STRUCTURE magazine | July 2022

STRUCTURE JULY 2022

NCSEA | CASE | SEI

WIND/SEISMIC

INSIDE: Museum of Motion Pictures CLT Design for Seismic Resistance UCSF Seismic Rehabilitation Renovation Project Lifts Building

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J U LY 2022

Contents JU LY 2022

Cover Feature

SEISMIC COMPLEXITY, SEISMIC INNOVATIONS

By Derrick Roorda, P.E., S.E., and Andrew Rastetter, P.E.

Critical systems and solutions for the Academy Museum of Motion Pictures in Los Angeles include an adapted former department store with retrofitted seismic shear walls, a 150-foot-diameter, orb-shaped spherical addition, three steel pedestrian bridges with sliding and pivoting connections, and structurally decoupled egress stair towers.

F E A T U R E S UCSF CLINICAL SCIENCES BUILDING SEISMIC REHABILITATION By Steve Marusich, S.E., and Andrew Salber, S.E. The circa 1932 Art Deco-inspired Clinical Sciences Building was not originally designed to withstand the significant earthquake ground shaking that it is expected to experience from the nearby San Andreas Fault. UCSF recently completed a seismic rehabilitation, incorporating vertically post-tensioned rocking shear walls to improve the seismic performance of the building.

RENOVATION PROJECT LIFTS BUILDING 12 TO NEW HEIGHTS – PART 1

By Jonathan Buckalew S.E., et al.

Building 12 in San Francisco was constructed in 1941 with two tall stories of riveted steel framing, wood plank floors, and a roof supported on long-span trusses.

The renovation project raises the entire building, adding a basement parking level and a partial 2nd floor. The building is retrofitted with new buckling restrained braces, floor diaphragms, and seismic and wind load collectors.

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, the Publisher, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

STRUCTURE magazine is not a peer-reviewed publication. Readers are encouraged to do their due diligence through personal research on topics.

STRUCTUREmagazine

THE EXCHANGE BUILDING

A NEW VISION FOR AN AGING MIDRISE BUILDING

By Cian Coughlan, MIEAust, CPEng

All Hallows School in Brisbane,

By Andrew Jimenez, C.E., Gina Beretta, S.E., and Marc Steyer, S.E.

Australia, is the oldest secondary

The vision for the new 633 Folsom

school in the state. The existing

building in San Francisco included

library was transformed into The

a vertical expansion of the existing

Exchange Building, requiring

structure wrapped in a striking new

wide-ranging modifications,

skin, doubling the height and square

extensions, and demolition.

footage of an existing building. The

Solutions included a hybrid CLT/

design team used a performance-

steel system, carbon fiber for

based approach. The lateral system

strengthening, and pile caps and

is atypical, comprising construction

ground beams with micropiles.

from three different eras.

C O L U M N S a n d D E PA RT M E N TS Editorial

A CASE for Risk Management

By Roger Parra, S.E.

Codes and Standards

Tornado Effects on Buildings

By Samuel Amoroso, Ph.D., P.E., S.E., Ezra Jampole, Ph.D., P.E., and Troy Morgan, Ph.D., P.E.

Structural Performance

Wind and Seismic Resistance

Structural Design

CLT Diaphragm Design for

By Scott Breneman, Ph.D., P.E., S.E., et al.

Shelter Skelter

By Tom Hadzor, P.E., S.E.

Risk Management

Disaster-Resilient Construction

By Sharyl Rabinovici, Ph.D., Teng Wu, Ph.D., and Mark Chatten, P.E.

Just the FAQs

FAQs on ASCE Standards

By Laura Champion, P.E., and Jennifer Goupil, P.E.

Structural Rehabilitation

Strengthening Concrete Columns

By Mo Ehsani, Ph.D., P.E., S.E.

InSights

3-D Printed Concrete

By Alexander Curth, M.Arch

Structural Failures

Treatment Plant

Failure of Equalization Basin at Water

By Hal K. Cain, P.E., and Michael A. Amos, P.E

Historic Structures

In Every Issue

Advertiser Index Resource Guide – Concrete Products NCSEA News SEI Update CASE in Point

The Schoharie Creek Disaster

By Frank Griggs, Jr., D.Eng, P.E.

Spotlight

Meeting Seismic Demands

Spotlight

SR99 Tunnel JULY 2022

EDITORIAL A CASE for Risk Management

By Roger Parra, S.E.

R

isk is an inevitable part of doing business. Active management of the risk is what creates consistently successful organizations. Risk management is defined as forecasting and evaluating financial risks and identifying procedures to avoid or minimize their impact. As structural engineers, we have a specific set of risks in doing business. We need to 1) identify the risks, 2) evaluate their relative significance, 3) develop procedures to minimize those risks, and 4) monitor the results. I like to view our risk exposure through the project life cycle lens.

Client and Project Selection The most important facet of risk management is client and project selection.

We all have our favorite clients, many of whom are demanding but fair and reasonable. Unfortunately, some clients also present problem after problem on their projects. We need to be willing to walk away from consistently problematic clients which present a disproportionate amount of risk compared to potential profitability.

As a result, you reduce your exposure and improve the mental health of you and your employees involved in those projects. It would be best if you consistently evaluated your clients at the end of each project.

New projects present new challenges and taking them on can be exciting. However, taking on projects outside your area of expertise significantly increases your risk. Therefore, be selective about the projects you are willing to take on. Be strategic if you accept projects outside your wheelhouse and implement measures to mitigate the risk.

Scope, Fee, Contract, and Negotiations Clearly define what you will be doing on the project and what you will not do. Develop a detailed scope of work for all your projects. By doing so, you clearly delineate additional services, and the scope functions as a guide for staff on the project. When developing fees, consider the level of effort and the value of your services. Do not negotiate yourself down on fee. Arrive at your fees using multiple methodologies. Develop a systematic way to gather historical project fee data to use as one metric in developing your fee. Finally, create consistent fee and scope development standards throughout your organization.

Sign fair contracts that proportionally assign the project risks. Contracts can be confusing and complicated. Start by using contracts developed

by American Institute of Architects (AIA) or Coalition of American Structural Engineers (CASE). Do not sign contracts you do not understand.

When negotiating contracts, always be willing to walk away from the project. Without this, you do not give yourself room for negotiation. Before you begin negotiations, you should have 1) terms you are willing to live with and accept as a show of good faith, 2) terms that you are okay with provided some change to the language is made, and 3) the terms that are deal-breakers and will cause you to walk away.

STRUCTURE magazine

When negotiating project fees, a reduction in fees should be accompanied by a proportionate reduction in project scope.

Design Documents and Construction Producing quality design documents inevitably reduces the potential for problems on a project. You must have a reasonable schedule and appropriate fees for producing quality documents. Put the right people with the proper tools on the project. Build on your past quality design documents and look for ways to improve your details and designs based on lessons learned.

Implement a quality assurance/quality management program you follow, one that does not just sit on a shelf/server.

The construction phase is where the rubber meets the road and, often, if we have problems with a project, they happen at this stage. If you selected the right client, project type, have adequate fees, good contract terms, and quality documents, you are in good shape but still need to manage your risk during construction. Take a proactive team-first approach to solving problems that are part of your scope. However, do not look to solve problems outside your scope or expertise. Provide a timely review of submittals and responses to RFIs as construction delay claims are almost always a portion of the dispute. If you have been involved in a claim, you know documentation is a critical aspect of the process and many hours can be spent retracing the issues under dispute. With text messages, instant messaging, emails, and various construction management software, there are many ways we communicate with the team during construction. Implement a detailed, repeatable procedure for documenting decisions and communications.

Culture of Risk Management There are many landmines to avoid at every stage of a project. Creating a culture of risk management in your organization increases your profitability and quality and thus reduces your claims. Culture is defined as a way of life, how we do things, customs, or habits. To develop these good habits, we must: • Develop the framework of risk management for the organization • Commit to managing the risk by both management and staff • Institute a training program for project managers and the soon-to-be project managers • Monitor successes and implement changes The business of structural engineering is unique, and I encourage you to learn more about best practices for managing risk by joining the Coalition of American Structural Engineers (www.acec.org/case).■ Roger Parra is a Senior Principal with Degenkolb Engineers in San Francisco, California, and is the Chair of the CASE Toolkit Committee (rparra@degenkolb.com).

J U LY 2022

7

CODES and STANDARDS Tornado Effects on Buildings Are Target Performance Objectives Consistent with Recent Damage Observations?

By Samuel Amoroso, Ph.D., P.E., S.E., Ezra Jampole, Ph.D., P.E., and Troy Morgan, Ph.D., P.E.

S

evere tornados struck the central and southern United States late on December 10, 2021. The heavy damage and the associated loss of life, which received extensive coverage by U.S. media outlets and piqued the general public’s interest, raised questions regarding the relative risks to structures from various natural hazards, including wind, tornadoes, earthquakes, floods, and fires. The damage from these tornados appeared to the casual observer disproportionate to the structural damage from other hazards such as hurricanes and earthquakes. Moreover, the tornado outbreak coincided with the release of ASCE 7-22, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, which includes a new Chapter 32 related to tornado loads and raised the profile of performance under tornado loads in the minds of practicing structural engineers. This article focuses on how risks associated with different hazards are considered by structural engineers in current design standards and whether the devastation observed in December 2021 is somehow inconsistent with these approaches.

Risks Across Hazards in ASCE 7 Load combinations in ASCE 7 were calibrated initially to provide a level of reliability implied by the existing building stock in service several decades ago. The reliabilities of structures in resisting dead and live loads are the benchmarks for reliability for all other non-seismic loads. Leaving aside code adoption and enforcement, this suggests that, despite improvements to codes and standards over time, the structural failure risk (as represented by the occurrence of first yield in members) embodied in buildings constructed 30 to 40 years ago should not substantially vary from those designed according to current standards.

Observations of damage, tabulations of property losses, and numbers of injuries and fatalities caused by structural failures in earthquakes, hurricanes, and tornadoes do not merely reflect the current status of modern building codes and standards. Instead, they represent a complex mixture of factors, including code/standard evolution, regional and local building code adoption, variations in local code enforcement practices, willingness to increase construction costs, and perhaps the influence of climate change (and the associated increases in the frequency of extreme weather events).

The wind speed and seismic acceleration maps were updated in ASCE 7-10 to reflect a more consistent risk target considering structural failure across hazards. The former uniform-hazard ground snow load maps were replaced with risk-targeted maps in ASCE 7-22.

The target reliabilities for structures of various risk categories are (helpfully) stated directly in Section 1.3 of ASCE 7-22. Notably, the bases for seismic and non-seismic loads are somewhat different.

Target reliabilities for the latter are for the failure of any member, whereas seismic reliabilities are based on the risk-targeted maximum considered earthquake (MCER). According to the standard, as described in Section 21.2.1 of ASCE 7-22, seismic design is expected to achieve a 1% probability of collapse within a 50-year period (or approximately a 2 × 10-4 annual collapse risk). On the other hand, the 8 STRUCTURE magazine

Figure 1. Structural damage to an Amazon facility in southern Illinois.

annual probability of member failure due to a non-seismic load in a Risk Category III building that does not lead to widespread damage progression is stated as being 1.25 × 10-5.

The tornado wind speed maps added to the 2022 standard, combined with the accompanying modifications to analysis methods for tornado loads, are intended to achieve similar levels of reliability for new structures compared to the wind load provisions. There were media reports of a compromise within the ASCE 7 standard committee that limited the application of the tornado provisions to only Risk Category III and IV structures due to concerns about added construction costs in the building industry. In a private communication to the authors, a member of the ASCE 7 committee stated that these media reports included misquotations of ASCE 7 committee members and that the decision to limit the tornado provisions to Risk Category III and IV structures was included after a study showed that only these structures would be controlled by EF-0 to EF-2 tornadoes. The commentary to ASCE 7 states, “Risk Category II includes the vast majority of structures, including most residential, commercial, and industrial buildings.”

Therefore, only a small minority of structures will be subject to the new tornado provisions. Nevertheless, Designers and Owners can elect to use the new provisions for any structure. ASCE 7 represents minimum standards that can be exceeded.

December 2021 Tornado Outbreak The tornados that struck the central and southern United States on December 10, 2021, inflicted heavy damage. Famous examples from southern Illinois and western Kentucky are shown in Figure 1 and Figure 2. In Mayfield, Kentucky, a candle factory was heavily damaged while many employees were reportedly working inside.

The observed damage in Mayfield was classified as EF-4. According to the Enhanced Fujita Scale, a damage-based intensity scale, the

wind speeds associated with EF-4 damage are estimated to be in the range of 166 to 200 mph.

The new tornado wind speed maps in Chapter 32 of ASCE 7-22 indicate that the occurrence of an EF-4 tornado at a building site in Mayfield, KY, corresponds approximately to an event with a 100,000year return period (Figure 3, page 10). The occurrence of wind speeds that severe coming from non-tornadic events correspond to events with 1,000,000 year return periods, which is well outside the range of return periods one should expect to reasonably predict.

The ASCE 7-22 Tornado maps show that the tornado wind speeds at Mayfield for 40,000-square-foot Risk Categories III and IV structures (i.e., 1,700 and 3,000 year return periods) are 82 and 101 mph, respectively, which are substantially lower than the corresponding non-tornadic wind speeds of 113 and 118 mph. These lower tornado wind speeds reflect that ASCE 7-22 considers EF-0 to EF-2 tornados and not EF-3 to EF-4 tornados such as those that occurred during the December 2021 outbreak. Whether the new tornado provisions will impact construction in a place like Mayfield depends on aspects of the tornado wind load calculations other than the wind speed itself.

Computation of Main Wind Force Resisting System (MWFRS) uplift pressures on the windward roof edge of a Risk Category III, flat-roofed building in Mayfield that is 20 feet high, 100 feet wide, and 400 feet long shows that the tornado provisions in Chapter 32 of ASCE 7-22 give lower design pressures than the non-tornado wind provisions in Chapters 26 and 27. The details of the calculation are provided in the Table (page 10). The augmentations of the exposure factor, internal pressure coefficient, and external pressure coefficients that account for special wind loading effects in tornados are not enough to make up for the significant differences in wind speed. Before the modifications to Chapter 32, the 82 mph tornado wind speed would produce only 53% of the pressure of the non-tornado wind speed of 113 mph, as the pressures are a function of the velocity squared. Therefore, for Risk Category III structures and below, the new tornado provisions either do not apply or would not provide more robust MWFRS designs than the conventional wind load provisions. This same analysis for a Risk Category IV structure shows that the tornado roof uplift pressures are 24% larger than those for non-tornado winds. Very few buildings fall into this category. The calculation of Components and Cladding (C&C) loads for roofs could exceed those for non-tornado winds due to variations in the tornado pressure coefficient adjustment factor that depend on roof zone and roof slope.

