures, and/or large sites with questionable

Figure 4. Continuous monitoring of vibrations adjacent to a shallow excavation.

52 STRUCTURE magazine

Figure 5. Continuous monitoring of construction vibrations in the building’s basement adjacent to a stone wall.

Figure 6. Continuous monitoring of movement across a pre-existing crack in brick masonry façade and overall building movement.

soils may justify the extra level of accuracy in predicting the effects of construction activities. Examples of soil-structure interaction analyses are shown in Figure 1 (page 50), Figure 2 (page 51), and Figure 3.

Best Practices

assessing the potential damage, and designing any necessary damage mitigation measures.

It is also important to develop a monitoring program that includes both vibrations and settlements as needed at the adjacent structures (or property lines near the adjacent structures) during construction. The monitoring can be implemented by deploying seismographs, tiltmeters, inclinometers, automated or manual surveying tools, piezometers, etc.

The monitoring program should be developed based on the level of damage risk, the degree of concern, and the expectation for frequency and accuracy of the data. Examples of monitoring measures are shown in Figures 4 thru 8.

The engineer in charge of the monitoring program should review the monitoring data as frequently as required by project needs to identify potential signs of movement to modify construction operations to avoid significant damage. The monitoring results should be reported and explained to the construction team and the neighbors in a timely manner so that the effects of the ongoing construction operations are clear to all parties involved.

Buildings experience cracks due to thermal expansions and contractions, changes in material properties through deterioration, and fatigue over time.

In addition, buildings can develop cracks due to structural overloading, foundation settlements, construction defects, etc. Differentiating between Figure 7. Continuous monitoring of brick façade movement with tiltmeter. wear-and-tear cracks and vibrationinduced cracks is often challenging without a thorough documentation of the building’s pre-existing conditions at the start of the adjacent construction project.

The pre-construction evaluation should identify all of the structures that may be at risk of being damaged by the adjacent construction.

These structures should be surveyed to document their pre-existing conditions, focusing on both structural and architectural features. The presence of historic buildings and buildings hosting artwork of value and/or sensitive equipment that may be negatively affected by conConclusions struction vibrations should also be recognized. Pilot studies can also Before a new construction project be conducted to understand how begins, it is good practice that the vibrations propagate (and attenuate) construction team educates adjaaround the specific construction site cent owners and residents about the and the foreseen vibration levels. type of construction activities that The pilot study generally consists of will take place, explain the anticistriking the ground with a backhoe pated vibration and noise levels that and recording the induced vibrations may be experienced, and outline at a number of locations around the Figure 8. Automated system for the survey of a bridge. what measures will be implemented construction site. to minimize the risk of damage.

Based on the information collected during the pre-construction Experience proves that informing neighbors about the evaluation, the geotechnical reports for the new construction and project and the potential effects of construction activities is any available foundation plans of the existing, adjacent structures often crucial to avoid public complaints and legal disputes.■ should be reviewed and analyzed to determine the damage risk from any planned construction activities. The construction tasks that may References are included in the online PDF version pose a risk to the integrity of adjacent structures should be identified, of the article at STRUCTUREmag.org. means and methods of construction with acceptable performance and less impact should be evaluated, and temporary or permanent Antonio De Luca is Senior Project Director with Thornton Tomasetti solutions to mitigate the effects of construction activities should be (adeluca@thorntontomasetti.com). analyzed. The adjacent buildings should be assessed for potential construction-induced damage. This assessment should consist of Meeok Kim is an Associate Principal with Thornton Tomasetti calculating the predicted construction-induced ground deforma(mkim@thorntontomasetti.com). tions, applying this deformation field to the building structures,

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engineer's NOTEBOOK Simple Capacity Checks for Commonly Used Steel Sections

By Hee Yang Ng, MIStructE, C.Eng, P.E.

S

teel sections can be used as columns, beams, and struts to resist axial forces (tension/compression), moments, and shears. They may come in the form of open sections (such as ASTM A992 – Wide Flange W-shapes beams/S-shapes beams) or closed sections (such as Hollow Structural Section [HSS] tubing).

Physical dimensions and weights designate most steel sections, i.e., American Wide flange shapes are designated by depth of section x weight per unit length (e.g., W24x76), and British Universal Beam shapes are designated by depth of section (mm) x width of a section (mm) by weight (kg/m) (e.g., UB610x229x113). On-site, it is also easy and convenient to measure a section’s physical dimensions of depth, width, and thickness. Sometimes, steelwork designers must quickly size up a section or estimate section capacities without immediate access to comprehensive section properties or code equations. This is where rules-of-thumb become helpful. This article summarizes a few rules-of-thumb to evaluate steel section capacities.

Section Properties When evaluating a section’s capacity, the most relevant section properties are the cross-sectional area, the moment of inertia, and the radius of gyration. A cross-sectional area is required for tension and compression capacity, the moment of inertia for moment capacity, and the radius of gyration for buckling resistance.

Since the density of steel is 490 lb/ft3 (7850 kg/m3), the crosssectional area of section (A) can be derived by dividing the weight per unit length by the density.

Similarly, the moment of inertia (I) and radius of gyration (r)can be correlated to section sizes (depth and width) and weight per unit length.

For ease of remembering, these can be approximated using the expressions in the Table. For example, a W24×76’s cross-sectional area is 76/3.4 = 22.4 in2.

The approximate moment of inertia WD2/20 can be obtained by summing up flange area multiplied by the square of distance from the centroid and assuming a proportion of the total cross-sectional area lies within the flanges. For W-shape sections, by assuming a ¾ proportion, a conversion factor of 20 is obtained which corelates well with the actual moment of inertia. For British Universal Beam (UB) sections, there is more area in the web and the moment of inertia may be closer to WD2/25.

Axial Load Capacity The axial load can be in tension or compression, acting axially along the longitudinal direction of a member. Load-carrying capacity is typically calculated using the stress multiplied by the cross-sectional

54 STRUCTURE magazine

Figure 1. Graph of pc column stresses against slenderness ratio.

area, with appropriate load, resistance, and safety factors applied.

Compression members have an added complication in dealing with slenderness where buckling becomes the governing failure criteria instead of crushing.

One common rule of thumb for compressive strength is: ½ the strength of materials (0.5 × yield strength, Fy) could be used for structural design for column members, i.e., members that are above the ground where they can be seen, and ⅓(0.3Fy) for piles, i.e., members that are below the ground where they are out of sight.

Since the 1800s and 1900s, when designing using allowable stresses, the ultimate (maximum) strength has to be reduced by a safety factor to arrive at a stress used for design. For example, mild steel having an ultimate strength of 65 kips per square inch (ksi) (448 megapascals, MPa) would have an allowable compression stress of 16.25 ksi (112 MPa) using a factor of safety of 4 (i.e., 0.25Fu). This seemingly high factor of safety is because ultimate strength, Fu, is used, which is higher than yield strength. Therefore, upon reaching yield, the material would enter into the plastic range and no longer be elastic.

In modern times, allowable stress design (ASD), governed by the American Institute of Steel Construction (AISC) Specification code, stipulates allowable stress for tension members to be 0.6Fy (yield stress) or 0.5Fu (tensile strength) on the gross and net effective area, respectively.

When steel sections are used as piles (e.g., H-piles), the allowable stress is ϕ(ecc)Fy/(LF), where ϕ, ecc, and LF are the stress-reduction factor, eccentricity, and load factor, respectively. Load factor is

taken as 2, which is in line with the super-structure member safety factor. Due to the uncertainties involved in underground piles, two additional factors were introduced: stress reduction and eccentricity. Φ and ecc are 0.85 and 0.7 respectively, which gives a product of 0.595. Together with a load factor of 2, the allowable stress for piles becomes 0.3Fy in accordance with Federal Highway Association publications.

Another common rule of thumb for the working compression load capacity of steel sections is: 1 kilogram (kg) of steel carries one ton (using a unit conversion of 1 ton = 1000 kg = 10 kilonewton, kN) of load. This rule of thumb is meant for British Universal Beam and Column (UB and UC) sections for S275 MPa (≈Grade 36) steel used as piles and therefore comes with a safety factor of 3. For example, a UB610x229x113 (mm × mm × kg/m) section would have a working load of 1130 kN and a nominal yield capacity of 3390 kN. This estimate is only for a short column without considering slenderness or buckling. The capacity for S355 MPa (≈Grade 50) steel can be estimated by simple proportion (i.e., applying a 355/275 factor).

In US customary units, the conversion factor is 8.8 (=1/490 pcf × 144 in2 × 50 ksi × 0.6) for A992 steel (Fy = 50 ksi). Therefore, the available axial compressive strength of a Grade 50 W14×109 section can be approximated as 109×8.8 = 959 kips.

To account for buckling, which must be considered, the slenderness ratio (c) = (Lc /r), which is the effective length (Lc) divided by the least radius of gyration (r) (the weak axis). For example, the legacy British Standard BS449:1948, The use of structural steel in building, used an allowable column compressive stress of 124 MPa (18 ksi) for a slenderness ratio of c = 20, which decreases to 25.9 MPa (4 ksi) for a slenderness ratio of c = 180. This is almost equivalent to a 15 MPa (2 ksi) decrease in stress capacity for every 20-increment increase in slenderness ratio (c) starting from c = 20 (Figure 1). Note that this curve is valid for steel with an ultimate tensile strength of 56 to 66 ksi.

When we compare two modern codes, namely the American Institute for Steel Construction AISC 360, Specification for Structural Steel Buildings, and the British Standard BS5950-2000, Structural use of steelwork in building, we can see that the column buckling curves for Fy = 50 ksi (345 MPa) steel are very similar. These real-life design curves provide lower column stresses than the theoretical Euler curve, where eccentricities and imperfections are not accounted for.

The deviation from the Euler curve is most pronounced at the curve transiting from buckling to the crushing failure mode. It should be noted that the yield strength of the material does not play a part in buckling. Changing the yield stress only affects the crushing stress portion of the curves where the slenderness ratio is at relatively lower levels. Besides the usual flexural buckling, designers should also be mindful that torsional or flexural-torsional buckling might become critical, especially for certain sections (e.g., asymmetric, singly symmetric, cruciform, or built-up sections, etc.). The legacy British Standard BS449-1948 curve is based on allowable column

Modern design codes contain detailed procedures in dealing with moment and shear for a wide variety of circumstances.

However, in its most basic form, moment capacity can be expressed as a product of yield stress and elastic section modulus. stress, which means that it is already factored. Readers need to be mindful of the various factors required by different codes and adapt accordingly when using simplified rules.

Another valuable formula to evaluate buckling capacity is the Rankine formula: 1 = 1 + 1 Nb Npl Ncr where: Nb = buckling capacity Npl = Fy × A where Fy = yield stress and A = area (plastic limit) Ncr = π2 EA/c2 where E = Young’s modulus and A = area (critical Euler curve) This formula is essentially an equation form of the combined interaction curve for crushing and buckling failure modes. The plastic limit term is constant for a particular section. At low slenderness ratios, 1/ Ncr becomes negligible, and the governing failure mode is crushing.

Conversely, when the slenderness ratio is high, 1/Ncr becomes the dominant term, and the governing failure mode becomes buckling.

Moment and Shear Capacity Moments and shear forces often co-exist together in a structural member in bending. Modern design codes contain detailed procedures in dealing with moment and shear for a wide variety of circumstances.

However, in its most basic form, moment capacity can be expressed as a product of yield stress and elastic section modulus.

Mn = Fy Sx The elastic section modulus is related to a section’s moment of inertia, which can be estimated from the section’s dimension and weight using WD/10 from (WD2/20)/(D/2). Designers should be aware that compact sections, which allow whole section yielding, may stretch the envelope further by adopting the plastic modulus in place of the elastic modulus. The shape factor, a ratio of plastic

Table of approximations.

Units

Area

Moment of inertia

Section Modulus

Radius of gyration

Imperial

A = W/500 (ft2) A = W/3.4 (in2)

I = WD2/20 (in4)

S = I/(D/2) S = WD/10 (in3)

rx = D/2.5 (in) ry = b/4.5 (in)

Metric

W/8000 (m2)

WD2/500 (cm4)

S = I/(D/2) S = WD/250 (cm3)

rx = D/2.5 (mm) ry = b/4.5 (mm)

W = weight per unit length; D = depth of section; b = width of section

APRIL 2022

55

The reader should note that the depth-to-web-thickness ratio for W-sections can vary quite widely from below 20 to above 60 (Figure 2).

Nonetheless, it is still possible to get a rough estimate from the weight-to-depth ratio (W/D), especially for W/D greater than 10.

For example, if the W/D ranges from 10 to 25, the depth to webthickness ratio is likely to range from 10 to 30. The higher the weight, the more likelihood of a thicker web and hence the inverse relation.

In many typical steel beam designs, moment is likely to be critical instead of shear. As an illustration, consider a simply supported beam of length L under uniform loading, w. The maximum moment is wL2/8 at midspan, and the maximum shear is wL/2 at the support. For any given w, the numerical value of the moment is equal to shear when span L is equal to 4. Beyond 4, a moment increases exponentially compared to shear when L increases due to the squaring effect of L. Inspecting a typical section capacity table for moment and shear resistance, it can be generalized that moment capacity will be critical for many typical cases (see sidebar).

Figure 2. Estimate of web-thickness from weight/depth ratios (W-sections). Wide flange column sections are those shapes included in AISC Table 4-1.

modulus against the elastic modulus, may range from 1.1 (I-beams) to 1.5 (rectangle), depending on the cross-sectional geometry. Beams without lateral restraint see a compromised moment capacity due to lateral-torsional buckling. In addition, there are also other aspects to consider, such as local buckling. High shear coincident with moment can reduce a section’s moment capacity. However, such cases are rare in practice, as most beams are designed as simply supported instead of continuous.

Allowable shear capacity is 0.6FyAv, where the shear area Av is the area of the web (depth times web thickness). Most column and beam section web thicknesses can vary considerably, so it is not easy to generalize. Nonetheless, for many shallower W-sections (less than 20 inches), 20% to 50% of cross-sectional area is assumed to fall within the web. The proportion becomes 30% to 50% for the deeper W-sections (deeper than 20 inches).

Summary Very often, a steel section’s most accessible information is its physical dimensions (depth and width) and weight (per unit length). Dimensions and weight can give valuable insights into the section capacity by applying simple rule-of-thumb calculations. However, designers need to be aware of the various load factors, resistance factors, and safety factors required by different codes and different design approaches when using such rules-of-thumb. In addition, it must be remembered that rules-of-thumb may not be one-size-fits-all or a definitive answer to a problem. It is the designer’s responsibility to appreciate the context, the limitations, and the boundaries for which such rules-of-thumb are applicable before relying on the result.■ Reference is included in the PDF version of the online article at STRUCTUREmag.org.

Hee Yang Ng is a Principal Engineer with a building control agency in the Asia-Pacific region.

Moment vs. Shear Criticality Examples S275 UB610x229x113 has a moment and shear capacity of 869 kNm and 1090 kN respectively. Applying the simple rules gives: Moment of inertia

= 113×6102/500

= 84095 cm4

Elastic section modulus

= 84095/(61.0/2)

= 2757 cm3

Moment capacity

= 275×2757000

= 758 kNm

Shear capacity

= 610/50×610×275×0.6 = 1228 kN

An ASTM A992 W12x30 (Fy = 50 ksi) has nominal moment (Fy*Z) and shear capacity (0.6*Fy*Aw) of 180 kip-ft and 96 kips, respectively. Applying the simple rules gives: Moment of inertia

= 30×122/20

= 216 in.4

Elastic section modulus

= 216/(12/2)

= 36 in.3

Moment capacity

= 50 × 36 /12

= 150 kip-ft

Shear capacity = (D/50)D(0.6Fy)

= 12/50×12×50×0.6

= 86 kips

Tip: Moment and shear capacity simplifies to 0.42WD and 0.6D2 respectively.

Simplified formula:

56 STRUCTURE magazine

Imperial

SI

Moment

0.4WD

0.01WD

Shear

0.6D

0.003D2

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APRIL 2022

57

guest COLUMN Cross-Laminated Timber A Guide to Selection and Specification

By Borjen (“B.J.”) Yeh, P.E.

C

ross-laminated timber (CLT) is a prefabricated, solid engineered mass timber panel. Because CLT is prefabricated, most components arrive ready to assemble and go together very quickly. CLT’s large-scale components enable faster construction, not only because of prefabrication but because fewer joints are needed between elements.

CLT also offers incredible strength. The crosswise arrangement of board layers lends integral structural stability to the panel, considerably increases dimensional stability, and ensures uniform load transfer to all sides for excellent structural capacities.

CLT’s panel size varies by manufacturer, but typical widths are 2, 4, 8, and 10 feet, with a thickness of 20 inches or less and a length of up to 60 feet. When selecting and specifying CLT, there are several key characteristics to consider, including allowable design capacities, grades, laminations, and adhesives.