Mismatch between Tornados and Other Hazards There is an apparent mismatch between tornado casualties and losses and those caused by other hazards. The damage and casualties from the December 2021 tornadoes were certainly newsworthy and appeared to the casual observer to be disproportionate relative to the impacts from other hazards. In fact, more people were killed between 1950 and 2011 by tornadoes than by earthquakes and hurricanes combined, and the Insurance Institute for Business & Home Safety (IBHS) reported that the insured losses from events involving tornadoes occurring between 1997 and 2016 were slightly larger than those for hurricanes and tropical storms. This range includes the especially active hurricane years of 2004, 2005, 2008, and 2012. One would intuitively guess that requiring structural engineers to deliberately consider tornado loads would reduce the disproportionate losses to life and property from tornadoes over the years as existing building stock is replaced. However, this may not be the case since the tornado provisions apply narrowly and may not produce MWFRS loads that control over non-tornado wind loads.

Figure 2. Candle factory in Mayfield, Kentucky, from October 2019 (top) and December 2021 (bottom). Courtesy of Google Earth and Maxar Technologies.

The authors are skeptical that the seeming mismatches between tornado impacts and impacts from other hazards are not a product of random chance. Earthquakes tend to cause many fatalities across a wide geographical region in a single event, but they do not happen very often. Similarly, a small number of hurricanes make landfall each year. On the other hand, tornadoes occur in large numbers, even if the majority of them are on the weak end of the spectrum. If an M8.0 earthquake were to occur on the San Andreas fault in Northern California, it is conceivable that accumulated earthquake damage and deaths could leapfrog tornadoes in an instant.

Given that the tornado damage that makes national headlines is often caused by events that we now would classify as quite rare (i.e., 10,000 to 500,000 year return periods) and that the contiguous United States has not yet experienced another seismic event comparable to the 1906 San Francisco earthquake, it is difficult to draw reliable conclusions regarding the apparent disproportionality of tornado impacts.

Impacts of Climate Change Our estimation of structural risks due to weather-related hazards represents backward-looking snapshots. The spatial distributions and return periods of severe events will evolve over time with an evolving climate.

continued on next page J U LY 2022

9

For example, recent research has shown that while the total number of tornadoes has not increased over the past few decades, their locations have. Tornados are occurring less frequently in the southern and central Great Plains and more frequently in the Southeast, Midwest, and Northeast of the U.S. Warmer oceans supply more energy to tropical cyclones, and storm severities associated with these weather systems are expected to increase.

Based on the latest research, we should expect regular, upward adjustments of design wind speeds for non-tornados and shifting contours on the tornado wind speed maps in ASCE 7 to keep up with climate trends and achieve the underlying risk targets in our designs.

Conclusions Figure 3. ASCE 7-22 Wind Speed Return Periods for Mayfield, Kentucky.

Table comparing MWFRS roof uplift pressures for a Risk Category III Structure in Western Kentucky.

Non-Tornado

Tornado

Risk Category III Wind Speed

V = 113 mph

VT = 82 mph

Exposure Factor h = 20 feet, Exposure Category C

Kh = 0.9 ASCE 7-22 Table 26.10-1

KhTor = 1.0 ASCE 7-22 Table 32.10-1

Topographic Factor

Kzt = 1.0 ASCE 7-22 Section 26.8

N/A

Ground Elevation Factor considering elevation of 480 feet above sea level

Ke = 0.981 ASCE 7-22 Table 26.9-1

Ke = 0.981 ASCE 7-22 Section 32.9

Velocity Pressure (Also, internal pressure qi for roofs)

qh = 0.00256 Kh Kzt Ke V2 qh = 28.86 PSF ASCE 7-22 Equation 26.10-1

qh = 0.00256 KhTor Ke VT2 qh = 16.88 PSF ASCE 7-22 Equation 32.10-1

Directionality Factor

Kd = 0.85 ASCE 7-22 Section 26.6-1

KdT = 0.80 ASCE 7-22 Table 32.6-1

Gust Effect Factor

G = 0.85 ASCE 7-22 Section 26.11

GT = 0.85 ASCE 7-22 Section 32.11

MWFRS External Pressure Coefficient for windward portion of flat roof

Cp = -0.9 ASCE 7-22 Figure 27.3-1 For h/L = 0.05

KvT Cp = (1.1) (-0.9) = -0.99 ASCE 7-22 Table 32.14-1 for MWFRS roofs

GCpi = 0.18 ASCE 7-22 Table 26.13-1 for enclosed building. Positive value chosen since uplift pressure on roof is being considered

GCpiT = 0.55 ASCE 7-22 Section 32.12.2 and Table 32.13-1 for partially enclosed building.

Positive value chosen since uplift pressure on roof is being considered

P = qh Kd G Cp – qi Kd (GCpi) P = -23.2 PSF ASCE 7-22 Equation 27.3-1

P = qh KdT GT KvT Cp – qi (GCpiT) P = -20.7 PSF ASCE 7-22 Equation 27.3-1

Internal Pressure Coefficient

Roof Pressure

10 STRUCTURE magazine

News coverage of recent tornados suggests that these events cause disproportionate structural and life-safety impacts compared to other hazards such as earthquakes or non-tornadic wind events. If the damage is disproportionate, we must answer questions about whether structural engineers are treating all hazards consistently. However, the new tornado wind speed maps indicate that the recent instances of severe tornado damage that have caught our interest were exceptionally rare, with return periods far exceeding what structural engineers typically consider in design. Moreover, the lack of casualty and property loss data for earthquakes limits our ability to say whether these impacts were indeed disproportionate to what would be produced by a similarly rare seismic event in a heavily populated area. Since we cannot definitively conclude that the December 2021 tornado impacts were disproportionate to other hazards or with current expectations of performance, we must ask whether the destruction was acceptable and whether design targets should be adjusted across the board so that scenes like those we saw in southern Illinois and western Kentucky last year are prevented in the future.■ References are included in the PDF version of the online article at STRUCTUREmag.org.

Samuel Amoroso is Senior Managing Engineer at Exponent’s Houston office.

Ezra Jampole is Managing Engineer at Exponent’s New York office.

Troy Morgan is Principal Engineer and Practice Director at Exponent’s New York office.

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structural PERFORMANCE CLT Diaphragm Design for Wind and Seismic Resistance Using SDPWS 2021 and ASCE 7-16

By Scott Breneman, Ph.D., P.E., S.E., Eric McDonnell, P.E., Bill Tremayne, P.E., S.E., Donovan Llanes, P.E., Jonas Houston, P.E., S.E., Eric Mengzhe Gu, Ph.D., P.E., Reid Zimmerman, P.E., S.E., and Graham Montgomery, P.E., S.E.

C

ross-laminated timber (CLT) has become increasingly prominent in building construction and can be seen in buildings worldwide (Figure 1). Specifically, CLT floor and roof panels have become relatively commonplace as a primary gravity force-resisting component.

Now, with the availability of the 2021 Special Design Provisions for Wind and Seismic (SDPWS 2021) from the American Wood Council (AWC), U.S. designers have a standardized path to utilize CLT floor and roof panels as a structural diaphragm.

AWC SDPWS 2021 SDPWS 2021 is the first edition to provide direct provisions for CLT use as an element in a diaphragm or shear wall. To differentiate between CLT and light-frame lateral force-resisting systems, it adopts the terminology sheathed wood-frame for light-frame diaphragms (SDPWS §4.2) and shear walls (SDPWS §4.3). In addition, it includes new sections for CLT diaphragms (SDPWS §4.5) and shear walls (SDPWS §4.6). The 2021 International Building Code (IBC) references SDPWS 2021.

Shear Capacity SDPWS 2021 has a single nominal shear capacity for each set of construction details, vn, defined in §4.1.4 for use with both wind and seismic design. From this nominal shear capacity, the Allowable Stress Design (ASD) and Load and Resistance Factor Design (LRFD) wind and seismic design capacities are determined by dividing by the ASD reduction factor, ΩD, or multiplying by a resistance factor, φD, for LRFD design, respectively, as summarized in Table 1.

CLT Diaphragms SDPWS 2021 §4.5 contains new provisions for the design of CLT diaphragms. When using these provisions, designers use an engineered approach to meet the required design loads defined by the building code (IBC) and ASCE 7, Minimum Design Loads and Associated Criteria for Buildings and Other Structures.

When designing CLT diaphragms, the general requirements for all wood systems in SDPWS §4.1 apply, including limits on when wood members can be used to resist seismic forces from concrete Table 1. SDPWS 2021 design capacity.

Loading

ASD Design Capacity vn ΩD

LRFD Design Capacity φDvn

Seismic

vn / 2.8

0.50 vn

Wind

vn / 2.0

0.80 vn

12 STRUCTURE magazine

Figure 1. The Catalyst building in Spokane, WA (MGA|Michael Green Architecture, Katarra).

or masonry walls in §4.1.5. However, the requirements specific to sheathed wood-frame diaphragms in SDPWS §4.2 do not apply.

SDPWS §4.5.4 Item 1 requires that diaphragm shear forces transfer between adjoining CLT panels and between CLT panels and boundary elements through dowel-type fasteners in shear. Dowel-type fasteners include nails, wood screws, lag screws, and bolts. In practice, nails and proprietary self-tapping screws are most commonly used in CLT diaphragm connections. SDPWS §4.5.4 Item 2 does not permit the diaphragm shear connections to transfer the diaphragm tension forces, such as at chords and collectors. Figure 2, Figure 3 and Figure 4 (page 14) show examples of diaphragm shear connections.

For diaphragm shear connections, the capacities of the dowel-type fasteners (nails and screws) in shear, Z, are calculated using the yield mode equations of the National Design Specification® (NDS®) for Wood Construction §12.3.1. Mode IIIs or Mode IV is required to control the capacity of the diaphragm shear connections. An adjusted design capacity, Z*, defined in SDPWS §4.5.4 Item 1, is the basis for the nominal diaphragm shear capacity of the connection. Z* is similar to the adjusted design capacity, Z´, in NDS Table 11.3.1, except the ASD and LRFD-specific adjustment factors, CD, KF, φ, and λ, are not applied.

Z* = Z x CM Ct Cg CΔ Ceg Cdi Ctn The nominal shear capacity per fastener is: Vn = 4.5 Z* A regular on-center spacing, s, in inches, is often specified for fasteners in such connections. Calculating the nominal unit diaphragm shear capacity (plf ) of such a connection is as follows: vn = 4.5 Z* (12 in/ft) / s The requirements and calculation method apply to the connections transferring diaphragm shear, including panel-to-panel,

Table 2. Force increase factors for CLT diaphragm components.

Component

Force Increase Factor γD Seismic

Wind

Chord splice connections between wood elements where the connection is using fasteners in shear controlled by yield mode IIIs or IV

1.5

1.0

Wood elements and connections between wood elements not meeting the above

2.0

1.5

Steel elements including connections between steel elements

2.0

2.0

panel-to-chord, and panel-to-collector connections. Figure 5 (page 14) shows an example of a CLT diaphragm with components and connections labeled for discussion. The diaphragm shear connections shown include (a) panel-to-panel connections not over framing, (b) panel-to-panel connections over a beam, (c) panel-to-collector connections, and (d) panel-to-chord connections.

The chords, collectors and their connections, (y) and (z) in Figure 5, and other structural components transferring shear, such as the CLT panels themselves, have different design requirements. These components and connections must be designed to a higher required capacity using a force increase factor applied to the diaphragm design force. The required force increase factors are found in SDPWS §4.5.4 Item 3, including Exceptions 1 and 2, and are summarized in Table 2.

The capacities of these diaphragm components are calculated using the provisions of the applicable material design method. The design capacities of wood chords, collectors, and their connections are calculated using the NDS, not the SDPWS nominal capacity (4.5 Z*) and reduction factors.

Figure 2. Panel-to-panel connections with splines at The Canyons building in Portland, OR (Kaiser+Path). Courtesy of Marcus Kauffman.

tolerances and intentional under-sizing of the panels for erection tolerances. When panels meet over framing, as shown in Figure 4, and the framing functions as a shear transfer component between the panels, gaps are of little consequence to the diaphragm shear behavior. However, if CLT panels act as a boundary element and panel-to-panel bearing is used to transfer compressive axial forces, the design must account for the presence of gaps. In such cases, excessive gaps need to be filled with a non-compressible material in the boundary element region to Important Detailing Considerations provide an axial load path.

When using a recessed spline to transfer diaphragm shear forces CLT or Framing as Boundary Element between CLT panels, as seen in Figure 3, Boundary elements for diaphragms there are two types of gaps – gaps include chords, collectors, their between CLT panels and gaps between splices, and their connections to the the shoulders of the recess in the CLT vertical lateral force-resisting system panel and the spline placed in the (VLFRS). Boundary elements in CLT recess. Under design-level loading diaphragms can include steel straps, events, CLT panels in diaphragms framing components supporting the mostly behave as rigid elements; CLT such as steel or timber beams, or however, they can shift in response to the CLT panels themselves. For CLT applied forces, creating deformations diaphragms supported by light-frame at the panel-to-panel connections. In walls, the top plates of the walls below areas of local compression between can act as chord and collector elements. panels, if the compression is resisted In practice, it is often advantageous to by the spline, as shown in Figure 6 use CLT panels, coupled with top side (page 15), it may create a prying or metal straps at panel breaks, as the pribuckling reaction in the spline, which mary chord and collector elements. This may reduce the shear capacity of the is especially true for glulam-supported spline connection. floor systems, as it can be challenging to It is recommended that, in spline transfer chord/collector demands across connections, the gap between panels beam-to-column connections. be not more than the sum of the gaps on each side of the spline to mitigate Connection Gaps and this behavior. This allows the panel Tolerances gap to close and achieve bearing before The details shown in Figure 3 and the spline side member gaps close Figure 4 may be constructed with completely. To follow this recommenmeasurable gaps between the panels. dation, specify that the width of the These gaps may result from fabrication Figure 3. Example CLT diaphragm panel-to-panel connection with a spline. recess in the CLT panels be larger than J U LY 2022

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Figure 4. Example CLT diaphragm panel-to-panel connections over framing below.

the specified width of the spline material. For example, a convenient set of dimensions is a 5⅞-inch-wide plywood spline with a ⅛-inch gap on each side and a nominal gap of 1⁄16-inch between panels.