Allowable Design Capacities There are five basic “E” grades, six “V” grades, and three “S” grades for CLT products and layups. E, V, and S indicate a CLT grade with a layup of E-rated or machine stress rated (MSR) laminations, visually graded laminations, or structural composite lumber (SCL) in the longitudinal layers, respectively. Visually graded or SCL laminations are used in the transverse layers for both E and V grades and S grades, respectively. Custom layups of CLT are also permitted, as stipulated in ANSI/APA PRG 320, Standard for Performance-Rated Cross-Laminated Timber.

The allowable stress design (ASD) methodologies reference design capacities for CLT grades with layups of three, five, and seven layers are shown in ANSI/APA PRG 320. The ASD reference design capacities for different CLT products, including custom grades and layups, are typically published in APA Product Reports (www.apawood.org/product-reports) or manufacturers’ literature.

Trademarks and Acceptance Chapter 10 of the 2018 National Design Specification® (NDS) provides design procedures, reference design values, and other information for CLT, while engineering design of connections using dowel-type fasteners in CLT is covered in Chapter 12 of the 2018 NDS. Sections were also added to the 2021 and 2018 International Building Code and International Residential Code regarding CLT used as different structural elements. Clause 8 in the Canadian Standards Association’s CSA O86-19 provides design procedures, resistance values, and other information for CLT used in Canada.

Note: National Design Specification® is a registered trademark of the American Wood Council.

Laminations Any softwood lumber species or species combinations recognized by the American Lumber Standards Committee under PS 20 or Canadian Lumber Standards Accreditation Board under CSA O141 with a minimum published specific gravity of 0.35 are permitted for use in CLT, provided that other requirements specified in ANSI/APA PRG 320 are satisfied. SCL should meet the requirements of ASTM D5456, Standard Specification for Evaluation of Structural Composite Lumber Products, and have an “equivalent specific gravity” of 0.35 or higher.

Lumber grades must be at least 1200f-1.2E MSR or visually graded No. 2 in the longitudinal layers and visually graded No. 3 in the transverse layers.

Moisture content is required to be 12 ± 3% for lumber and 8 ± 3% for SCL at the time of CLT manufacturing unless a lower moisture content is specifically qualified in accordance with the standard.

Adhesives In the U.S., adhesives used for CLT manufacturing are required to meet ANSI 405, Standard for Adhesives for Use in Structural Glued Laminated Timber, with the exception that some gluebond durability tests are not required. This is because CLT manufactured according to ANSI/APA PRG 320 is limited to dry service conditions, and some gluebond durability tests are designed for adhesives in exterior applications. In Canada, CLT adhesives must meet the requirements of CSA O112.10. In both the U.S. and Canada, CLT adhesives must meet ASTM D7247 for heat durability and CSA O177, a smallscale flame test. In addition, CLT adhesives must comply with the requirements for elevated temperature performance in accordance with the full-scale compartment fire test specified in Annex B of ANSI/APA PRG 320.

We are excited to see a lessening of the pandemic’s impact on our authors and advertisers and expect to have more normal page counts in the future.

As a result, we have taken the opportunity to showcase the work that other organizations do in supporting SEs by reinvigorating our Guest Column program. It is a pleasure to have these organizations add to STRUCTURE’s knowledge base. If your organization would like to submit an article proposal, please contact Chair@STRUCTUREmag.org.

58 STRUCTURE magazine

General CLT shall be furnished and installed following the recommendations provided by the CLT manufacturer and the engineering drawing approved by the engineer of record. Permissible details shall be in accordance with the engineering drawing.

Manufacture 1) Materials, Manufacture, and Quality Assurance – Product quality shall conform to ANSI/APA PRG 320, Standard for Performance-Rated Cross-Laminated Timber.

2) Trademarks – CLT products conforming to ANSI/APA PRG 320, Standard for Performance Rated Cross-Laminated Timber, shall be marked with CLT grade, CLT thickness or identification, mill name or identification number, the certification agency logo, and “ANSI PRG 320.” The top face of custom CLT panels with unbalanced layup used for roof or floor shall be marked with a “TOP” stamp.

3) Protection for Shipment – Members shall be protected with a water-resistant covering for shipment.

Find more information on the selection and specification of structural engineered wood products in the APA Engineered Wood Construction Guide, available as a free download at www.apawood.org.■ This article, all or in part, has been previously published in the APA Engineered Wood Construction Guide, Form E30, December 2019. It is reprinted with permission.

An example CLT trademark.

Borjen (“B.J.”) Yeh is Director of Technical Services Division, APA - The Engineered Wood Association in Tacoma, WA. He is an ASTM Fellow, a recipient of ASTM Award of Merit and L.J. Markwardt Award. (borjen.yeh@apawood.org).

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APRIL 2022

59

structural INFLUENCERS Career Reflections

By Carol Post, P.E., S.E., LEED AP

U

pon hearing I was retiring, my Cal Berkeley friend and colleague, John Dal Pino, Chair of the STRUCTURE magazine Editorial Board, asked if I would share some thoughts about my nearly 40 years (WHAT? – how did that happen?) practicing as a Structural Engineer. He suggested many topics, and these were the ones that resonated most with me.

What makes engineering an interesting and rewarding profession?

At our core, we are problem solvers, not just M = wL2/8 but our predisposition to improve the world around us. Be it a project workflow, the technical challenges for projects without precedent, or even the culture in our office, we want tomorrow to be a better day. In all my years, I do not recall a boring day. The years have been a mixture of knowns and unknowns, peppered with physical triumphs in the built environment based on the intangibles of codes and software modeling. Most importantly, it has been interacting with some of the most intelligent people in the working world – both colleagues and clients.

Many professions draw young minds into a particular field of work with the lure of high-paying jobs. All too often, those career paths hit rough patches in your mid-40s when the next generation of 20-yearolds easily replaces acquired skills. In structural engineering, we are constantly learning something new and challenged daily with computerized solutions and the physical considerations of construction. While I have had my share of sleepless nights worrying about a design that may be pushing typical limits, I have never been bored or felt that creativity was being hampered. The knowledge one accrues is not easily supplanted by the next new graduate.

Structural Engineering is the right career for individuals who enjoy people, solving problems, and with a knack for driving efficiency. The most successful SE leaders are intellectually curious about mathematics, physics, and materials and want to explore how their focus integrates with architecture and constructability. They are capable of being the smartest ones in the room, delighted to share their knowledge, as well as being the person most willing to learn something new from someone else.

Adage: If you love what you do, you will never work a day in your life.

60 STRUCTURE magazine

What was your approach to developing and mentoring staff?

As the mother of three sons, I would say that domestic experiences of raising children and hiring their caregivers involved some of the same skillsets for hiring and developing staff. As it applies to Thornton Tomasetti, I came to understand that I could learn as much from the younger staff as they were going to learn from me.

I think most could see that I saw their unique value and that I was trying to encourage them to advance with their natural talents rather than assume one size fits all.

I tried to follow some fundamentals when mentoring staff. Early in an SE career, it is important first to become proficient with your technical skills and dabble in understanding good project management by observing more senior staff at internal and external meetings. This proficiency is often followed by assuming project management responsibilities. Many of us at Thornton Tomasetti helped develop bespoke training modules for our Project Management (PM) staff, which was critical as the firm size grew. I also encourage PMs to use selfguided training by reading books or attending presentations authored by subject matter experts. By honing in on PM skills, including communication and “reading the audience,” staff are better prepared to take on the next growth in their career, which is often developing client relationships and trusted partners and winning work.

The listening skills and pursuit strategies learned at this stage of one’s career may lay the foundation for becoming part of your firm’s senior leadership. At all stages, it is different phases of problem-solving. It begins with physics and ends with sustaining and growing your firm’s value in the AEC community.

Takeaway: Empathy is the key to success with staff and clients.

Are there tips for developing and maintaining a clientele?

Developing trusted partners is akin to playing the long game. More than 20 years ago, I began my outreach to a new client for our firm. It started with a letter to the President (yes, the kind with an envelope and stamp).

When he did not respond, I sent another letter. Long story short, he eventually must have asked the Facilities Group to give me some small project – which was the design of a “rusty lintel.” That start and my professional persistence (a term coined by the new client) have developed into many exciting and rewarding projects. But, equally important has been my deliberate transition of this client to the project managers who did the heavy lifting for the last two projects. I have made it evident that they are the next generation to deliver the quality of service that began two decades ago.

Words of Advice: The best business development strategy is professional persistence and showing your value by discovering your client’s needs.

How does one work successfully within and leading large organizations?

Four years ago, I became the first Chief Quality Assurance Officer at Thornton Tomasetti. Unlike executing project work with clients, I needed to learn how to listen, engage, and empathize with my colleagues as I pursued the cultural changes for our quality and risk management goals. More importantly, there should be an underlying desire for continuous improvement when leading large organizations, coupled with a passion for delivering it. I have been fortunate enough to have spent the last 25 years at a firm that allowed me to explore and develop solutions on various initiatives. My key to success has been to have a road map and a willingness to pivot when others offer betterments to my original plan.

Secret revealed: If you believe in something, do not give up on the idea. Instead, find ways to help others see your vision by challenging yourself to understand their perspective.

What does the profession do right, and what does it do wrong?

Structural engineers are some of the most intelligent and diligent people in our society. Our dedication to the public’s life, safety,

and well-being are not easily executed, especially for the most sophisticated projects. We also fall short on attracting the best and keeping them in our profession. Our pipeline seems to be decreasing, and it is particularly low as it applies to people of color.

This needs to be solved by design professionals and stakeholders working together to attract and sustain minorities who enroll in Bachelors and Masters degree programs.

We are generally modest and humble professionals, which is a good trait. However, it is commonly bemoaned by many in the SE world that our compensation does not match “our value.” While I agree, I also do not feel enough of us are taking the time to explain what we mean by this to the ultimate clients – owners. The only way to fix a problem is to address it head-on and authentically work with an owner to demonstrate that we are there to listen to and solve their problem. We must believe in our value and sell it to our clients by educating them about our worth.

There will always be professionals whose greatest asset is a low fee.

Rather than getting drawn into that business model, those in our profession who have more to offer need collectively to enlighten the world. “The bitterness of poor quality remains long after the joy of a good price.”

Forward Thinking: ADEPT – Advancing Diversity in Engineering Pipeline Talent is an idea I have advocated for over the last year. If interested, I am looking for resources and partners.

Thank you for allowing me to reflect on a career that unfolded in ways I could never have imagined when I graduated from an all-girls Catholic high school in New Jersey. I am grateful, pleased, proud, and energized to begin a new era in my life using the skills and experiences from a rewarding career as a Structural Engineer.■ Carol Post currently serves as Thornton Tomasetti’s Chief Quality Assurance Officer. Prior to this, she led the firm’s Higher Education Market Sector in her role as Principal (cpost@thorntontomasetti.com).

APRIL 2022

61

historic STRUCTURES Silver Bridge Failure 1967 (aka Point Pleasant Bridge)

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng, P.E., P.L.S.

T

he Silver Bridge across the Ohio River connecting Point Pleasant (West Virginia) and Gallipolis (Kanauga, Ohio) was opened to traffic on May 30, 1928, along with a sister bridge connecting St. Mary’s (West Virginia) and Gallipolis (Ohio), also across the Ohio River. It was called the Silver Bridge due to the color of paint used and was built just north of the intersection of the Kanawha River with the Ohio River. They were both vehicular bridges built as toll bridges by the Gallia County Ohio Bridge Company. The federal government’s approval was required, and bills were submitted to House and Senate for consideration. In December 1926, The The original and final design of Florianopolis Bridge.

Corps of Engineers informed the Bridge Company they must, in accordance with the act, submit a “plan of the bridge showing the River, NYC), and Quebec Bridges (St. Lawrence River). He prolength and height of spans; width of draw openings; position of posed to build his bridges by stiffening the chains rather than the piers, abutments, fenders, etc.” deck. In 1903, a Panel of Engineers selected to evaluate chains for the On December 29, 1926, plans were submitted to the Corps, which Manhattan Bridge determined, “the chains have a decided advantage appointed a Board of Engineers to review them. After a public hearing, in the accessibility of all parts for inspection and protection, [emphasis the Board recommended approval of the plans with minor revisions. added] as well as in economy and rapidity of erection and they are The plans drawn by J. E. Greiner were for a conventional suspension to be preferred to wire cable whenever the cost of the chains is not bridge with wire cables, two lanes of traffic, and a sidewalk located materially greater.” Lindenthal was unsuccessful in having any of inside the side trusses. On April 28, 1927, the American Bridge his plans for a chain bridge accepted. In 1922, H. D. Robinson and Company, which made the low bid, submitted plans for approval to David B. Steinman were selected as design engineers for a bridge at the Corps. They were rejected as they did not conform to the approved Florianopolis (Brazil) to link the mainland with the island of Santa plan. They submitted revised plans on May 2, 1927, which were Catarina. Originally, they designed a standard wire cable suspension approved. A new District Engineer took over and requested a set of bridge. However, with pressure from the American Bridge Company plans and specifications for the bridge. In a letter accompanying the (which had developed a new high strength eyebar) and the bridge plans, Greiner wrote to the District Engineer, owners, they changed their design to a chain.

“In accordance with your request, we are sending you a set of plans They also adopted Lindenthal’s pinned tower rocker bearings at the and specifications for the Point Pleasant Highway Bridge. The cable foot of the two-dimensional towers. There were four 2- x 12-inch bars design calls for a straight wire cable. In asking for bids on the super- per chain, and the main span was 1,113 feet 9 inches long. structure, two alternatives to this, namely, a long lay wire rope cable The same thing happened at Point Pleasant, where the American or heat-treated I-bars, may be bid upon. We expect to have these bids Bridge Company convinced Greiner and the Bridge Company that in within a short time, after which the a similar design was safe and less type of cable will be decided upon, expensive than a conventional wire and we will forward you additional cable suspension bridge. American plans covering this part of the work.”

Bridge had used their eye bars on The Bridge Company selected the the Three Sisters Bridges across American Bridge Company to build the Allegheny River in Pittsburgh the bridge using a chain made of eyebetween 1924 and 1928, using an bars. Eye-bar suspension bridges had alternating nest of nine and eight been built in Britain, starting with eyebars. They claimed their new Thomas Telford, Samuel Brown, “heat-treated carbon steel…would and Isambard Brunel. In the United allow the individual members of the States, James Finley was the first to bridge to handle more stress.” Along use chains for his suspension bridges. with the two eye-bars sharing the Gustav Lindenthal became a propoload, the steel could easily handle the nent for iron chain bridges for his 4-million-pound load. The newly proposed North River (Hudson) treated chain steel eye-bars had an Bridge, Manhattan Bridge (East Point Pleasant Bridge pin joint above stiffening trusses. ultimate strength of 105,000 pounds

62 STRUCTURE magazine

Silver Bridge profile.

per square inch (psi) with an elastic limit of 75,000 psi along with a maximum working stress of 50,000 psi.

A huge pin passed through the eyes and linked each set of bars to the next. Each chain link consisted of a pair of 2- x 12-inch bars, and an 11-inch-diameter pin connected the links. The length of each link varied, up to 55 feet, depending upon its location on the bridge. With the higher strength of the steel, only two bars per panel, as noted, were required. Earlier chain bridges, both built and proposed, had many bars. For instance, Lindenthal’s chains for his proposed Manhattan Bridge had up to 20 links.

Construction of the Silver Bridge superstructure began in late 1927 and opened to a grand celebration on May 30, 1928. On June 21, 1928, the District Engineer made a final inspection and reported that the conditions of the permit “have been fully complied with and the work completed in substantial accordance with the approved plans.”

The bridge had side spans of 380 feet and a main span of 700 feet with short approach spans. The total length of the bridge was 1,750 feet.

On December 31, 1951, the bridge became a toll-free structure when West Virginia purchased it. Only cosmetic changes were made at the time of the purchase. Other major inspections were made in 1955, 1961, and 1965, all of which determined the bridge was safe.

However, these inspections were visual-only, including the deck, trussing, and chain links. Since only the exterior surfaces of the links could be observed, there was no way of determining if corrosion, cracking, etc., of the interior, not visible, surfaces of the eyes were present.

that the eye of one link had fractured, setting up a series of events that led to the failure. They issued an interim report on October 4, 1968, stating, “it was determined that the fracture in suspension chain eyebar 330 (north bar, north chain, Ohio side span) was essential to the catastrophic stage of the collapse, but the cause of the failure had not been determined…When the north eyebar chain was separated at joint C13N, total collapse of the bridge was a certainty due to its design with towers resting on rocker seats.”

In the final report (available online) issued on December 16, 1970, NTSB broke it up into four parts, A. With respect to the Sequence of Events in the Collapse of the Point Pleasant Bridge.

B. With Respect to the Elements Which Contributed to the Failure.

C. With Respect to the Implications for the Safety of Other Bridges.

D. With Respect to the Status of Bridge Inspection and Maintenance.

In Part A, they had seven conclusions, five of which are given below.