Design Loads and Force Increase Factors The SDPWS CLT diaphragm design method requires designing the diaphragm shear connections to the diaphragm design forces specified by the building code, labeled here as Fdesign. The 2021 IBC references ASCE 7-16 for the derivation of the appropriate wind and seismic diaphragm design forces and requirements. Following the SDPWS, the remaining components of the diaphragm are designed to increased diaphragm design forces, γDFdesign.

A common question regarding CLT diaphragm design is how to use the SDPWS force increase factors in conjunction with various amplification factors of ASCE 7 for seismic design. For seismic design, calculate the diaphragm design force following ASCE 7-16 §12.10.1.1, which can be described in equation form as: Fdesign = max(Fpx,Fx) + Ω0Fx_transfer

Figure 5. Typical CLT diaphragm components and connections.

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Where Fpx is the inertial diaphragm design force calculated using ASCE 7-16 Eq. 12.10-1 through 12.10-3, Fx is the force in the diaphragm from the structural analysis of the seismic lateral force-resisting system, and Fx_transfer is the transfer force through the diaphragm, when applicable, from one VLFRS component to another. These diaphragm transfer forces occur at out-of-plane offset irregularities of the VLFRS, such as horizontal structural irregularity Type 4 of ASCE 7-16 Table 12.3-1.

Following standard practice, as presented in the Structural Engineers Association of California’s Structural/Seismic Design Manual Vol 1, the redundancy factor, ρ, of ASCE 7 is equal to 1.0 to determine the non-transfer design forces on the diaphragm. Therefore ρ > 1.0 is not included in the diaphragm design force, Fdesign.

Where a significant seismic diaphragm transfer force, Fx_transfer, occurs, ASCE 7-16 §12.10.1 can require applying the vertical seismic forceresisting system overstrength factor, Ωo, to the seismic transfer forces within the diaphragm design force. In this case, the overstrength factor, Ωo, is within Fdesign for the diaphragm design and thus cumulative with the SDPWS force increase factor γD: γDFdesign = γDmax(Fpx,Fx) + γDΩ0 Fx_transfer For structures in Seismic Design Categories C through F, the required design forces for collectors potentially include amplification by Ωo per ASCE 7-16 §12.10.2. However, if the structure is entirely braced by light-frame wood shear walls, the exception to §12.10.2 applies, and Ωo does not apply to the collector design. Otherwise, the collector, collector splice connections, and collector connections to the VLFRS must be designed to the maximum of the three enumerated force levels in §12.10.2 and the increased diaphragm design forces required by SDPWS 2021, shown in Table 3.

As the load increase factor in SDPWS 4.5.4 Item 3 applies to the diaphragm design forces, the amplification of forces by Ωo per ASCE 7-16 §12.10.2 for collector design is not cumulative with the load increase factor, γD, of the SDPWS. This is different than amplification of diaphragm transfer forces by Ωo per ASCE 7-16 §12.10.1, as the transfer forces are part of the diaphragm design forces.

For structures in Seismic Design Categories D, E, and F with certain structural irregularities, the design forces for collectors and connections between the diaphragm and vertical elements may need to be increased by 25% per ASCE 7-16 §12.3.3.4. Similar to the Ωo of §12.10.2, this is an increase in the required strength

Table 3. Design force requirements for collectors, collector splices, and connections to VLFRS.

Source of Design Force

Diaphragm Flexibility

Required Design Force*

SDPWS 2021 requirement for all collectors SDPWS load increase

γ γD max(Fpx,Fx) +γγDΩ 0 Fx_transfer

ASCE 7-16 §12.10.2.1 SDC C through F, when not entirely braced by wood light-frame shear walls 7-16 12.10.2.1 Item 1

Ω 0 Fx + Ω 0 F x_transfer

7-16 12.10.2.1 Item 2

Ω0 Fpx,eq 12.10−1 + Ω0 Fx_transfer

7-16 12.10.2.1 Item 3

Fpx,eq 12.10−2 + Ω0 Fx_transfer

*Use the maximum of applicable forces.

Figure 6. Example of inappropriate detail with a significant gap.

of specific components and not an increase in the diaphragm design force and is therefore not included in Fdesign. For collectors, collector splices, and connections to the VLFRS, γD is 2.0 for seismic design.

Therefore, the design force, γD Fdesign, required by SDPWS, will always be greater than 1.25 Fdesign when required by ASCE 7. For the CLT diaphragm shear connections from the CLT panels directly to the collectors and VLFRS, γD does not apply in the SDPWS; therefore, these shear connections must be designed to 1.25 Fdesign if triggered in ASCE 7-16 §12.3.3.4. Figure 7 shows an example of a collector strap from a CLT diaphragm to core walls.

In ASCE 7, several specific diaphragm types can qualify prescriptively as flexible or rigid. CLT diaphragms are not included in the prescriptive categories. ASCE 7 does allow a diaphragm to be idealized as flexible when: δMDD >2 ΔADVE

Where δMDD is the maximum in-plane diaphragm deflection; ΔADVE is the average deflection of adjoining vertical elements of the VLFRS. ASCE 7 does not provide a method to idealize a diaphragm as rigid by analysis; however, IBC §1604.4 and SDPWS §4.1.7 address this approach.

Given the large size and high in-plane stiffness of CLT panels, it is usually the case that CLT diaphragms can be idealized as rigid if used with wood structural panel shear walls or steel or concrete moment frames. When CLT diaphragms are used with concrete shear walls or steel braced frames, sometimes the diaphragms can be idealized as rigid and sometimes as flexible. Alternative options include a semi-rigid or an envelope analysis.

Further Information This article provides a brief summary of the new CLT diaphragm provisions in SDPWS 2021 and recommendations from the authors on their implementation. The upcoming CLT Diaphragm Design Guide published

by WoodWorks will provide detailed information, including the design of collector and chord details, full examples, and pre-calculated tables of connection capacities. For questions and free technical support related to mass timber and light-frame wood buildings, contact the WoodWorks regional director nearest you (woodworks.org/project-assistance) or email help@woodworks.org.

Acknowledgments Primary funding for the development of this document was provided by the U.S. Endowment for Forestry and Communities and USDA Forest Service.■ Portions of this article were previously posted on the WoodWorks website, March 2022. It is reprinted with permission.

Scott Breneman is Senior Technical Director – Mass Timber with WoodWorks – Wood Products Council (scott.breneman@woodworks.org).

Eric McDonnell (Principal, eric.mcdonnell@holmes.us), Bill Tremayne (Principal, bill.tremayne@holmes.us), Donovan Llanes (Project Engineer, donovan.llanes@ holmes.us), Jonas Houston (Principal, (jonas.houston@holmes.us), and Eric Mengzhe Gu (Structural Designer, eric.gu@holmes.us) are all with Holmes.

Reid Zimmerman is the Technical Director for the Portland, Oregon office of KPFF Consulting Engineers (reid.zimmerman@kpff.com).

Figure 7. Collector strap from CLT diaphragm to core walls at Platte Fifteen building in Denver, CO (Oz Architecture/KL&A Engineers and Builders). Courtesy of JC Buck.

Graham Montgomery is Vice President and Technical Director with Timberlab (graham.montgomery@timberlab.com).

J U LY 2022

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structural DESIGN Shelter Skelter

Adoption of IBC 2018 Shakes Up Storm Shelter Requirements

By Tom Hadzor, P.E., S.E.

W

ith the 2018 Edition of the International Building Code (IBC) being adopted in more jurisdictions across the country, some designers in storm-prone areas may be surprised that their next project requires a storm shelter. Section 423 of IBC 2018 now requires that structures housing critical emergency operations and certain Occupancy E buildings incorporate storm shelters in accordance with the International Code Council and National Storm Shelter Association’s Standard for the Design and Construction of Storm Shelters (ICC 500). The code requires projects such as police stations and elementary Figure 1. Adapted version of ICC 500-2014, Figure 304.2(1). schools (with occupant loads over 50) located in parts of the country with potential tornado wind speeds requirements than ICC 500, so different terminologies are used. of 250 mph to incorporate a storm shelter. Although some design- FEMA P-361 states that safe rooms provide near-absolute protection ers may think their projects are not typically prone to tornados, this for occupants during a storm event. All safe rooms are storm shelters, requirement affects a large portion of the country, as shown by the but not all storm shelters are safe rooms. This difference is because dark shaded area in ICC 500-2014, Figure 304.2(1) (Figure 1).

FEMA requires additional “Funding Criteria” above and beyond the Outside of the prescriptive design requirements such as occupant minimum requirements of ICC 500 when FEMA funds are used to load and site location, the structural requirements can be challenging construct a safe room. Additional detail on the differences between to meet and even more challenging to implement. Like all structures, the ICC 500 storm shelter and the FEMA P-361 safe room criteria creating a load path from the roof structure to the foundation is can be found in FEMA P-361-(2021) Section A1.3.3. paramount for the design of storm shelters. Due to the extreme loading conditions, connections between the roof structure and the ICC 500 Storm Shelter Requirements load-bearing wall system, and connections between the wall system and the foundation, need to be structurally adequate and construct- Many design professionals have been designing storm shelters for ible. This article highlights some of the requirements of the ICC 500, some time due to state laws. For instance, the State of Alabama began the critical elements of storm shelter design, and some typical issues requiring storm shelters to be incorporated into all public K-12 schools that arise during construction that can be mitigated using good com- beginning in 2010 (Figure 2). In addition, Alabama required storm munication between the design team and contractor. shelters in all public 2-year and 4-year higher education institutions a year later. Now that similar requirements have been incorporated into the IBC, more design professionals and owners must become Terminologies familiar with the storm shelter requirements.

First, it is important to discuss some of the nomenclatures used for ICC 500 has requirements for most design disciplines associated enclosures that help protect people from storms. ICC 500 uses the term with a storm shelter project. Bathrooms, egress and ingress, signage, storm shelter. This term refers to detached buildings or rooms/areas within emergency power, and ventilation are just some of the non-structural host buildings designed and constructed according to ICC 500. The ICC requirements for a storm shelter. The architect, structural engineer, 500 is the standard that establishes minimum requirements to safeguard and owner should all be involved in determining the location of the public from high winds associated with tornados and hurricanes. the storm shelter. The owner must regularly maintain storm shelter Technically, suppose an enclosure does not meet all of the requirements elements such as doors and operable louvers. Owners will also be of the ICC 500 (architectural, structural, mechanical, electrical, and fire involved in educating the occupants and performing drills to ensure protection). In that case, it should not be referred to as a storm shelter. the shelter is used correctly.

The Federal Emergency Management Agency (FEMA) uses the term The structural requirements for storm shelters are similar to standard safe room. Similar to storm shelters, this term refers to wind-refuge building design, but the loading conditions are significantly higher. For areas that meet the requirements of FEMA’s Safe Rooms for Tornados gravity loads, the roof live load is 100 pounds per square foot (psf) comand Hurricanes (FEMA P-361). FEMA P-361 has more stringent pared to the typical 20 psf, and increased rain loads should be considered

16 STRUCTURE magazine

Figure 2. Auburn High School – Athletics Building featuring storm shelter design; Exterior (left), Interior (right).

in combination with this roof live load. In addition to the increased roof using interior baffles to ensure that no wind-borne debris can enter live load, the designer must also consider laydown and falling debris the MEP penetration zone during the storm event. hazards. These additional loads would include the collapse of an adjacent structure or host building. Therefore, these loads should be added to the Critical Structural Elements uniform roof live load, and impact factors for laydown and falling debris should be considered per ICC 500.

As all structural designers know, one of the most critical items for For lateral loading, loads are calculated similar to standard build- an effective design is the load path. Load path cannot be stressed ing design, with the following exceptions: Design wind speeds for enough for storm shelter design. The extreme loads from the storm tornados are per ICC 500 Figure 304.2(1), and similar figures are event must be transferred from the roof to the ground in the most given for hurricanes; wind loads shall be based on Exposure Category efficient way possible. Although the individual element design (roof, C for tornado shelters, and Exposure Category B is not permitted wall, foundation) may be straightforward, the connections are typifor hurricane shelters unless excepted per ICC 500; the directionality cally the elements that present the most trouble in completing the factor, Kd, shall be taken as 1.0, and the topographic factor, Kzt, shall load path. Figure 3 shows a non-shelter roof structure attaching to the not exceed 1.0. The enclosure classification for storm shelters can storm shelter cap at a wall bearing condition. The further discussion be determined per ASCE 7, Minimum Design Loads and Associated below references specific items in Figure 3.

Criteria for Buildings and Other Structures. However, atmospheric presMultiple structural systems can be used for storm shelter design. sure change (APC) must be considered for tornado shelters. Based on Wall systems typically consist of concrete or concrete masonry unit experience, pressure change is typically hard to manage with venting. (CMU) construction. Roof systems typically have a solid concrete Therefore, most tornado shelters are designed as partially enclosed cap or concrete topping to meet the wind-borne debris requirements (GCpi = ±0.55), so APC can be neglected per ICC 500. Finally, shield- of ICC 500. Still, the supporting structure can range from metal ing effects from adjacent structures cannot be considered. ICC 500 deck on steel beams or joists, precast concrete hollow core, or precast assumes that the host building and adjacent structures are destroyed concrete beams. The National Wind Institute at Texas Tech University during a storm event, and the shelter is fully exposed. (https://bit.ly/390w5Wa) has produced significant testing data on Besides standard building structural design, additional requirements include wind-borne debris for horizontal and vertical surfaces, connections of host building elements (non-shelter) to shelter elements, and penetration requirements. The designer should study these atypical requirements to ensure that the shelter performs effectively and protects its occupants.

Some architectural elements affect the shelter envelope, and the designer should be involved with the specification and detailing around these elements. For instance, storm shelter doors, windows, and louvers are to be rated per ICC 500. Similarly, penetrations for mechanical, electrical, and plumbing elements larger Figure 3. Non-shelter roof structure attached to storm shelter cap at deck bearing condition. than a specific size must be protected J U LY 2022

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the different roof and wall assemblies shelter have a 90-degree turn upon enterrelated to wind-borne debris. ing the shelter. Please reference Figure 4 The roof systems should be designed for for a representative baffle detail. both the downward gravity load and the uplift loads associated with the increased Moving Towards wind speed. The designer should expect Construction conditions where the uplift forces control the design. When using beams/joists to In addition to construction documents support the roof system, the designer for the storm shelter, design professionshould provide a way for the slab to posials also need to put together a Quality tively attach to the supporting beams Assurance Plan. This plan ensures that for uplift forces, and the beams/joists the contractor knows the storm shelter should be designed for negative bending. requirements and critical elements assoIn Figure 3, please note the headed studs ciated with the design. attached to the joist that positively attach Typically, the storm shelter design the deck/slab system to the joists. Due to must also be peer-reviewed for the the extreme loading, the joist was given disciplines involved (typically architecan increased seat depth to handle shear tural, structural, mechanical, electrical, loads, and uplift forces were provided and fire protection). The peer reviewers to the joist manufacturer to incorporate must be independent and employed into the design. by the owner or the owner’s agent.

The connection of the roof system Figure 4. Storm shelter baffle detail.

Peer review reports must be submitto the wall system can be challenging, ted to the authority having jurisdiction depending on the thickness of the wall system. As discussed previ- (AHJ) as part of the construction permitting process. ously, the designer should expect uplift to control the design. In this Finally, according to the IBC, special inspections are required. It is reccase, the roof-to-wall connection should be designed to carry these ommended that shelter elements be considered for special inspections for loads. In referencing Figure 3, there are two mechanisms to transfer wind resistance according to the IBC. According to the typical standard the uplift forces to the wall: the attachment of the joist seat to the of care, structural observations are also recommended, where special embed plate and the L-shaped reinforcement bars that attach the attention should be paid to the critical shelter elements and connections. concrete topping slab to the wall system. One of the primary failure Finally, communication with the contractor is critical for a sucmechanisms for this type of connection is at the bottom or top of the cessful storm shelter project. The design team should communicate bond beam, so the designer should pay close attention to lap lengths the important elements associated with the shelter, answer questions for development of the reinforcement. The use of hooked bars can that arise in the field, and offer “lessons learned” based on previous aid in developing the bars in more restrictive conditions. experience. Some common issues that arise in the field include: The wall systems used for storm shelter design should be designed • Additional baffles having to be installed for electrical conduit for both in-plane and out-of-plane wind loads. Particular attention penetrations because the electrical subcontractor was unaware should be paid to jamb and header reinforcement around openings. of the maximum penetration size and drilled an oversized hole In addition, the designer should coordinate anchorages of door and in the wall. window assemblies with the wall design. Typically, CMU is fully • Connections of host building elements to the shelter having to grouted to help with wind-borne debris requirements and add weight be removed and re-installed because the installer was not aware to the structure for foundation design purposes. of the requirements of ICC 500 and provided a connection A geotechnical engineer typically dictates foundation systems for storm that was excessive. This connection would lead to additional shelters based on a subsurface exploration of the specific project site. force being transmitted to the shelter during a storm event and Regardless of the system, foundations are typically controlled by uplift for can compromise the design. storm shelters. Similar to the other connection interfaces, the attachment • Improper lap lengths used for connections of roof systems to of the wall system to the foundation is critical to the shelter’s performance. wall systems and wall systems to foundation systems.

Referring to Figure 3, please note the light gage trusses shown for the non-shelter roof support. For this particular project, the rest of Forward Thinking the building had a gabled roof carried over the top of the shelter. This condition falls under ICC 500 Section 304.9, where a host-building As the IBC 2018 Edition continues to be adopted in more and element was connected to the shelter. For these types of conditions, more jurisdictions, design professionals must familiarize themselves the code requires that the shelter is designed to carry the loads asso- with the requirements of ICC 500 and FEMA P-361. Additionally, ciated with the ultimate capacity of the connection in combination architects and engineers alike should be educating owners with the prescribed shelter loads. Any delegated design must be aware on the importance of storm shelters and the costs associated of this requirement, as over-design of these connection elements can with them.■ impact the performance of the storm shelter.

Tom serves as a Senior Project Manager and Branch Manager for the Finally, penetrations larger than a specific size (reference ICC 500 Nashville office of LBYD Engineers, a civil and structural engineering firm based Section 309.1 in 2014 Edition, and Section 306.6 in 2020 Edition) in Birmingham, AL. Tom has contributed a storm shelter design example to the must be protected. This protection is typically accomplished using a Structural Engineers Association of California’s Wind Design Manual published steel baffle, a 4-sided box on the interior of the shelter that would stop in 2018 and currently participates in the ICC 500 workgroup that reviews any wind-borne debris from traveling into the shelter during a storm proposed changes to the ICC 500 (thadzor@lbyd.com). event. This arrangement requires that most utilities coming into the 18 STRUCTURE magazine

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RISK management Disaster-Resilient Construction Rating Systems Drive Systemic Change

By Sharyl Rabinovici, Ph.D., Teng Wu, Ph.D., and Mark Chatten, P.E.

I

n December 2021, the Wells Fargo Foundation made a $300,000 grant to the U.S. Resiliency Council (USRC), a nonprofit that develops natural hazard building performance rating systems. The grant is focused on USRC’s implementation of a rating system for buildings under straight-line wind and storm surge risks.

Rating systems related to energy and environmental impact are now plentiful and standard in the tool kit of design professionals.

They have proved a powerful tool in popularizing sustainable design approaches and investment forward worldwide. However, credible, consistent, and accessible performance metrics for disaster resilience have proved more elusive. The rarity of natural hazard rating systems is a problem because the downtime, injuries, and economic disruption due to damaged buildings in major natural events can be devastating and often do not match up with what the public expects or wants.

The good news is that building more resiliently is more feasible, affordable, and in-demand than ever before, which is good for the engineering profession and owners, tenants, the environment, and our interconnected economy. The accelerating pace and severity of climatedriven hazards compel engineers to do more to help their clients and communities avoid and recover more quickly from natural disasters and the human and financial devastation that accompany them.

Answering the Call Building on efforts such as the work of the Structural Engineers Association of Northern California’s (SEAONC) Existing Buildings Ratings Committee from 2006 to 2014, the ten-year FEMA-funded P-58 project, and recommendations from an Applied Technology Council stakeholder workshop in 2011, the USRC was founded as a 501(c)3 nonprofit organization in 2011. The formation of the USRC has been technically and financially supported by over 100 engineering firms, industry leaders, and professional associations such as the National Council of Structural Engineers Associations and the International Code Council.

A Portland, Oregon building owner receives a US Resiliency Council placard in 2019.

20 STRUCTURE magazine

The WindHub researchers at the University at Buffalo developed an actively controlled multiple-fan wind tunnel to study structural performance during extreme wind events.

Courtesy of Douglas Levere, 2019.

With the help of all its partners, USRC implemented a rating system for earthquake risks in 2017. Like what the U.S. Green Building Council did with its Leadership in Energy and Environmental Design certification (LEED), the system empowers design teams, owners, and investors with scientifically backed yet straightforward information for understanding and improving building performance. “For earthquakes, a USRC rating is the fastest straight-talk risk management service we can provide,” says Jay Kumar of Partner Engineering and Science, Inc., a national provider of seismic assessments for real estate investment due diligence and a USRC Founding Member firm.

Disaster Resilience is the New Sustainability Building performance rating systems can help achieve a world where humans not only have a low impact on the environment, but the environment also has a low impact on us. Wind-related hazards, the focus of USRC’s newest development effort, are major sources of damage and disruption across the nation and worldwide and an accelerating threat due to climate change. The U.S. experienced eighteen billiondollar-plus disasters in 2021, with total damage estimated at over $100 billion. Over three-fourths of that total was related to storms, most notably Hurricane Ida. The grant from financial giant Wells Fargo speaks to the important progress being made in the resilient design field and the potential for rating system information to help markets adjust to very real and increasing natural hazard risks. “What USRC is doing aligns with Wells Fargo’s goals of creating systemic change in how people invest in the built environment,” says John Moon, Vice President of Climate-Aligned Philanthropy and Partnerships Lead for Wells Fargo’s Institute for Sustainable Finance.

Members of the USRC Technical Advisory Committee, or TAC, hope that developing the wind rating system helps the wind engineering community’s attitudes catch up with the leadership earthquake engineers have shown on performance-based design. A rating system can also attract attention to wind’s outsized disaster impacts, catalyze change in construction and insurance practices, and support local adoption of effective codes.

Benefits to Practicing Engineers and Their Clients Engineers can use a wind rating system to work with clients to choose performance goals and evaluate design options to reach those goals in balance with other project objectives and constraints.

One example is using a resilience rating system to provide value to sophisticated clients who want to pursue a more thorough and enduring kind of sustainability. Many end clients, like owners, are aware of the issues but struggle to articulate their goals into something that an engineer can help them achieve. A rating system can become the language for translation by converting the sometimes complex problems and solutions of wind risk into everyday language, as LEED did for energy and the environment. Working with a rating system can provide value in a project’s planning and assessment stages, even if the owner does not elect to pursue formal certification later.

Financial institutions and large portfolio holders are also becoming aware of how climate change affects their overall risk exposure. Natural hazard rating systems complement other Environmental, Society, and Governance (ESG) assessments for investors seeking to understand and manage those risks and improve their climate resilience. In addition, rating systems like USRC’s that utilize performance-based methodologies can accelerate understanding of which building elements are the largest potential sources of damage. Regarding wind risk, for example, USRC TAC members found that choices about cladding are often critical.

By describing outcomes in “plain-speak” on a relative scale, rating systems also help educate clients on needs and possibilities more easily, especially when using state-of-the-art methodologies. For project managers who have to go back and report to superiors about the rationale for complicated design decisions, it can help tremendously to have a framework for conveying the options and expected differences in outcomes.

Development of the USRC-Wind System After the release of the American Society of Civil Engineers (ASCE) Pre-standard for Performance-Based Wind Design in 2019, USRC began a two-year technical advisory process to develop a building performance rating for straight-line wind risks, following the approach it pioneered for earthquakes. The system is primarily scaled to analyze larger commercial, mixed-use, condominium, and apartment buildings.

The technical advisory process involves bringing together a top-tier group of expert volunteers, unified in the goal of crafting a topquality, evidence-driven, open-source rating system. USRC plays the role of a convenor that drives the process forward and brings an understanding of the practicalities of rating system implementation.

The USRC-Wind TAC has utilized the best-available pre-standards, methods, and models and is now researching to document system validity, costs, and benefits via case studies.

In 2022, new efforts are underway to involve stakeholders in rolling out the system so it can influence investment and construction practices in the real world. System success depends on broadening out to educate owners, allying with key influencers, and involving passionate change agents in the built environment field who can help drive system adoption, particularly architects. Wind risk is about the integrity of the whole building, and with so many elements, the architect has a central role.

Key Elements USRC-Wind will allow design professionals to measure the expected performance of new designs in progress and older structures built to earlier building codes and quantify performance from one to five stars in each of three dimensions of expected performance: safety, repair cost, and functional recovery. The first version of USRC-Wind will address straight-line wind (including hurricanes) and coastal storm surge flooding. Future versions may include tornados and other wind hazards.

USRC-Wind is built on open-source engineering science developed

by ASCE and FEMA to provide credibility and consistency to performance evaluations regardless of their location. Among the challenges that USRC-Wind addresses are comparing the performance of buildings of different ages, built according to different codes. By using an objective evaluation methodology with three different performance metrics, not just compliance with local regulations, owners, lenders, and insurers can assess the relative risk of properties or investments consistently across any jurisdiction nationwide.

Engineers interested in learning more about the technical basis of the USRC Wind Rating system can refer to the ASCE Pre-standard for Performance-Based Wind Design and FEMA’s HAZUS User Manual.

Gaining Traction and Benefiting Whole Communities Adoption of building performance rating systems into government and organizational practices is key to their power to transform market behavior. The growth of LEED, for example, was greatly accelerated when local, national, or public-sector entities began establishing policies that require the use of a specific rating system and set minimum acceptable certification levels. In the private sector, a business or institution may specify a rating system as part of their corporate social responsibility policies, a specific project requirement, or a strategy to attract the interest and loyalty of customers and clients who share those values.

Natural hazard rating systems can also support policymaking that changes on-the-ground realities for risk-exposed owners and neighborhoods. As a result, there is a high potential to use USRC’s earthquake and wind systems at the portfolio and community scales. For instance, systems can help target and raise the bar for resilience investments in Low- and Moderate-Income (LMI) areas where exposures to enduring and emerging physical risks can have an outsized impact.

Wells Fargo’s John Moon says, “We are proud to support innovative entities like USRC who are paving new pathways to mitigate risk around emerging physical impacts of climate change.” Having reliable, standardized metrics available can help catalyze investment in resilience by corporate clients and in the communities they serve.

Mr. Moon continued, “Wells Fargo Foundation made this investment because we believe in the power of rating systems to characterize the expected performance of buildings against natural hazards.

Investors need that information to drive an equitable transition to a more climate-resilient society, and that means creating actionable intelligence that helps us build better where it matters most.”■ Sharyl Rabinovici is Director of Strategic Communications at the U.S. Resiliency Council (sharyl.rabinovici@usrc.org) Teng Wu is an Associate Professor at the Department of Civil, Structural, and Environmental Engineering, University at Buffalo (tengwu@buffalo.edu).

Mark Chatten is Vice President at Strategic Initiatives and a Principal with RWDI (mark.chatten@rwdi.com).

J U LY 2022

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just the FAQs FAQs on ASCE Standards What You Always Wanted to Ask

By Laura Champion, P.E., F.SEI, F.ASCE, and Jennifer Goupil, P.E., F.SEI, M.ASCE

T

his quarterly article addresses some of the questions received about structural standards developed by the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE). In addition, questions received from engineers, building officials, and other design professionals are often considered to develop future editions. Following are some questions received by SEI and responses to clarify the provisions.

ASCE/SEI 7: Minimum Design Loads and Associated Criteria for Buildings and Other Structures When do footings need to be interconnected with ties?

Q: ASCE 7-16 Section 20.3.1 Condition 1 states that soils vulnerable to potential failures, such as liquefiable soils, are classified as Site Class F. There is an exception that allows the ground motions (SDS, SD1) to be determined assuming the site is Site Class D, instead of F, and if the period is less than 0.5 seconds. Section 12.13.8.2 (and the International Building Code) requires footings that bear on Site Class E or F soils to be interconnected with ties. Section 12.13.9.2.1.1 has additional stricter tie requirements if the site has liquefiable soils with lateral spreading, bearing loss, or differential settlement issues. It is my understanding that structures with periods less than 0.5 seconds are still considered to be on Site Class F soils, and the requirements to tie footings together in Sections 12.13.8.2 and 12.13.9.2.1.1 still apply. Does Section 12.13.8.2 still require individual spread footings on sites with liquefiable soils to be interconnected with ties when the period of the structure is less than 0.5 seconds? Furthermore, does Section 12.13.9.2.1.1 still require individual spread footings on sites with liquefiable soils subject to lateral spreading, bearing loss, or differential settlement to be interconnected with ties when the period of the structure is less than 0.5 seconds?