1) The total collapse of the structure required the failure of some element in the supporting chains or towers. The directions in which the towers fell indicate that this failure was at the Ohio tower or west of this point in the Ohio side span, and in the north chain or its supporting elements.

2) Examination of the Ohio tower wreckage showed no failure in the north leg, and the laboratory examination of the fractures in the Ohio north chain bent post and gusset plate Bridge Failure and Investigation U7N showed these fractures to have occurred from excessive or abnormal loads beyond those which were possible from On a cold winter’s evening, December 15, 1967, the Silver Bridge the loading on the structure just prior to collapse. The only collapsed into the Ohio River under a full load of Christmas shopremaining failure in the chain or its supporting elements in pers, large trucks, etc. Observers stated, “the sound of the collapse the Ohio side span, which was like that of a shotgun” and “the could have led to collapse, is a bridge just keeled over, starting slowly failure of some element in the on the Ohio side and then folding like chain itself. a deck of cards to the West Virginia 3) The joint at C13N, the first side.” In an instant, sixty-four people joint in the north chain west in 32 vehicles fell into the river, and 46 Although the bridge had served of the Ohio tower, began to of them died from drowning or being for 39 years without incident, separate because of the brittle crushed by the falling bridge. After fractures in eyebar No. 330 rescuing those they could and retrievmany still blamed the engineer (the northerly bar of the pair ing the bodies of those that died, the in the north chain connectnatural question arose as to how such and contractor in the belief there ing pins at joints C11N and an instantaneous collapse could occur. was no equivalent of the Statute of C13N). Subsequent to this The National Transportation Safety fracture, eyebar No.33 (the Board (NTSB) began its analysis Limitations for the bridge engineer. southerly bar of this pair) slid shortly after the failure. One of the off the south end of the pin, first steps was to remove the wreckage causing complete separation of from the river and lay it out in a large the north chain at this point. nearby field for analysis. They found

continued on next page APRIL 2022

63

Silver Bridge.

The most significant positive result of the failure was the creation of the National Bridge Inspection Program, which has identified many bridges that were failing or likely to fail in the future if no corrective action was taken.

4) With respect to the brittle fracture in eyebar No. 330, the laboratory work has shown that: a) The small crack which existed prior to the collapse was large enough to account for the brittle fracture in the special steel of which the eyebars were made at the stress level computed to exist at this location, without any additional dynamic effects. b) This small crack probably initiated at a small corrosion pit. c) The crack grew to critical size by the joint action of stresscorrosion cracking and corrosion fatigue. The available evidence is not sufficient to permit a definite conclusion as to which mechanism was predominant.

5) The small size of the critical crack in eyebar No. 330, and its location on the inside surface of the hole, precluded it being found while the structure was intact by the inspection techniques used, or by any other inspection technique available at this time for use in the field on heavy structures, without disassembly of the joint.

In Part B, two conclusions stand out, 6) The point of high stress was not accessible for inspection.

7) The use of only two eyebars per link in the eyebar chain. This made the total failure of the chain inevitable once the fracture occurred in eyebar No. 330. Had there been three or more eyebars per link, there would have been the possibility that the failure of one bar would not have led to disaster.

NTSB concluded with Parts C and D promoting greater bridge inspection. This resulted in the passing of a part of the 1968 Federal-aid Highway Act, from which National Bridge Inspection Standards (NBIS) were adopted by the Federal Highway Administration (FHWA) on April 27, 1971. It requires that all public bridges with spans over 20 feet be examined every two years and, if they are considered to be at a higher risk, more frequently.

The Safety Board did not specifically assign blame to J. E. Greiner or the American Bridge Company. Lawsuits against the Federal Government were unsuccessful. Lawsuits were also filed against the West Virginia Department of Highways. Still, they were disallowed by the court that ruled, “the collapse could not have been anticipated or 64 STRUCTURE magazine

foreseen by the respondent in the exercise of reasonable care.” Lawsuits against American Bridge and Greiner were filed, and in August 1973, they agreed to pay $950,000 in a settlement. Although the bridge had served for 39 years without incident, many still blamed the engineer and contractor in the belief there was no equivalent of the Statute of Limitations for the bridge engineer. The engineer must, it seems, within the bounds of economics, build for the ages.

Lessons Learned Conclusion #5 has since been partially addressed with the availability of many non-destructive testing techniques that might have identified the crack if used on the Silver Bridge. The discovery of the crack probably would not have saved the bridge but could have prevented the loss of life. The most significant positive result of the failure was the creation of the National Bridge Inspection Program, which has identified many bridges that were failing or likely to fail in the future if no corrective action was taken.

Most engineers learn in college that a chain is only as strong as its weakest link. One statement of the Board that pertains to this truism is, “The designer, therefore, must be careful in deciding which influences should be assumed to occur simultaneously and must make appropriate reductions in the factor of safety for highly improbable conditions.” With only two bars in each link, it could be asked, did the designer take into consideration “highly improbable conditions” such as the failure of a single bar in a single link? This is sometimes called the What If question.

The collapse also led to the closure and demolition of the St. Marys Bridge as a safety measure. Engineers could not assure the public that the same thing would not happen to it as had happened to the Point Pleasant Bridge. A replacement for the Point Pleasant Bridge, a cantilever span, was completed one mile downstream in 1969 and is called the Silver Memorial Bridge.■ Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having restored many 19 t h Century cast and wrought iron bridges. He is now an Independent Consulting Engineer (fgriggsjr@twc.com).

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65

SPOTLIGHT Efficient Structural Elements Define Stunning Architectural Form

T

he David Rubenstein Forum at the University of Chicago is a new center for intellectual exchange, scholarly collaboration, and special events. The 97,000-square-foot facility consists of a 2-story podium and a 10-story tower of stacked “neighborhoods” with a zinc-and-glass exterior. A 285-seat auditorium sits above the podium, and a large multipurpose space on the 2nd Floor, called the University Room, can accommodate groups of up to 600 people. The top of the tower features a flat-floor multipurpose space and offers stunning views of the campus, the Midway Plaisance, the city skyline, and Lake Michigan.

At 166 feet tall, the Forum rises over 100 feet above the surrounding tree line, striking a dynamic form that symbolically ties together the history of Chicago’s Midway Plaisance, the University’s heritage and ambitions, and the architectural legacy of the area. Conceived as a scholarly retreat, the design promotes openness and diverse interactions, from the mixed program at the base to the serene City View Room on the 10th Floor.

The stacked neighborhoods of the tower are staggered to varying degrees, posing a complex structural challenge, with cantilevers up to 40 feet. The structural system is composed of a series of simple individual structural elements stacked in a harmonious form to efficiently create the desired open spaces. Incorporating post-tensioned concrete into the structural design proved crucial to achieving the long spans, cantilevers, and column-free spaces that the architectural team envisioned.

Several core design principles informed the structural systems of the David Rubenstein Forum, the key to helping meet the University of Chicago’s and the Architect’s goals for the project. These included: • Openness to capture views of the University, of the Chicago skyline, and Lake Michigan and allow for free stacking of the “neighborhoods” and flexibility of event planning, without the need for expensive structural transfers.

• Simple Use of Familiar Materials enables the efficient construction of conventional structural elements and technologies commonly used in the Chicago concrete market.

• Unique Form to “stitch” the north and south sides of Chicago together while 66 STRUCTURE magazine

simultaneously taking advantage of the prominent site location on the Midway Plaisance to create a highly visible landmark.

• Balance to achieve the cantilevers to the north and south while minimizing the structural work required in the building’s interior core.

The innovation of the structure lies in its simplicity. The efficient form also made it cost-effective, utilizing common materials and systems that could be constructed in a familiar manner.