A: The exception to ASCE 7-16 Section 20.3.1 Condition 1 does not automatically allow the site class to be set at Site Class D for structures with a fundamental period of vibration equal to or less than 0.5 seconds. One must apply the rules of Section 20.3 to determine the site class and the corresponding values of Fa and Fv. The site class might very well end up being Site Class E. This exception, however, does change the site class.

Considering the requirement for footings to be interconnected with ties, there are two triggers that require the use of foundation ties. ASCE 7-16 Section 12.13.8.2 requires that individual spread footings founded on Site Class E or F soils be interconnected with ties. An additional trigger requiring foundation ties is found in ASCE 7-16 Section 12.13.9. The foundation tie requirements of Section 22 STRUCTURE magazine

12.13.9 are triggered by the structure being founded on liquefiable soils and not by site class. Depending on the amount of movement and bearing capacity loss (see the exception to Section 12.13.9) predicted from liquefaction, Section 12.13.9 may require foundation ties to be provided.

Specifically, the requirement for interconnecting ties for spread footings of Section 12.13.8.2 is triggered by the site class. If the structure is founded on Site Class E or F soils, interconnecting ties are required. If the exception to Section 20.3.1 Condition 1 allows the soil to be reclassified as Site Class D for a structure with a fundamental period less than or equal to 0.5 seconds, then the requirement for interconnecting ties for spread footings of Section 12.13.8.2 does not apply.

Furthermore, the requirement for interconnecting ties for spread footings of Section 12.13.9 is triggered by a liquefiable site and not by site class. Individual spread footings on sites with liquefiable soils subject to lateral spreading, bearing loss, or differential settlement may still be required to be interconnected with ties even when the period of the structure is less than 0.5 seconds.

Where is the seismic base of the building located?

Q: Does ASCE 7-16 Commentary Section C11.2 allow for the seismic base of a building to be located near grade level?

A: As noted in ASCE 7-16 Commentary Section C11.2, the location of the seismic base is affected by several factors. ASCE 7-16 Commentary Section C11.2 states, “For typical buildings on level sites with competent soils, the base is generally close to the grade plane.” So, depending on the specific factors of the structure and location in question, the seismic base of the building can indeed be located near the grade plane. ASCE 7-16 Commentary Section C11.2 also gives a number of examples where this is not the case. You must exercise your professional judgment in determining the location of the seismic base. As noted in C11.2, it is conservative to use the lower elevation when in doubt.

Is ASD conversion of wind speeds still allowed in ASCE 7-16?

Q: Is the Allowable Stress Design (ASD) factor conversion of 0.6 still allowable (or appropriate) in the ASCE 7-16 standard? We understand that one of the major changes between ASCE 7-10 and ASCE 7-16 is an overall reduction in the wind speed, so we wanted to confirm if it is still appropriate to use 0.6 to convert the ultimate speeds to ASD values, for example, to use ASD product tables.

A: The quick answer is, “Yes, the 0.6 factor is still current with ASCE 7-16 to convert ultimate design pressures to allowable stress design pressures.” The 0.6 factor comes from the inverse of the load factor that is rounded down. To be more precise, a conversion of ultimate wind speed

to allowable stress design wind speed can be determined by multiplying the ultimate wind speed from the maps in ASCE 7-16 by 1/1.6 or 0.625.

What is the difference between reducible live load for serviceability and reducible live load for strength design?

circumstances does ASCE/SEI, its affiliates, officers, directors, employees, or volunteers warrant the completeness, accuracy, or relevancy of any information or advice provided herein or its usefulness for any particular purpose.

ASCE/SEI, its affiliates, officers, directors, employees, and volunteers expressly disclaim any and all responsibility for any liability, loss, or damage that you may cause or incur in reliance on any information or advice provided herein.

Q: In ASCE 7-16, Appendix CC, Section CC.2.1 describes a load combination of D + 0.5L for serviceability that states involving settlement or similar ‘long-term’ or ‘permanent’ effects. What is the justification or intent Laura Champion is a Managing Director of the Structural Engineering for reducing the live load by 50%? Is it to account for the transient nature Institute and Global Partnerships at the American Society of Civil Engineers. of live loads, which is suggested by the terms ‘long-term’ and ‘permanent’?

Jennifer Goupil is Director of SEI Codes, Standards, and Technical Initiatives A: There is a significant difference between the loads used to evaluate at the Structural Engineering Institute of the American Society of Civil the strength limit state and the serviceability limit state. Live loads Engineers. given in Chapter 4 of ASCE 7 are intended for evaluating the strength limit states and are intentionally higher than live loads that have been measured during various live load surveys. The live loads in Chapter 4 represent the maximum loads that the structure may see during the life of the structure. The load combinations in Section 2.4 are intended for use in evaluating the strength limit state and not the serviceability limit state.

In Appendix CC, short-term effects, where the full live load is used for evaluation, are described, and long-term effects, where 50% of the live load is used for evaluation, are also described. The use of long-term and short-term is related to the probability that the full magnitude of the load will be present over a given period.

Specific examples are best used to explain these points. For example, cracks in drywall will occur under the full live load • Concrete Repair Mortars even if the live load is only present for • Corrosion Protection a brief period. However, the long-term • Construction Grouts settlement of a structure is not affected by • Waterproofing short durations of the full live load and is • Sealants and Joint Fillers best evaluated under a reduced live load.

• Coatings and Sealers The amount of live load assumed to be • Epoxy Adhesives present to evaluate the serviceability limit • Decorative Toppings state should ultimately be based on pro• Cure and Seals fessional judgment and knowledge of the • Densifiers intended use of the structure in question.

• Structural Strengthening Products Appendix CC makes recommendations based on the experience and knowledge of the practicing structural engineers that make up the ASCE 7 Your single-source provider for restoration, committee.■

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If you have a question you want to be considered in a future issue, send it to sei@asce.org with FAQ in the subject line. Visit asce.org/sei to learn more about ASCE/SEI Standards.

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Visit www.mapei.us for details on all MAPEI products.

This article’s information is provided for general informational purposes only and is not intended in any fashion to be a substitute for professional consultation. Information provided does not constitute a formal interpretation of the standard. Under no J U LY 2022

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structural REHABILITATION Strengthening Concrete Columns An Economical Solution for Non-Ductile Frames

By Mo Ehsani, Ph.D., P.E., S.E.

M

any older buildings include columns that require strengthening. Several scenarios could cause this. In coastal regions and aggressive environments, for example, the corrosion of reinforcing steel results in loss of capacity of the columns. In other cases, the poor quality control during the original construction may have resulted in low compressive strength in the concrete. The author has been personally involved with the retrofit of two such buildings in Florida, where the concrete compressive strength has been below 1500 psi, only a fraction of the strength specified in the design documents. Some of the collapsed Champlain Tower investigations in Surfside, Florida, have also mentioned the “powder-like” concrete in the columns as a potential contributing factor to that failure.

Figure 1. Capacity design concept.

In yet another scenario, before the late 1970s, concrete frames were commonly designed with the beams being stronger than of plastic hinges that can form at the ends of the beams dissipate sigthe columns. When subjected to lateral forces, for example, during an nificant energy, leading to a more desirable ductile failure. In 1983, in earthquake, plastic hinges can form at the ends of such columns. In the recognition of this behavior, ACI-318 required the ratio of the sum worst case of weak columns, flexural yielding can occur at both ends of of the flexural capacities of the columns to those of the beams to be all columns in a given story, leading to the column sway mechanism larger than 1.2. It is well recognized that keeping this ratio even larger and collapse of the building. This is shown with the dashed line in than this specified minimum improves the frame’s overall performance.

Figure 1. In contrast, when the flexural capacity of the columns exceeds Many older buildings in seismic regions constructed prior to the early that of the beams, the failure of the frame is more ductile (beam sway 1980s fail this test and have been designated non-ductile structures. mechanism), as shown with the solid line in Figure 1. A large number For example, in Los Angeles, over 1300 buildings are the subject of an

Figure 2. PileMedic laminates coiled in 4-foot-wide rolls for shipment.

24 STRUCTURE magazine

Figure 3. Original column and beams.

Figure 4. Retrofitted column section and epoxy-coated laminate being wrapped around the column.

ordinance called Mandatory Earthquake Hazard Reduction in Existing Non-Ductile Concrete Buildings. These building owners must retrofit their structures and address these shortcomings over a 25-year time frame that began in late 2017.

This article provides a new solution for enhancing both the axial and flexural capacity of such columns. Implementing the technique is relatively easy, leading to a fast and economical solution with minimal disruption to the occupants. An additional feature of the repair is its small footprint, which minimizes floor space loss due to such modifications.

Fiber Reinforced Polymer Solutions The author introduced the concept of repair and strengthening of structures with Fiber Reinforced Polymer (FRP) products in the late 1980s. In that original approach, known as a wet layup, sheets of carbon or glass fabric are saturated in the field with epoxy. They are bonded to the external surface of the structural element, such as beams, columns, and walls. Within several hours, the materials harden and reach a strength 3 to 4 times that of steel. In addition, the FRP serves as additional tension reinforcement that can contribute to the flexural and shear resistance of the host structure. Numerous applications of this system over the past two decades attest to the advantages of these products.

Most of these applications have been for flexural and shear strengthening of beams. In such cases, the maximum moment is typically at midspan, and there is sufficient distance to the end of the span to develop the full capacity of the FRP. On the other hand, applications of wet layup FRP in columns have been chiefly for confinement and shear strengthening. The maximum bending moments in columns occur at the floor levels. Because FRP cannot be easily extended through the floors, it is difficult to achieve significant axial and flexural enhancement of columns with these products. Furthermore, externally bonded FRP does not increase the stiffness of the column that much. This contrasts with the strengthening of beams, where there is appreciable gain in stiffness of the member after FRP is applied. These shortcomings can be overcome using relatively new FRP laminates.

Over a decade ago, the author introduced a new type of FRP laminate with applications in strengthening columns or piles and pipes. These laminates are constructed with specially-designed equipment. Sheets

of carbon or glass fabric up to 9 feet wide (2.7m) are saturated with resin and passed through a press that applies uniform heat and pressure to produce the laminate (Figure 2). The laminates offer several significant advantages compared to the fabrics used in wet layup applications, as listed below: a) Using a combination of unidirectional and/or biaxial fabrics, the laminates provide strength in both longitudinal and transverse directions; the tensile strength of these laminates can reach 155 ksi (1070 MPa). b) The laminates can be made as thin as 0.03 inches (0.76 mm); this allows them to be bent around a corner with a radius of 2 inches (50 mm) (Figure 2). c) The laminates are manufactured in plants under high-quality control standards; this improves the quality of the finished construction. d) The strength of the laminates can be tested before installation; this assures the design engineer that the specified strength is met, eliminating delays for corrective actions. e) The repairs can be completed much faster in the field.

Figure 5. Interaction diagram for the original and retrofitted column.

continued on next page J U LY 2022

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Figure 6. Details of lateral ties for the longitudinal bars within the depth of the beam.

f ) The number and pattern of the layers of fabrics in the laminates can be adjusted to produce an endless array of customized products that can significantly save construction time and money.

Since the introduction of this system, many agencies have conducted independent tests to verify the efficacy of these laminates for a range of applications. These include a study funded by the National Science Foundation (NSF) and Caltrans for fast repair of earthquake-damaged bridge piers, a study funded by the Nebraska Department of Roads for strengthening deteriorated timber bridge piles, and another funded by the Texas DOT for the repair of corrosion-damaged steel H piles.

However, the most significant investigation was a 3-year study by the U.S. Army Corps of Engineers, which resulted in the military selecting a laminated product to repair submerged piles worldwide. In addition, the U.S. Navy’s website reported that the product was used to repair concrete piles in Ukraine (www.tinyurl.com/PLM-UKR). The U.S. Army Corps of Engineers and the Federal Emergency Management Administration (FEMA) have also singled out these laminates in their 2013 Field Operations Guide as the selected product for repairing columns and piles that may be damaged in a disaster, including hurricane, earthquake, terrorism, and more.

Figure 7. Samples of spacers that can also be used to position longitudinal bars.

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A proprietary system was developed to use these laminates to construct a shell around the column to create a small annular distance.

Reinforcing bars can be placed within this cavity before filling with concrete or grout. The jacket serves as a stay-in-place form that facilitates the construction process and provides significant shear reinforcement and confinement for the column.

Design Example The following example illustrates the application of this technique to retrofit non-ductile frames. The existing frame (Figure 3, page 25) consists of an 18-inch x 18-inch square column, reinforced with eight No. 8 bars. Lateral reinforcement is No. 4 at a 12-inch-spacing along the column height. For simplicity, it is assumed that beams in both directions are 14 inches wide x 26 inches high, and they are reinforced with a pair of #10 bars at the top and bottom. Concrete compressive strength is 4000 psi, and steel reinforcement is Grade 60.

The nominal moment capacity of the column is Mcol = 219 k-ft, and for the beams is Mbeam = 276 k-ft. Therefore, the flexural strength ratio, MR, can be checked as: MR =

2Mcol = 0.79 < 1.2 2Mbeam

This ratio does not meet the minimum value of 1.2 set by today’s standards and requires flexural strengthening of the column. The corners of the column that do not include any reinforcing steel can be cut and removed to minimize the enlargement of the column and loss of floor space. Two new No. 8 bars can be placed at each corner, and these bars extend to the floor above through the slab. Plastic spacers are installed on the column to define the annular space. In this example, 1.5-inch-long spacers are placed in the middle of the column sides (Figure 4, page 25 ).

In this case, PileMedic® laminates are supplied in 4-foot-wide rolls to any desired length (Figure 2).

These laminates are 2 to 3 times stronger than steel.

Typical detail requires the laminate to be wrapped two complete times plus an 8-inch overlap around the column (Figure 4 ). The laminate is cut to the desired length, and an epoxy paste is applied; the laminate is wrapped around the column and bonded

to itself to create a 2-ply shell at a distance of 1 to 2 inches from the face of the column (Figure 4). Additional 4-foot laminates are similarly installed and overlap the previous shell by 3 to 4 inches to cover the full height of the column. Finally, the annular space between the column and the PileMedic jacket is filled with concrete or grout using a pump or the tremie method.

The interaction diagram for the retrofitted column has been calculated and is compared to the original column in Figure 5 (page 25), assuming the grout strength to be 4000 psi. The axial capacity of the column has been enhanced, but its flexural capacity has also been increased to 485 k-ft. Therefore, the flexural strength ratio for the retrofitted frame is: MR =

2Mcol = 1.76 > 1.2 2Mbeam

This is significantly larger than the minimum value of 1.2 and ensures that any plastic deformations are concentrated at the beam ends.

Furthermore, in the above calculations, the increase in compressive strength of the concrete in the column due to confinement by the FRP shell was conservatively ignored. Thus, the actual increase in flexural capacity of the column is even higher than presented in this example.

The jacket is a stay-in-place form that expedites construction, but it also acts as supplementary steel ties, which is a shortcoming in these columns. The laminate provides the equivalent of No. 4 ties at a spacing of 3 inches which is far stronger than the original design.

Depending on the type of laminate used and the number of wraps (2 or more), this contribution can be increased even more. In addition, the confinement provided by the shell increases the compressive strength of the original column and the newly placed grout and the ductility

of the column. Lastly, the impervious FRP shell prevents moisture and oxygen ingress in corrosive environments and significantly lowers the column’s corrosion rate, prolonging the structure’s life. The FRP itself is non-corroding, so any future rain or moisture simply runs off the surface without causing any damage to these jackets.

For the joint region within the depth of the beam, steel ties can be epoxy anchored into the core of the column to provide support against buckling for the newly installed longitudinal column bars (Figure 6 ).

This region can subsequently be encased in concrete. Spacers have been developed that ensure the proper width of the annular space. These can also hold the longitudinal column bars in the desired location (Figure 7 ). The beam-to-column connection must be checked, which may indicate a need for an enlargement of that region. However, an earlier study has demonstrated that as the flexural strength ratio increases, the required lateral ties in the joint region may be relaxed.

The footprint of the proposed retrofit is very small. In this example, the column dimensions were increased by only 3 inches, while the flexural capacity of the column was more than doubled. The entire system is comprised of lightweight materials that can be taken to any floor of the building using passenger elevators. The estimated cost to retrofit a typical column is well below $10,000.■ References are included in the PDF version of the online article at STRUCTUREmag.org.

Mo Ehsani is President of QuakeWrap, Inc. and Centennial Emeritus Professor of Civil Engineering at the University of Arizona (mo@quakewrap.com).

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UCSF CLINICAL SCIENCES BUILDING

Seismic Rehabilitation

By Steve Marusich, S.E., and Andrew Salber, S.E

The UCSF Clinical Sciences Building façade. Courtesy of Bruce Damonte.

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CSB exterior view circa 1930s.

28 STRUCTURE magazine

he Art Deco-inspired Clinical Sciences Building (CSB), located on the University of California, San Francisco (UCSF) Parnassus Heights Campus, originally served as the school of dentistry when it opened in 1932. In the 80 years since, CSB has been in continuous service, providing much-needed clinical, office, academic, and research space for the campus. Unfortunately, being approximately 5 miles from the San Andreas Fault, the building is expected to experience significant earthquake ground shaking, which it was not originally designed to withstand. To reduce this risk, UCSF recently completed a seismic rehabilitation of CSB to extend the life of this vital building for at least another 80 years.

Existing Structural System The seven-story, 108,000 gross square-foot building consists of 4-inch-thick, one-way concrete slabs supported by a complete steel frame founded on concrete spread footings. The façades are 7-inch-thick, board-formed punched concrete walls. Lateral loads are resisted primarily by the nonductile concrete façade and additional 6-inch-thick concrete walls around interior stair and elevator cores.

Performance Objectives Previous seismic studies indicated that CSB posed a severe risk to occupant safety in a major seismic event. Given its historical significance and the limited land available on campus, the University opted to seismically rehabilitate the building. The project had to meet the University’s minimum seismic standards, similar to the Basic Performance Objective for Existing Buildings per ASCE 41-13, Seismic Evaluation and Retrofit of Existing Buildings.

However, UCSF also sought to improve functional recovery time in a moderate earthquake and repairability in a major earthquake to align the building’s performance with its intended use supporting faculty at the adjacent mission-critical hospital. After coordination with UCSF’s Seismic Review Committee, the Damage Control Structural Performance Objective (S-2) in a moment magnitude 7.5 deterministic earthquake on the San Andreas Fault was added to the overall criteria to achieve UCSF’s goals. Since functional recovery is highly dependent on the performance of the nonstructural systems, the criteria also targeted the Operational Nonstructural Performance Objective (N-A) for systems critical to re-occupancy.

Vertically Post-Tensioned Rocking Shear Walls The project utilized vertically post-tensioned rocking shear walls to improve the seismic performance of the building. Unlike conventional shear walls, rocking walls tend to remain essentially elastic, with the walls rotating in a rigid body manner around the base.

This allows the rocking walls to distribute lateral deformations more uniformly over the height of the building. For CSB, the

Post-tensioning anchorage and ducts.

Rocking shear wall elevations.

use of rocking walls resulted in distributed cracking throughout the façade rather than concentrated areas of damage, which was beneficial for re-occupancy after a major earthquake.

The walls only have a limited amount of reinforcement and vertical post-tensioning connecting the walls to the foundation to allow the rocking motion. The reinforcement provides hysteretic damping, while the vertical post-tensioning provides additional stiffness and an elastic re-centering force.

In traditional rocking walls, the reinforcement is typically internal rebar that is unbonded for a few feet above the rocking plane to limit strain demands on the bars. However, for CSB, external damping devices were used to allow easier access for post-earthquake inspections. In addition, because of the modest vertical velocities, hysteretic devices were preferable to viscous dampers.

Consequently, vertically oriented buckling-restrained braces (BRBs) were provided near the ends of the walls. The BRBs were sized to fit within the available floor-to-floor height and limit the ultimate strains to less than 3.0%. The BRBs were attached to the walls with base plates using fully-tensioned threadbars to reduce the possibility of slip at the wall interface.

An unbonded multi-strand post-tensioning system was used for the walls to maximize the available vertical post-tensioning force. For similar bar strain reasons, the vertical post-tensioning was unbonded and ran in ducts the entire height of the walls. The tendons were anchored just above the foundation and at the top of the wall in locations that could be easily accessed in the future. Because the tendons are unbonded, a fully encapsulated system was used for J U LY 2022

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corrosion protection. The vertical post-tensioning was proportioned so that the restoring force would overcome the BRBs compressive strength, ensuring re-centering of the wall.

Wall-to-Slab Connections Interaction between the rocking walls and the existing structure was an essential consideration in the design. The ends of the walls were anticipated to uplift as much as 2 inches. Where the rocking walls were adjacent to interior framing, the existing steel connections were determined to have sufficient capacity to accommodate the anticipated rotational and axial demands without compromising gravity support. However, this approach was not possible where the rocking walls abutted the existing historic façade.

In these locations, the façade would resist the rocking motion resulting in undesirable damage to the façade and Comparative response of conventional and rocking walls. adjacent framing connections.

Laterally stiff but vertically flexible connections were developed It was designed to allow the wall and surrounding floor to move to connect the walls to the adjacent structure to resolve this issue. independently in the vertical direction while enforcing deformaA bent plate was used based on its reliability and predictability. tion compatibility in the horizontal directions. The bent plate was welded to the adjacent collector beams and bolted to the rocking walls to transfer lateral forces. The bent plate was covered with nonstructural concrete to create appropriate acoustic and fire separations.

Conclusion

Flexible wall-to-slab connection detail.

With the rehabilitation project now complete, the Clinical Sciences Building provides UCSF with a modern, inviting, and seismically safe facility to continue its mission of “Caring, Healing, Teaching, and Discovering.” The innovations implemented into the design were only possible through the openness and collaboration of UCSF, the design team, and the contractor. As UCSF works to revitalize its Parnassus Heights campus, CSB will feature prominently as a vital reminder of the University’s past as they look to the future.■ Steve Marusich is a Principal and the Director of Technical Excellence with Forell | Elsesser Engineers in San Francisco (s.marusich@forell.com).

UCSF CSB shear wall layout.

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Andrew Salber is Senior Engineer with Forell | Elsesser Engineers in San Francisco (a.salber@forell.com).

Renovation Project Lifts Building 12 to New Heights Part 1: Design Process

By Jonathan Buckalew S.E., Anthony Giammona S.E., Michael Gemmill S.E., Andreas Schellenberg, P.E., Ph.D., and Joe Maffei S.E., Ph.D.

S

an Francisco’s Pier 70 is located on the city’s southern waterfront on the San Francisco Bay, an area with a rich history of industry and shipbuilding dating back to the 19th century.

Over time, these uses diminished, and many of the buildings fell into disrepair. In 2015, the Port of San Francisco began plans to redevelop the area around Pier 70 to create up to one thousand housing units and two million square feet of office space. Figure 1. Pre-construction (left), Lifted building (right, Courtesy of Plant Construction).

Building 12 is a large and prominent building constructed in 1941 with two tall stories of riveted steel protect the area against sea-level rise and make the new bottom story framing and wood plank floors and a roof supported on long-span a below-grade basement. The lifting of the building required 66 coltrusses. With tall and wide crane bays, the first floor was used for umns to be jacked up simultaneously in small increments over 7 days building ship hulls. The upper floor, called the “mould loft,” was used (Figure 1). A separate article dedicated to construction engineering to build the wooden formwork (molds) around which the steel plate (shoring, temporary bracing, and lifting) will be published in a future hulls were shaped. The 240-foot-wide building had only two interior issue. A building section before and after the lift is shown in Figure 2. lines of columns, with floor and roof spans of 50 to 100 feet. The building’s roof has a distinctive “square sawtooth” profile with vertical Existing Structural System clerestory windows on the north and south sides of each sawtooth, providing natural light at the loft.

The existing roof and 3rd floor are wood sheathed floors supported by The renovation project raises the entire building off its original steel beams that span to 7-foot-deep steel trusses that vary in length from foundation by 10 feet to add a basement parking level and a partial 50 to 100 feet. The new 2nd and 1st floors consist of a new slab-on-steel 2nd floor under the mold loft, which becomes the third floor. In addi- deck over steel beams. Additional lines of columns were added in the tion, the building is retrofitted with new buckling restrained braces, new framing to break up the 100-foot spans. At the basement, new floor diaphragms, and collectors for seismic and wind loads. The steel columns support the lifted existing columns. The entire perimeter renovations maintain the historic and industrial feel of the building of the basement has a concrete retaining wall. All of the existing footwhile creating a modern space for commercial tenants. ings were replaced with new footings and grade beams. The existing The two added floors nearly double the building’s square footage. building lateral system consisted of tension-only angles, column knee After the building construction, including the lift of 10 feet to create braces, and a truss moment frame (the bottom chords of the long span a new bottom story, 10 feet of fill was added to the entire locale to trusses connect to the existing columns). continued on next page

Figure 2. Section before and after the Building 12 lift.

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One of the project’s most significant challenges was a lack of existing structural drawings.

Detailed site investigations were conducted to confirm the sizes and connections of the longspan trusses, built-up steel columns, and wood flooring. Figure 3 highlights the complexity of the existing construction. Luckily, all of the structure is exposed, but it required special equipment to get close enough to determine member sizes by hand with a tape measure. Material testing consisting of yield strength testing and weldability analysis was performed on the steel elements to validate design assumptions.

Due to the lack of existing documents, the project team elected to do a 3-D scan of the building to better define existing elements in the architectural model. The scan indicated that the existing columns were all slightly off (fractions of an inch) from the straight grid lines. The design accounted for these discrepancies prior to construction. After the design Figure 3. Complex existing conditions at the first story (left) and roof trusses (right). was complete and construction began, the building was lifted. When the columns were resurveyed after the The final design imposed the drifts on the truss frame analysis model building lift, it was found that they had shifted from the original 3-D and retrofitted the truss where excessive forces occurred. Three models scan up to 2 inches in the worst locations. This generated numerous were used to design the building. The first model included gravity special conditions in the project: 1) basement wall reinforcement had elements that resisted lateral loads to determine the drift compatreduced clearances, 2) column base plate anchors were out of place, ibility forces in the existing trusses and columns. The second model and 3) steel beam bolt holes did not align with column shear tab holes. The design and construction team learned that when lifting or moving large structures, designs should account for these tolerances and, where possible, the connection details should allow for field adjustments. The contractor should survey the column locations after the structure has been moved and incorporate those into the shop drawings. More discussion on construction tolerances will be included in Part 2 of the series.

Seismic Analysis and Retrofit The project considered using a non-prescriptive seismic design with nonlinear response history analysis (NLRHA) instead of a code-prescriptive design. This would improve the understanding of seismic performance and reduce brace sizes by explicitly accounting for the truss moment frame. The downsides were that it added design costs and triggered a seismic peer review. While the savings associated with reduced member sizes could offset the additional design fees, the peer review would add an additional step to the approval process, creating a risk to the schedule.

The project ultimately decided on a prescriptive design.

A three-dimensional finite element model was built using SAP2000 (Figure 4) and included gravity elements that contributed to the building lateral resistance (i.e., large steel trusses). A two-stage linear dynamic response spectrum analysis was used to design the retrofit. The flexible upper portion of the structure is a buckling restrained braced frame (BRBF) system with a code-prescriptive value of R = 8. The rigid lower portion of the building is a special concrete shear wall system designed for overstrength loads from the BRBF system. All floor diaphragms were modeled as semi-rigid with appropriate properties for timber roof, timber floor (3rd floor), and slab over steel deck (1st and 2nd floor).

Drift compatibility created a problem for the design. Even though the BRBs are designed to resist 100% of the lateral load, forces develop in the existing gravity trusses and columns as the building drifts. The project team explored removing the bottom chord at the column to create a pinned condition, but this resulted in excessive deflections.

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Figure 4. Structural analysis model in SAP2000.

Figure 5. Reinforcing of existing column at BRB connection.

Figure 6. Third-floor assembly detail.

“zeroed out” the stiffness of gravity elements so 100% of the load went to the BRBF system. This model governed the BRB design.

The last model applied overstrength factors (BRB upper bound axial capacity divided by the design axial demand) to design collectors and expected brace forces.

Structural Detailing The existing built-up columns created challenges for detailing the BRB connections, particularly at the location of the existing crane beams, which occur near the added 2nd floor and complicate the geometry.

Figure 5 shows a typical condition at the second floor. The BRB gusset plate (orthogonal to the two-dimensional drawing) is centered on the upper column, which is offset from the centroid of the lower column.

Horizontal continuity plates were placed above and below the BRB gusset plate. Vertical cover plates were welded across the existing column’s flange tips and also welded to the continuity plates. The column cover plates and added splice angles were designed to transfer loads between the offset column flanges and provide continuity for the maximum forces that the braces could deliver.

The third-floor assembly, shown in Figure 6, also presented detailing challenges: • The 2x diagonal plank flooring of the original mold loft, viewed from the underside, had historical and architectural value that needed to be maintained, • 1.5 inches of gypcrete was needed to satisfy acoustic requirements, and • New 2x tongue and groove (T&G) sheathing and plywood were required for the new vertical design loads and the floor diaphragm’s seismic force path. The new 2x T&G sheathing on top of this floor system makes an architectural statement in the sky-lit space of the former loft. Although it is not placed diagonally like the original plank flooring, it reflects the original materials at the loft.