The main support for the Tower is provided by a pair of 12-inch-thick, conventionally reinforced concrete side walls that are skewed from neighborhood to neighborhood. These side walls, and the slabs that are connected LERA Consulting Structural Engineers was an Outstanding to them, form a stiff box-like structure Award Winner for the David Rubenstein Forum, University of that cantilevers out to the north and Chicago project in the 2021 Annual NCSEA Excellence in south by as much as 40 feet. The backStructural Engineering Awards Program in the Category – New and-forth cantilevers are balanced to Buildings $80M to $200M. help minimize the amount of bending induced into the building’s central core, enabling the use of thin 8-inch-thick compressive strength of 8,000 psi or higher. In shear walls around the elevators and egress addition to using high strength concretes, the stairs, where space is limited, and coordina- modulus of elasticity and a low 28-day drying tion of systems is critical. Four rectangular shrinkage limit were specified to minimize concrete columns also support the slabs, the likelihood of developing restraint-towhich are located just outside of the central shortening cracking due to the robust side core to preserve the open concept. walls and additional rebar placed in the slabs An unbonded post-tensioning system is used in along the walls. The General Contractor and the floor slabs, which form the top and bottom Concrete Contractor were involved early in a of the stacked boxes, to enable the long, column- design-assist phase, together with input from free 65-foot spans and reduce the tensile stresses the post-tensioning supplier, to recommend in the top of the cantilevered boxes. These long, ways to simplify the construction of the concolumn-free spans are an essential component crete structure. of the functionality of the building, maintaining As with the Tower, the principle of openness openness throughout. The typical slab thickness remained key at the Podium. However, the is 10 inches and is only thickened where needed, clear open span over the University Room is such as where the side walls are skewed and do 105 feet, compared with the 65-foot clear not align from the floor above to the floor below spans in the Tower. The design uses steel and at interior-to-exterior transitions. The result roof trusses, which in turn rest upon, and is a long-span slab that is efficiently constructed are transferred by, a system of cantilevered with simple, flat formwork that allows for a post-tensioned slabs and beams at Level 2 more efficient distribution of ductwork due to to accomplish this span. To allow for a high no interior beams. ceiling at the interface with the façade, the Additionally, the structural design took trusses are set back from the south edge of advantage of the high-strength concretes the University Room, which also available in the Chicago construction market. requires the steel roof beams to canAll concrete used in the superstructure has a tilever from the trusses.■ APRIL 2022

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SPOTLIGHT Investigation and Creative Structural Solutions

T

he historic Alberta Bair Theater renovation was an exercise in forensic investigation, creative detailing and problemsolving, and construction coordination. Lack of original drawings, combined with a previous remodel, required extensive field research to document existing construction and load paths. Close coordination during construction was necessary to validate designs and deliver final details. Due to rigging capacities, the theater was also at risk of not booking new acts. Creative structural solutions were implemented to nearly triple the technical rigging capacity of the theater, all while reducing the net loading on existing members.

The Alberta Bair Theater was constructed in Billings, Montana, in 1931 and consisted of board-formed concrete with clay tile infill walls, riveted steel trusses, and a combination of nail-laminated timber and cast concrete roofs. Unfortunately, no existing drawings were available that represented the existing construction, design loads, material strengths, or details.

A remodel in 1985 made significant changes to the theater. The stage area was expanded into the seating, and a five-degree skew was added to the stage. The counterweight rigging system was moved outside of the footprint of the building and cantilevered off the existing walls. A catwalk and a light rigging system were suspended from the original trusses. Walls and beams were cut and moved to allow new framing and openings.

Structural drawings from this remodel were available, providing clues about the original construction.

A forensic investigation was performed early in the design phase to gather as much information as possible about the structure. Load paths, sizes, and details were documented as best as practicable. Destructive testing and removal of finishes were not permitted as the theater was still hosting performances while the renovations were being designed.

New details and framing plans were designed with flexibility as most all verifications had to be performed as demolition progressed. Work was sequenced to uncover original conditions, verify dimensions and substrates, verify engineering designs, and provide final coordinated details to fabricators and other subcontractors.

A significant component of the remodel work was to increase the technical rigging capacity of the theater from a limited 40,000 STRUCTURE magazine

pounds to 125,000 pounds to accommodate modern touring shows. In addition to upgrading the rigging capacity over the stage, it was requested to provide new catwalk circulation and an additional 15,000 pounds of rigging capacity out over the seating directly in front of the stage.

The 1980s remodel complicated solutions for the rigging upgrades. By adding the skew to the front of the stage, none of the original Cushing Terrell was an Outstanding Award Winner for the Alberta and remodel framing was Bair Theater project in the 2021 Annual NCSEA Excellence in parallel or orthogonal. Steel Structural Engineering Awards Program in the Category – Forensic/ clashes were prevalent. The Renovation/Retrofit/Rehabilitation Structures under $20M. decision was to re-frame the fly loft entirely to better connect original and remodel construction. and strengthened. In addition, discrete Requested rigging upgrades and other modi- members and connections were stiffened to fications presented a monumental challenge. accommodate new reactions. Finally, some The 5 and 10% allowable stress increases per- cantilevered elements and tension-only mitted by the International Existing Building members were modified to provide direct Code (IEBC) would not provide capacity. bearing and redundant load paths as addiIt was not deemed appropriate to use those tional safety measures. stress increases on the remodel of a historic Rigging capacity was increased in front of the structure. Creative solutions were needed. stage by inserting several trusses. The framThe original stage roof construction was cast ing was erected on the theater floor and then in place concrete. A deep beam originally lifted into the final position by a blind crane spanned over the stage opening and supported pick outside the theater. The trusses spanned both the fly loft floor and roof. The previous independent of existing framing and were supremodel removed the deep beam to enlarge the ported on each side by new framing. One side stage. Additionally, a metal roof was built over was supported by a new, full-height concrete the original concrete roof to accommodate the shear wall added to the structure following a new framing. This left the original concrete lateral analysis of the original structure. The roof suspended over the stage, providing no opposite end was supported by a new column gravity support or diaphragm action. hidden behind a new acoustical reflector.

After analyzing the modified load paths, Finally, the building was wrapped in a glass the decision was made to remove the original façade. This required a custom structural soluconcrete roof. This significantly lightened the tion to support the coiled steel fabric fins that weight of the existing construction. A robot was mimic a theater scrim being lifted at the street brought in to work in the tight spaces and safely corner to reveal the start of a show. Inside the removed the concrete from overhead. More than glass façade, the raw structure was left in place 100,000 pounds of concrete were removed. The to celebrate the historic terra cotta detailing. removal allowed the fly loft to be re-framed and This project was an exercise in investigation the rigging capacity to be increased to 95,000 and creative problem-solving. All finished, the pounds, all with a net reduction in loads and renovations to Alberta Bair Theater extend stresses on existing members. the historic life of a community fixture. In The modification and rigging increase addition, the structural rigging upgrades, required verification of loads paths, and while invisible to guests, allow the both original construction and previously theater to provide acts previously remodeled framing needed to be re-evaluated not possible.■ APRIL 2022

67

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NCSEA

NCSEA News

National Council of Structural Engineers Associations Congratulations to the 2021 Excellence in Structural Engineering Outstanding Projects and Award Winners

NCSEA announced the winners of the 2021 Excellence in Structural Engineering Awards at an awards celebration on Wednesday, February 16th, at the NCSEA Structural Engineering Summit in New York City. These awards were presented in eight categories, each with one or two Outstanding Project Winners. Congratulations to all!

Learn more about these projects in the March issue of STRUCTURE, and watch the awards presentation video at https://bit.ly/3KMn1kV. Would you like to see your firm’s prized project in this list of winners for the 2022 awards program? Visit the NCSEA website to learn more.

Category 1: New Buildings < $30 Million Outstanding Project Red Rocks Amphitheater Stage Roof Replacement Morrison, CO | Martin/Martin, Inc.

Category 5: New Bridges or Transportation Structures Outstanding Project State Route 99 Alaskan Way Viaduct Replacement Program Seattle, WA | HNTB Corporation

Outstanding Project DC Southwest Library – Washington, D.C. | StructureCraft

Award Winner Gerald Desmond Bridge Replacement – Long Beach, CA | ARUP

Category 2: New Buildings $30 Million to $80 Million Outstanding Project Taiyuan Botanical Garden Domes Taiyuan, Shanxi Province, China | StructureCraft

Award Winner Sea-Tac International Arrivals Facility Pedestrian Walkway SeaTac, WA | KPFF Consulting Engineers

Award Winner The George W. Peavy Forest Science Center Corvallis, OR | Equilibrium Consulting Inc

Category 6: Forensic/Renovation/Retrofit/Rehabilitation Structures up to $20 Million Outstanding Project Alberta Bair Theater – Billings, MT | Cushing Terrell

Award Winner 2461 Broadway – New York, NY | WSP USA

Award Winner Apple Park – Greenville, NC | Collins Structural Consulting, PLLC

Category 3: New Buildings $80 Million to $200 Million Outstanding Project Stanford Center for Academic Medicine – Stanford, CA | HOK

Award Winner Brent Spence Bridge Emergency Repair Covington, KY to Cincinnati, OH | Michael Baker International

Outstanding Project David Rubenstein Forum, University of Chicago Chicago, IL | LERA Consulting Structural Engineers

Award Winner Sperry Chalet Reconstruction Glacier National Park, MT | JVA, Inc.