Special design considerations and detailing were needed. Thicker plywood was used to ensure field nailing did not penetrate through the existing sheathing and be visible below.

Boundary nails were not long enough to transfer lateral loads directly to the steel beam nailer, so two sets of boundary nails were required. One set was from the plywood to the diagonal sheathing and another from the diagonal sheathing to the nailer, which was attached to the steel beam with welded threaded studs. The diagonal sheathing was not parallel to the plywood edges. The contractor carefully coordinated plywood nails to avoid gaps between existing sheathing.

Conclusion This project was a great success. The design, lifting, and construction teams worked through challenging as-built conditions that literally moved during the course of the project. The final product (Figure 7 ) showcases the simple beauty of structural engineering for all to see. It also allows the continued use of this building while celebrating a piece of history.■ All authors are with Nabih Youssef Structural Engineers (NYASE) or Maffei Structural Engineers (MSE) in San Francisco: Jonathan Buckalew is a Senior Project Engineer at NYASE (jbuckalew@nyase.com).

Anthony Giammona is a Vice President at NYASE (agiammona@nyase.com).

Michael Gemmill is the Managing Principal at NYASE (mgemmill@nyase.com).

Andreas Schellenberg is an Associate Principal at MSE (andreas@maffei-structure.com).

Joe Maffei is the Founding Principal at MSE (joe@maffei-structure.com).

Project Team Owner: Port of San Francisco Developers: Brookfield Developers Structural Engineer: Nabih Youssef & Associates Structural Engineer Sub-Consultant (lateral system design): Maffei Structural Engineering Architect: Perkins & Will, Pfau Long General Contractor: Plant Construction

Figure 7. Completed project of the main atrium (left) and third-floor “mould loft” (right).

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The Academy Museum of Motion Pictures in Los Angeles joins two structurally unique buildings: a renovation and seismic improvement of the 1930s May Company Building with the addition of the new, spherical shaped theater. The adapted former department store connects with the spherical addition by three steel pedestrian bridges with sliding and pivoting connections, and structurally decoupled egress stair towers. Courtesy of Gabe Guilliams, Buro Happold.

Seismic Complexity, Seismic Innovations Structures for the Academy Museum of Motion Pictures

By Derrick Roorda, P.E., S.E., LEED AP, and Andy Rastetter, P.E., LEED AP

T

The three steel-and-glass pedestrian bridges linking the two buildings demand exceptionally flexible connections. The structural connection strategies allow one building to move and the other to stand completely still. Courtesy of Gabe Guilliams, Buro Happold.

34 STRUCTURE magazine

he Academy Museum of Motion Pictures opened to the public in September 2021, promising to “invite the world into an unparalleled experience of the arts, sciences, artists, and social impact of moviemaking.” Designed

by Renzo Piano Building Workshop with Gensler as executive architect, the Academy Museum of Motion Pictures comprises a significant renovation and seismic improvement of the 1930s May Company Building, renamed the Saban Building, with the addition of the new spherical shaped Geffen Theater.

Visitors enter the 250,000-square-foot Saban Building through the Fairfax Avenue entrance and Sidney Poitier Grand Lobby, which connect to the museum’s exhibition galleries, the 288-seat Ted Mann Theater, Shirley Temple Education Studio, Debbie Reynolds Conservation Studio, Fanny’s restaurant and café, and the Academy Museum Store.

The 45,000-square-foot Geffen Theater, a spherical structure, holds a 1,000-seat hall and adjacent function area, the Dolby Family Terrace.

The museum is the largest in the world devoted to exploring films and film culture and is the only such museum in Los Angeles.

Two Distinct Structural Solutions

collectors required carefully cutting the existing concrete while leaving the perpendicular slab reinforcing in place before carefully installing the large bars that tie the floors to the new shear walls.

Different tasks were required to design the new, 150-foot-diameter theater addition, an orb-shaped sphere. Through an in-depth analysis by the design team, the best solution was determined to be a novel baseisolation system. This unusual design consists of four mega-columns in a system that seismically isolates the structure from the ground. Above, the theater volume is framed by spherical, reinforced concrete walls, capped by concrete slab-on-deck above a spider web grillage of steel beams, which effectively create a very shallow dome to complete the primary structure and provide support for the glass-enclosed terrace above.

To construct the spherical theater, the first step was to erect the temporary steel structure, which supported the precast panels that

Two architecturally different and structurally unique buildings were joined to create a single museum experience. The theatre building and its terrace, for example, are constructed of reinforced concrete. This material was chosen in order to form and cast its precise curves. A partial steel structure was erected to support the exterior precast concrete panels, providing formwork for the structural shotcrete on the inside. A steel and glass canopy encloses the terrace and completes the spherical form.

While in good condition overall, the original May Company Building structure – redeveloped as the Saban Building – presented various challenges for this adaptive reuse. The structural engineers worked with the architects and preservation firm to assess sections of the structure worth being conserved and reused within the historic building. For example, the interior areas of the original structure were in very good condition. However, some portions of the roof slab showed evidence of significant water intrusion, which had caused damage, such as calcification of the concrete and rusting of the reinforcing steel. These sections were selectively removed The theater volume is framed by spherical, reinforced concrete walls, capped by concrete slab-on-deck above a spider for replacement with new reinforced concrete. web grillage of steel beams, which effectively create a very shallow dome to complete the primary structure and The building exterior also presented evidence of provide support for the glass-enclosed terrace above. Courtesy of Gabe Guilliams, Buro Happold. localized water intrusion, particularly damaging connections of the stone façade to the concrete walls. The engineers, architects, and conservation experts from John Fidler Preservation Technology conducted a thorough inspection of all exterior stone panels, identifying all damaged anchors in the building envelope that needed to be replaced.

The Saban Building’s retrofit to meet modern codes required studies and design modifications of the structural elements to improve its capacity and limit movement. Inspections of the existing spread footings showed they were in very good condition and required no changes. The steel gravity framing was also inspected and modeled for analysis. The structural retrofit of the building core consists of seven new interior shear walls extending through the main floor levels. The concrete shear walls rest on new 12-inch micropiles extending to about 75 feet below grade. Micropiles were also used separately to resist hydrostatic pressure and tie down the slab-on-grade below the Ted Mann Theater.

The structural retrofit also included the installation of a system of collectors within the existing The 45,000-square-foot theater building holds a 1,000-seat hall and adjacent function area, the Dolby concrete floor slab thickness. Constructing the Family Terrace. Courtesy of Gabe Guilliams, Buro Happold.

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provided the cladding for the exterior.

Second, the 24-inch-thick reinforced shotcrete walls were placed using the precast as formwork. Third, the steel grillage and slab-on-deck were placed, with the perimeter steel ring beam allowed to slide on top of the shotcrete walls. Fourth, the temporary shoring of the steel dome grillage was released, allowing the ring beam to stretch, which it must do to act as a tension ring. By allowing this expansion, the engineers ensured that the exterior precast was not subjected to the resulting outward displacement, which would have caused it Hinges are included at two locations on the columns supporting the stair structure, to prevent lateral loads from being to crack extensively. Finally, the perim- transferred through them and to accommodate the movement of the theater building without damaging the stair structure. eter of the shallow dome was connected to the shotcrete walls, thereby completing the superstructure. An of mass of the theater structure, with three translational masses and Ansys model was employed to study the spherical concrete theater three rotational inertias. construction sequence and advise the construction team with loading The seismic simulations totaled 84 runs, accounting for seven diffor temporary shoring design and a sequence for their safe removal. ferent ground motions at four orientations (-45, 0, 45, 90) and three X and Y-axis scalars. This was repeated for two different earthquake force levels, DBE and MCE. Then, a separate detailed model of the Analysis and Design superstructure was created in Ansys to complete the design of the concrete reinforcing, using the horizontal shear from SAP analysis Structural and Seismic Performance of the Concrete Shell applied as a horizontal acceleration on the concrete mass.

A combination of finite-element analyses was employed to study the The four mega-columns extend 15 feet below grade, where they forces on this unique form and assess the gravity and seismic forces are supported by two 7-foot-thick pile caps, each with 43 auger cast on the mega-columns and shell structure. Triple-friction pendulum piles of 24 inches in diameter, which extend 90 feet below grade. The (TFP) base isolating bearings were used to mitigate the effect of novel design, employing only eight base isolators set 15 feet above seismic forces on the structure, effectively reducing the seismic force grade, is exposed to visitors as a unique design element. To address on the shell structure above the isolation plane by 80%. An extensive the seismic challenges, the base isolators allow the spherical structure analysis was used to determine the ideal location of the TFP bearings to move up to 30 inches in any direction during an earthquake – yet with respect to the center of mass of the structure and determine at the same time, the historic Saban Building next door is designed the required bearing properties. A series of nonlinear time history to limit movement. And these two different structural approaches are analyses were conducted with simulated seismic ground motions to observably “tied together” with the pedestrian bridges. determine peak building accelerations and movements. Nonlinear response history analyses were conducted in accordance with ASCE Bridging Disparate Structures 7-10, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, to determine maximum bearing displacements and Connecting the base-isolated structure with an existing building introhorizontal shear transmitted to the superstructure above. These analy- duced significant challenges for the structural design team. With two ses were conducted in SAP with a simplified model of the structure very different buildings conceived to be one museum – and designed that included the TFP bearings over pinned supports, rigid links to to move very differently during earthquakes – the connections between individual support reaction points, and a lumped body at the center the two structures, mainly pedestrian bridges – demanded exceptionally flexible connections. The team developed connection strategies to allow one building to move and the other to stand completely still.

The Saban Building connects the new David Geffen Theater with three steeland-glass pedestrian bridges. The bridges are different sizes: the Casey Wasserman Bridge at mezzanine level is the longest at 64 feet long by 16 feet wide and about 8.5 feet tall. Visitors access the theater from the Saban Building via this link.

The mezzanine bridge is provided with gravity support by the Barbra Streisand Bridge on the fifth floor, which spans 60 feet long, 12 feet wide, and about 8.5 feet in height, crossing over from the top floor The finished column hinges were cast from high-grade carbon and stainless steel alloys, providing a streamlined construction and fabrication process and ensuring an aesthetically pleasing final product. of the Saban Building to the Dolby Family 36 STRUCTURE magazine

Terrace above the theater. An additional second-floor bridge spans 45 feet and is 8 feet wide and 8.5 feet tall, providing back-of-house access into the area behind the theater screen.

These primary pathways to and from the Saban Building and David Geffen Theater are anchored to the Saban Building and designed to pivot and slide, moving with the theater building during an earthquake. At the Saban Building, the bridges connect to pivot and bearing connections that provide vertical and horizontal restraint while also allowing the bridges to rotate about a vertical axis when the theater building moves during an earthquake.

At one corner of each bridge, a cylindrical pivot-type connection acts as the center of rotation and locks the bridge to the Saban Building. The bridges’ other corners at the In an earthquake, the theater building moves on the base isolators below, and these hinges allow the stair columns to “lean over” so that the upper portion of the stairs can go “along for the ride.” These joints are Saban Building interface are fitted with simple bearing hinged in two perpendicular axes, effectively allowing rotation and translation in any direction. connections that let the bridge elements move freely in horizontal directions.

The stair towers are supported on the ground though they must also Sliding tracks connect the bridges at the spherical theater building connect to the theater building’s exterior at multiple floor levels for and provide restraint in the vertical and horizontal east-west directions lateral stability. This raised a fundamental design challenge to accomtransverse to the bridge. Yet they also afford free movement in the modate the differential movement anticipated between the theater longitudinal direction. When the theater building moves east or west, building and the ground during seismic events – up to about 30 the sliding tracks push on the bridges and cause them to rotate about inches in any direction, as noted above. the pivot connections at the Saban Building. If the theater building The structural designers evaluated solutions to decoupling the stair moves to the north or south, the bridges slide along the tracks, allowing systems carrying gravity and lateral loads: vertical gravity loads are the theater to move freely without applying any load to the bridges. transferred directly to the ground through long slender columns that The team modeled the wireframe geometry generated in Rhino. This anchor to stationary foundations, which are not seismically isolated. freeform surface modeling software allows for mathematically precise The slender stair columns are 6-inch by 6-inch-wide flanges – I-shaped representation and is then transferred to the structural engineering soft- sections – with unbraced lengths of approximately 25 feet. Lateral ware RISA 3D for analysis and design. More detailed portions were earthquake loads are transferred to the theater building’s structure modeled with finite-plate elements to assess deflection and vibration. through beams at each floor level, which also push on the stairs when Ultimately, all the connections were custom designed to accommodate the theater building moves on its isolators. the forces and displacements calculated as part of the structural analyses Hinges are included at two locations on each column to prevent of the theatre building and the historic Saban Building, in concert with lateral loads from being transferred through them and to accomanalyses of the bridges. modate the movement of the theater building without damaging The rotating and sliding connections are exposed as part of the the stair structure. In an earthquake, the theater building moves on architectural expression. For this reason, the engineers supported the the base isolators below, and these hinges allow the stair columns to architect’s goal of achieving high levels of both form and function, “lean over” so that the upper portion of the stairs can go “along for requiring some give-and-take in the design process. the ride.” These joints are hinged in two perpendicular axes, effectively allowing rotation and translation in any direction.

Inspired by the basic universal joint, the column hinges created Decoupled External Egress Stairs for the stairs were custom designed and fabricated by Cast Connex A stair tower on either side of the spherical theater building is designed to accommodate the anticipated displacements of the theater to provide required emergency egress from the terrace and the theater. building and the calculated gravity forces from the stairs above.

The finished column hinges were cast from highgrade carbon and stainless steel alloys, providing a streamlined construction and fabrication process and ensuring a high-quality final product that is aesthetically pleasing. The hinges will perform over the life of the building.

Construction Assistance

Triple-friction pendulum (TFP) base isolating bearings were used to mitigate the effect of seismic forces on the structure, effectively reducing the seismic force on the shell structure above the isolation plane by 80%.

Buro Happold’s engineers also aided in developing temporary structures for the construction phase, including engineering shoring and formwork for temporary support of the theater’s concrete structure during construction. The team used the same analysis model employed to design the concrete shell to study the distribution of shoring loads during different construction phases and to determine a support strategy in collaboration with the shoring subcontractor and supplier. The team identified heavy-duty J U LY 2022

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shoring locations based on recommendations for installing load cells.