Category 4: New Buildings Over $200 Million Outstanding Project Rainier Square – Seattle, WA | Magnusson Klemencic Associates

Category 7: Forensic/Renovation/Retrofit/Rehabilitation Structures over $20 Million Outstanding Project Savannah Plant Riverside Project Savannah, GA | Browder + LeGuizamon and Associates

Award Winner 425 Park Avenue – New York, NY | WSP USA Award Winner Conrad Washington, DC Washington, D.C. | Thornton Tomasetti, Inc.

Outstanding Project Syracuse University Stadium–New Roof Project Syracuse, NY | Geiger Engineers Category 8: Other Structures Outstanding Project Little Island – New York, NY | ARUP Award Winner East End Gateway – Entrance Canopy, MTA C&D – New York, NY Skidmore, Owings & Merrill, in association with AECOM Award Winner Moynihan Train Hall Skylights New York, NY | Schlaich Bergermann Partner

follow @NCSEA on social media for the latest news & events!

68 STRUCTURE magazine

News from the National Council of Structural Engineers Associations

Wanted – Expert Speakers, Exhibitors, and Attendees Join Us in Chicago for the Structural Engineering Summit in November

Mark your calendars for November 1-4 in Chicago, and meet us in the windy city to network and learn with those who know wind loads best! The NCSEA Structural Engineering Summit offers unrivaled educational opportunities with leading experts, unique networking opportunities at the trade show, inspirational keynote speakers, and a celebration of structural engineering ingenuity and service like no other.

Would you like to be one of those expert speakers? Learn more about the Summit Call for Abstracts and submit your presentation abstract by the April 15th deadline at https://bit.ly/2022SummitAbstracts.

Our trade show is filling fast – 30 exhibitors are already signed on for networking and fun with structural engineers. If you would like to join our growing exhibitor list, please visit https://bit.ly/Summit2022Prospectus to learn more.

We had great fun at the Summit in New York in February – you can see for yourself by checking out the photos at https://bit.ly/SummitNYCPhotos.

Let’s continue the celebration and top-notch educational experience in Chicago in November – see you there!

Diversity in Structural Engineering Scholarship Program The NCSEA Diversity in Structural Engineering Scholarship was established by the NCSEA Foundation to award students who have been traditionally underrepresented in structural engineering (including but not limited to Black/African Americans, Native/Indigenous Americans, Hispanics/Latinos, and other people of color).

Please spread the word! Applications are being accepted now through April 30th. Multiple scholarships will be awarded to junior college, undergraduate, and/or graduate students actively pursuing structural engineering degrees and careers.

Please visit www.ncsea.com/awards/scholarship to submit a scholarship application.

NCSEA Webinars

Visit www.ncsea.com/education for the latest news on upcoming webinars and other virtual events.

April 12, 2022

Ground Improvement and the Structural Engineer

April 21, 2022

Shotcrete in Concrete Codes and Standards

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APRIL 2022

69

SEI Update Learning / Networking

Join us to connect with friends and colleagues, interact with and learn the latest from the experts, browse cutting-edge publications and resources, and bring back new ideas to improve your practice. Check out the excellent program of sessions, speakers, and exhibitors, and register at www.structurescongress.org. We look forward to welcoming you!

The 2022 SEI Standards Series Review changes from ASCE 7-16, the Digital Products/Hazard Tool, and join the discussion with the expert standard developers.

1.5 PDHs per session.

• May 12, 2022: ASCE 7-22 Seismic • June 9, 2022: ASCE 7-22 Wind & Tornado • July 14, 2022: ASCE 7-22 Snow/Rain • September 8, 2022: How & Why to Use ASCE 7-22 in Your Practice Learn more and register at https://collaborate.asce.org/integratedstructures/sei-standards.

NEW in the ASCE Bookstore

2022 Fazlur Rahman Khan Distinguished Lecture Series Friday, April 29, 2022 – 4:30 pm EST Increasing the Resilience of Highway Bridges Under Multiple Hazards Including Earthquake, Tsunami, Corrosion and Climate Change by Mitsuyoshi Akiyama, Professor and Chair of the Department of Civil and Environmental Engineering, Waseda University, Tokyo, Japan.

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Lectures are in-person at Lehigh University and live-streamed. Register at www.lehigh.edu/~infrk.

Errata 70 STRUCTURE magazine

SEI Standards Supplements and Errata including ASCE 7. See www.asce.org/SEI. To submit errata, contact sei@asce.org.

News of the Structural Engineering Institute of ASCE Students and Young Professionals

Last Call for Students

Apply by April 25 for New Scholarship to Electrical Transmission and Substation Structures Conference October 2-6 in Orlando Expand your career opportunities, connect with leaders, and get involved in this important industry that innovates for critical global infrastructure. The scholarship was made possible by the SEI Futures Fund. www.etsconference.org

Advancing the Profession Sign up for updates to receive the latest safety information and news from Collaborative Reporting for Safer Structures – US, UK, and Australasia – and see the latest CROSS UK Safety Alert on Safety issues associated with balconies. www.cross-safety.org/us

Thank You, SEI Futures Fund Donors!

Thank you to all that generously support the SEI Futures Fund. Your gifts fund SEI strategic initiatives – exceeding $250,000 in FY22. Learn more and give at www.asce.org/SEIFuturesFund.

FOUNDERS John L. and Karen E. Carrato Jon D. Magnusson

JAMES LAURIE VISIONARY CIRCLE John L. and Karen E. Carrato Sarmad (Sam) and Ina Rihani Odeh Engineers, Inc.

LEGACY SOCIETY Phillip L. Gould Silky S. K. Wong

2021 Annual Gift Support PHILANTHROPIST SOCIETY $10,000+ Odeh Engineers, Inc. CHAMPION SOCIETY $5,000-9,999 Severud Associates Consulting Engineers GUARDIAN SOCIETY $2,500-4,999 James R. Harris LEADER SOCIETY $1,000-2,499 William F. Baker, Jr. Cary H. Beatisula Glenn and Judy Bell Joseph G. Burns John L. and Karen E. Carrato Edward M. and Mary F. DePaola and Family Thomas A. DiBlasi Anne M. Ellis Theodore V. Galambos Ramon E. Gilsanz Daniel L. Lavrich

Sarmad (Sam) and Ina Rihani Jon A. Schmidt Donald R. Scott J. Greg Soules Anonymous SUPPORTER CLUB $500-999 Ahmad K. Abdelrazaq Robert E. Bachman Randall P. Bernhardt Laura E. Champion Joseph G. DiPompeo Bruce R. Ellingwood Roberto Leon Brian P. Quinn Victor E. Van Santen Silky S. K. Wong Beverly M. and Loring A. Wyllie, Jr. FRIEND CLUB $250-499 John Cleary Suzanne Fisher

Jerome F. Hajjar Matthew B. Kawczenski Norma Jean Mattei Ivan Ramirez Stephen S. Szoke Dennis and Nancy Tewksbury Paul Z. Zia DONOR CLUB $100-249 Walter L. Allen, Jr. Johnny M. Aquino Suzanne Aultman Anthony C. Cerino Peter H. Chase Donald Dusenberry Thuy Fontelera Phillip. L. Gould Jennifer Goupil John O. Grieshaber Jeremy Isenberg Linda M. Kaplan Takahiko Kimura Peter L. Lee

SUSTAINERS Anne M. Ellis Jennifer Goupil Julian J. Lineham J. Greg Soules Julian J. B. Lineham Kishor C. Mehta Harry W. Shenton, III Stephanie L. Slocum William H. Thorpe James P. Wacker Mehdi S. Zarghamee United Way of Midland County Anonymous (3) $1-99 Soussan Bathaee John W. Fisher Subhash V. Kulkarni Jon D. Magnusson Peter D. Marxhausen Sher A. Mirza Heather C. Neri Andre Newinski Sergio F. Plaza Michael W. Salmon Hongyu Zhou Anonymous

Follow SEI on Social Media: APRIL 2022

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CASE in Point Join a Coalition

The Coalition of American Structural Engineers (CASE) is a dedicated community in ACEC committed to advancing the business practices of structural engineering through education, networking, and the development of critical business resources.

CASE is open to all ACEC members. Coalition membership is firm based and all firm employees are invited to take advantage of the membership.

For more information contact Michelle Kroeger at coalitions@acec.org

Before it’s too late – contribute to this year’s CASE scholarship!

Your monetary support is vital in helping CASE and ACEC increase scholarships to those students who are the future of our industry. All donations toward the program may be eligible for tax deduction and you don't have to be an ACEC member to donate!

Donate today: www.acecresearchinstitute.org/scholarships.

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News of the Coalition of American Structural Engineers Events in Structural Engineering ACEC Coalitions Winter Meeting

At the Winter Meeting held in San Diego, CASE members, along with fellow ACEC Coalitions, participated in two educational sessions: Remote Monitoring Using Today’s Technologies – how new remote sensing technologies are changing the way civil engineers do business, and Navigating the Challenges of Working with Out of State Employees – how firms can manage legal, financial, and human resource activities for out of state employees.

CASE and CAMEE held a joint roundtable with about 35 members in attendance. Coalition roundtables are a great way to bring together principals from a diverse collection of engineering firms to discuss real-time business practice issues affecting our profession. The audience contributes discussion topics and then votes on 2 or 3 to open up to the floor. The topic that generated the most interest by far was remote employees. Remote work is widespread among member firms, and much back and forth was held on implementation strategies and the impact remote work has had so far on workflow, communication, culture, mentoring, hiring, retention, compensation, tax issues, and business/ professional licensing – to name a few!

In the remaining time, the group also touched on current trends in professional liability coverage limits, claims on design-build projects, and other topics. If you get the chance, join CASE for our next roundtable at the ACEC Annual Conference and Legislative Summit in Washington, D.C., on May 23rd.

ACEC Annual Convention & Legislative Summit May 22-25, 2022, Grand Hyatt Washington, D.C.

Join fellow structural engineers to discuss the latest trends, challenges, and opportunities facing your markets.

The Roundtable is open to everyone. National meetings provide attendees an opportunity to obtain information about issues that affect the industry through informative education, networking, and exhibits. Registration is open now. Early Bird Registration ends Thursday, April 21, 2022. Go to www.acec.org/conferences.

Have you seen what’s new in CASE Publications?

Did you know? CASE has tools and practice guidelines to help firms deal with a wide variety of business scenarios that structural engineering firms face daily. So whether your firm needs to establish a new Quality Assurance Program, update its risk management program, keep track of the skills their young engineers are learning at each level of experience, or need a sample contract document – CASE has the tools you need!

Check out the most popular CASE tools and guidelines: • CASE 962-D – A Guideline Addressing Coordination and Completeness of Structural Construction Documents (2020). Since the mid1990s, owners, contractors, and design professionals have expressed concern about the level of quality of structural construction documents. They have observed that the quality of these documents has deteriorated, resulting, at times, in poorly coordinated and incomplete design drawings. Inadequate and/or incomplete design drawings often result in inaccurate competitive bids; delays in schedule; a multiplicity of requests for information (RFIs), change orders, and revision costs; increased project costs; and a general dissatisfaction with the project. The Council of American Structural Engineers (CASE) has prepared this Guideline to address these concerns. This book discusses the purpose of the guideline, the background behind the issues, the important aspects of design relationships, communication, coordination and completeness, guidance for dimensioning of structural drawings, effects of various project delivery systems, document revisions, and closes with recommendations for development and application of quality management procedures. In addition, the Guideline includes Drawing Review Checklist.

• CASE Agreement #6 – An Agreement between Client and Structural Engineer for a Structural Condition Assessment. The purpose of this document is to provide a sample Agreement for structural engineers to use when providing a structural condition assessment directly to a client. For example, this may be required for upgrading the structure for an increase in imposed loads; for damage from fire, wind, or earthquake; for seismic retrofitting; for historic preservation or change in occupancy; or for adding new structures upon or adjacent to an existing structure.

• CASE #2 – An Agreement Between Client and Structural Engineer of Record for Professional Services. This agreement form may be used when the client, e.g., owner, contractor developer, etc., wishes to retain the Structural Engineer of Record directly. The contract contains an easy-to-understand matrix of services that simplify the “what is included and what is not” questions in negotiations with a prospective client. This agreement may also be used with a client who is an architect when the architect-owner agreement is not an AIA agreement.

You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

Follow ACEC Coalitions on Twitter – @ACECCoalitions.

APRIL 2022

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structural LICENSURE SECB Passing the Torch By the SECB Board of Directors

I

n 2003, the Structural Engineering Certification Board (SECB) was established with a simple mission with three straightforward goals: Promote structural engineering licensure (SE) in all jurisdictions; determine the unique and additional qualities (beyond a professional engineering license) necessary to practice structural engineering; and provide the engineering profession, the public, and other stakeholders with a way to identify engineers with these unique and additional qualifications. Originally formed through a group of past presidents of the National Council of Structural Engineers Associations (NCSEA) as an interim step towards SE licensure, SECB eventually became a financially stable organization with certificate holders in all 50 states and a roster of more than 1,200 certified structural engineers. While states would not initially recognize this new SECB certification, the vision was to create momentum within the structural engineering community for higher credentialling that would serve as a model or bridge for SE licensure adoption in more states.

Now, almost 20 years later, the time has come for the profession to reaffirm its focus on our end goal of nationwide SE licensure. Starting a new chapter is always challenging, as it begins by ending one.

But the first chapter, SECB’s work, has helped us figure out where we want to go – and now is the time to go there. In September of 2021, the SECB governing board, in conjunction with the governing boards of NCSEA, the Structural Engineering Institute (SEI), the Coalition of American Structural Engineers (CASE), and the Structural Engineering Licensure Coalition (SELC) agreed that the need for the interim step no longer exists and that it is now time for our profession to focus solely on a direct path toward SE licensing.

SECB will therefore be closing its doors effective March 31, 2022, coinciding with the expiration of all current dues.

Although its progress has been slower than desired, SECB has had some success in reaching its goals. Its initial efforts – starting with its very creation – brought the SE licensure into broader awareness within the profession. Position papers, webinars, and articles brought the issue into focus and spurred a lively debate. Increased inclusion of the thoughts and opinions of practicing structural engineers allowed SECB to define the benefits of an SE designation better and refine its certification criteria.

Significantly, SECB was instrumental in establishing the national SE exam. After first considering a suggested minimum curriculum for structural engineering degrees, the board decided instead to specify any accredited engineering degree in conjunction with passing a 16-hour structural engineering exam, to be taken by candidates after passing their fundamentals of engineering (FE) and principles and practice (PE) exams. This accommodated the variety of widely available degree programs and, not coincidentally, those represented by engineers already practicing.

But at the time, only the state engineering boards in California, Washington, and Illinois offered a 16-hour structural engineering exam, which made the testing requirement problematic. Creating a new exam was beyond the resources of SECB – not to mention its desired timeframe – so the board looked to other organizations for assistance. The UK-based Institution of Structural Engineers, whose membership exams are rigorous and demanding, was consulted with the idea of authorizing SECB to administer the IStructE exams.

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SE Licensure Efforts NCSEA – National Council of Structural Engineers Associations https://bit.ly/3pMTm2G SEI – Structural Engineering Institute (ASCE) https://bit.ly/3ty9DJW SELC – Structural Engineering Licensure Coalition https://bit.ly/3pPgeP8 Eventually, the board turned to the National Council of Examiners for Engineering and Surveying (NCEES), which follows a similar mission of promoting engineering and surveying and develops uniform standards for state licensing of these disciplines. In fact, NCEES’s Model Engineering Law was an early inspiration for SECB’s work.

NCEES had already developed national FE and PE exams and administered the tests for most state licensing boards. NCEES now offers the 16-hour SE exam to any state that chooses to use it.

As for separate licensure of structural engineers, twelve states (Alaska, California, Georgia, Idaho, Nebraska, Nevada, Oklahoma, Oregon, Utah, and Washington) have either a partial practice act or a title act (restricting what types of buildings require a structural engineer’s design, or restricting the use of the SE title, respectively). Hawaii and Illinois have full practice acts. Additionally, thirteen states (Alabama, Arizona, Delaware, Louisiana, Maine, Massachusetts, Minnesota, New Hampshire, New Mexico, South Dakota, Texas, Vermont, and Wyoming) maintain a roster designation of engineers whose professional license was obtained with a structural emphasis, usually through an NCEES structural examination.

And yet, there is still much to be done. In addition to instituting SE licensure in the states and territories that do not have it now – the most apparent goal – it remains critical to establish and promote the value of a structural engineering license. Structural engineers are highly educated and are trained technicians and motivated professionals who are devoting their careers to increasingly complex solutions to their clients’ needs while advancing the state of their art. Recognition of the specific expertise and contributions that structural engineers can offer elevates the profession and brings a better understanding of its importance to public safety and well-being.

And now, the time has come for SECB to pass the torch to SEI, NCSEA, CASE, and SELC in the quest for national SE licensure. As a final act, the SECB governing board urges all structural engineers, individually and especially as members of professional organizations such as SEI, NCSEA, and CASE, to continue the discussion in all available venues to give the profession the prestige it deserves.■ APRIL 2022

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