The spherical temporary steel support structure of dozens of vertical trusses gave the first indication to the outside world of the scale and unique form to soon appear on this Los Angeles site. The design capacity of the sloping concrete slab supporting the interior of the theater was evaluated for forces induced by the vertical trusses and was strengthened as required. This allowed the construction of the truss radii and, after that, the final roof substructure.

In addition, the engineers also evaluated modular solutions, as for the bridge structures, and segmentation strategies for efficient flat packing in combined In an earthquake, the spherical theater building moves on a novel base-isolation system with four shipping containers, as used for modular or tempo- mega-columns in a system that seismically isolates the structure from the ground. rary buildings. The glass-and-steel bridge structures were fabricated in Europe and then flat-packed and shipped to the site. adaptive-reuse project undertaken by the Academy of Motion Picture Arts and Sciences. “Adaptive reuse projects extend the lifespan of existing structures, reduce waste, conserve resources, and create a smaller carbon Sustainability and Carbon Reduction footprint than new buildings, as they relate to materials manufacturing Arriving outside of the northern entrance to the completed museum, and transport,” the group stated in a press release. visitors approach the impressive sphere through The Walt Disney Significant to the design of the structural system, the project demonCompany Piazza. This open space presents the thoughtful landscap- strates the embodied carbon benefits of reusing most of the existing ing designed by artist Robert Irwin and works by LRM Landscape May Company Building structure. Buro Happold’s analyses show Architects and the project’s civil designer, KPFF Consulting Engineers. that this reuse approach resulted in an embodied carbon savings of With jacaranda trees, Mexican fan palms, dwarf southern magnolia roughly 50%. While some of these savings may be overshadowed trees, karo shrubs, and vinca minor, the lush plantings hint at the by the investment in other museum structures and operations, the green design systems that suffuse this new destination.

Academy is commended for making positive use of a classic Los The Academy Museum of Motion Pictures has garnered LEED Angeles building that outlived its original use.

Gold certification by the U.S. Green Building Council for its two The Academy Museum of Motion Pictures opened its doors to buildings. Focused on reducing carbon emissions and sustainability the public in late September 2021. Behind its civic and cultural at all scales, Buro Happold contributed to this achievement through significance, the new complex reflects specific structural engineering best practices in structural engineering. contributions that benefit future works, offering input on seismic The Academy Museum’s commitment to green building goals and its solutions for existing and new buildings. adaptive reuse of the 1939 May Company Building – known widely The lessons learned in the design and construction of these paired as an iconic example of the 1930s Streamline Moderne design move- buildings, one historical and one on the cutting edge of architecment associated with Art Deco – were central to achieving substantial ture and engineering, include ideas for retrofitting structures and sustainability gains. The 300,000-square-foot campus marks the third designing new ones. In particular, the project contributes to the practice of structural engineering with a few valuable approaches, especially in reconciling the very different structural solutions required by two newly interconnected structures. These include efficient ideas for strengthening a 1930s steel-andconcrete structure for modern seismic performance and a unique base-isolation system for a new building that allows for significant displacement in an earthquake while supporting a 150-foot-diameter orb-shaped theater.

Inside the buildings, visitors find exhibitions on the history, arts, and sciences of moviemaking. For the engineers, creating With t PS=Ø® With PS=Ø® the vessel for this valuable visitor education was an exceptional experience in No leave-outs Reduces costs PS=Ø® eliminates: making visible how the structural Pour strips Accelerates construction ACI-permitted Type 1 & Type 2 solution works.■ Wall restraint Improves safety ICC-approved Expansion joints

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The Exchange Building

Adaptive Reuse at All Hallows School

By Cian Coughlan, MIEAust, CPEng, NER, RPEQ

The Potter building’s new feature “Drum” stair. Images courtesy of KANE Constructions.

T

he historic All Hallows School is an all-girls secondary school located in Brisbane, Australia. Founded by the Sisters of Mercy in 1861, All Hallows is the oldest secondary school in the state.

Over the years, Brisbane has transformed from a small penal colony formed in 1824 to the third-most populous city in Australia. As the city grew, with high-rise towers now casting their shadows over the campus, so did All Hallows School. The campus has evolved with contemporary architecture seamlessly integrating with the original heritage-listed structures. The Exchange Building continues

this legacy by combining environmentally sustainable design with inspirational learning opportunities.

The Concept All Hallows School identified a need for additional space for their staff and students. But as real estate was limited in this historic, inner-city campus, they looked to their well-loved library, the Potter Building, to provide the required area. The needs included a senior study center, IT support services, offices, a variety of modern teaching and collaboration spaces, and a rooftop function area.

Fulton Trotter Architects were tasked with the challenge and developed the concept of The Exchange Building. The scheme utilized large internal voids and interconnecting stairs to create fluid, interconnecting volumes, with two new external stairs providing a vibrant focal point for the building. The Potter Building would require wideranging modifications, extensions, and demolition to achieve the architectural vision.

Existing Structure

Southern elevation prior to the installation of cladding and curtain wall.

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The original Potter Building was constructed in 1972, consisting of a three-story concrete framed structure with unreinforced brick infill and a brick stair core to the western end. In 1978, the building was extended with the addition of another floor. However, once the contractor was on-site and began peeling back the layers, it became evident that the building was subjected to numerous structural modifications

over the decades that were undocumented, increasing the challenges involved with retrofitting the building.

The original building framing was typically 36-inch-deep × 21-inchwide reinforced concrete beams running north-south at 16.5 feet on-center, supported by 16-inch-square columns. The typical slab was four-inch-thick reinforced concrete. Level two was the lowest suspended level of the four-story building, with plan dimensions of approximately 180 feet × 82 feet. The building stepped at the eastern end, with reduced floor plate dimensions of approximately 141 feet by 52 feet for levels three and four.

Site Challenges The site was heavily constrained as it was in the middle of an active schooling environment with numerous buildings surrounding it. Access to the site is also extremely limited, with only two means of approach: 1) Along a narrow laneway to the west, entered via a heritagelisted and protected stone archway. Not only would the risk of damaging the archway with its limited head height be enormous but, to further complicate matters, the arch is accessed from Ann Street, a very busy, one-way, major four-lane thoroughfare through Brisbane. Obtaining road closure permits from Brisbane City Council would be difficult and expensive.

2) Through a much wider and more accessible entrance to the North. Unfortunately, this entrance also has its limitations as larger construction vehicles cannot make it to the site due to an overhead pedestrian link bridge with limited head clearance located along the route.

To Demolish or Not to Demolish The proposed scheme involved extending the largest floor plate at level two and significantly extending the floor plates above to create a consistent area for each floor of the building, with the original building to the west remaining largely untouched. Finally, an entirely new floor was to be introduced.

The reading stair near completion.

The reading stair. Level 03 under construction with a large void and feature reading stair.

Early schemes considered a traditional structural system of new reinforced concrete (RC) walls and columns with post-tensioned (PT) flat plates. However, as the design progressed, it was determined that existing columns and footings would require strengthening to accommodate the heavier structure. Alternatively, the entire existing building within the footprint of the new development would need to be demolished and re-built.

In addition, the very restricted laydown and working space of this site placed significant restrictions on a conventional solution of insitu concrete framing. Therefore, prefabrication and weight were two primary considerations that drove the final structural solution.

Bligh Tanner, the Structural Engineer for the project, proposed a lightweight Cross Laminated Timber (CLT)/steel hybrid structure to limit the additional weight applied to the existing structure. The hybrid system consists of CLT panels supported on steel beams, with the panels connected to the beams by screwing through pre-drilled holes in the steel flange from below. The lightweight system allowed footings and columns to be reused without expensive and disruptive strengthening. The reduction in weight also meant lower seismic demand, and therefore fewer bracing walls were required, allowing for more uninterrupted spaces and greater future flexibility. It also enabled the majority of the existing floorplates to be retained. A hybrid CLT/steel system was adopted as the original structural grid determined the new column positions. The resulting spans were too large for an entire mass timber solution to be economical.

The CLT and steel system also assisted with the site access and storage constraints. The prefabricated elements could be delivered to a laydown area off-site as required and lifted into place with a single tower crane. The CLT and steel construction resulted in less dust, noise, and workers on-site. In addition, it eliminated the queues of concrete trucks, which would be required for large concrete pours and, therefore, significantly reduced the general disruption to the school. However, in-situ concrete was still utilized for new footings and to form a complex cylindrical Drum structure for the external stairs and new shear walls to the front of the site as the complex geometry was better suited to the flexibility J U LY 2022

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Footings The original building footings were investigated and confirmed to have been founded on strong rock. The lightweight hybrid solution allowed the existing footings to be maintained. Pile caps and ground beams with micropiles were adopted for the new building’s vertical and bracing elements.

Micropiles were chosen due to head height restrictions as a small rig could be used to install the piles. This resolved the site access and head height constraints. Plan dimensions for the pile caps and ground beams could also be kept to a minimum, reducing the amount of excavation required and potentially undermining adjacent existing footings. Micropiles could also efficiently resist the uplift forces under the new bracing elements.

Level 03 under construction.

Construction Surprises

Carbon fiber strengthening to level 03 RC beams.

of concrete. Pre-cast concrete panels were also adopted at the back of the site to form a bracing wall. The panels were cast in lengths to suit the tower crane lifting limits with their vertical edges stitched together with steel plates field welded to plates cast into the panels.

Strengthening Where the existing structure was being altered or the design loads increased (for example, where an internal span became an end span due to demolition), carbon fiber was provided to strengthen the structure without impacting floor-to-floor heights or service routes. Strengthening consisted of pultruded carbon fiber reinforced polymer laminates bonded to the soffit and top side of beams and slabs. Several beams were also strengthened for shear with sheets of unidirectional stitched carbon fiber fabric. In one case, where the post-demolition deflections required a higher degree of control, a beam was instead strengthened by providing a steel beam below and pre-loading the beam with a flat jack to take over the dead load acting on a concrete beam. Once the new steel beam had taken the load, the space between the steel and concrete was packed with steel shims and grouted. The adjacent span of the concrete beam was then able to be demolished.

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Several surprises were uncovered once the contractor, KANE Constructions, was established on-site and began to strip the existing building finishes. Those surprises included a previously assumed slab on grade being suspended over a crawl space and old floor voids filled with sections of concrete slabs cast on metal deck. It was also discovered that one of the floor plates had been extended previously, with an additional bay being added.

It was determined that an external beam that would have previously been a gable end beam supporting brick cladding was, at some stage, turned into an internal beam.

Additionally, the floor levels had been raised not once but twice on the same floor, with two distinct layers of topping uncovered, totaling six inches. Consequently, the design needed to be revisited as the assumption of an eight-inch slab and 36-inch beam was incorrect. The beam depths on this floor were six inches less than previously assumed, and their loads had already been significantly increased.

Two of these beams showed signs of distress with flexural cracking at their midspan. The beams were propped, GPR scanned, and concrete removed locally to confirm reinforcement sizes. It was determined that the bars had not yielded, and the beams’ capacity was sufficient to continue strengthening as planned once all cracks had been epoxy injected. Steel dowel bars were also drilled and epoxied vertically from the top surface and into the beam to allow composite action between the topping and the beam.

The successful delivery of this complex and adaptive reuse project has delivered additional space and functions for the original Potter Building while minimizing the impact of new construction on the environment and saving a lot of construction waste going to landfills.

Completed in June 2022, the new Potter Building will be the first education building in Queensland constructed using a hybrid CLT/Steel system.■ Cian Coughlan is a Senior Structural Engineer at Bligh Tanner’s Brisbane office.

A NEW VISION for an Aging Midrise Building

Renovation and Vertical Expansion Breathes New Life into 633 Folsom

By Andrew Jimenez, C.E., Gina Beretta, S.E., and Marc Steyer, S.E.

B

uilt in 1967, the original building at 633 Folsom Street in San Francisco was developed by The Swig Company as a modern office building for the telecommunications giant AT&T. Fast forward to the 2010s: AT&T had long since moved out of the space, and the aging midrise building was no longer attractive to tenants looking for a corporate home in San Francisco’s South of Market (SoMa) district.

So instead of selling the outdated building or redeveloping the site with new construction, Swig saw an opportunity to give the building a facelift, nearly double its square footage and height, and make it the star of their portfolio. The vision for the new building included a vertical expansion of the existing structure wrapped in a striking new skin. To make this distinctive concept economically feasible required a dedicated design and construction team capable of an equally unique approach.

Challenges Call for a Courageous Solution The vertical expansion itself posed several significant design challenges. To compete with newer office buildings, the new floors would need to offer long-span, flexible workspaces, much wider than the 20-foot column grid in the 60-year-old structure. At the lower levels, the existing floor-to-floor height left little room for new mechanical systems, let alone any structural strengthening. In addition, the building’s engineering was outdated, particularly with respect to modern seismic detailing. Finally, assessing the new structure was complicated by the multiple layers and eras of construction. After careful consideration, the design team met these challenges by thoughtfully reimagining a new structural layout and designing efficiencies afforded by a performance-based approach using nonlinear response-history analysis.

Innovative Structural Approach

Figure 1. Structural Stacking Diagram – Exploded diagram showing stacking structural elements.

Doubling the height and square footage of an existing building is not trivial from a structural perspective.

Before conceptualizing the new design, Tipping, the structural engineer for the project, had to fully understand the existing structure and architectural program constraints. The structural design had to consider these limitations in framing and walls to support the expansion (Figure 1). continued on next page

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The building consisted of two-way lightweight concrete slabs supported by normal-weight concrete columns and ten-inch-thick normal-weight concrete core walls. The columns were supported by spread footings and the central core walls on a mat slab. The original lateral system comprised concrete moment frames around the perimeter and concrete shear walls at the core. In the mid-1990s, a voluntary seismic retrofit was performed to strengthen the existing core walls and mat foundation. A twelve-inch-thick layer of shotcrete was overlaid on four existing concrete walls above grade. Meanwhile, micropiles were added to abandoned basement hallways and repurposed as anchor blocks to provide additional overturning resistance.

The vertical addition called for an expanded elevator-and-stair core adjacent to the original building’s vertical circulation core.

This became the natural location for new concrete shear walls, adding strength and stiffness to the existing floors (Figure 2).

The core expansion was designed to be independent of the existing core without coupling beams connecting the two. Coupling would have required intrusive and costly structural detailing with relatively little benefit to structural performance. Consequently, the addition of collector elements was required at the existing floors to engage the new core walls under seismic loading. To maximize ceiling height and avoid conflicts with MEP routing, carbon-fiber-reinforced polymer (CFRP) was applied to the top of the slab to complete the load path from the existing diaphragm to the new core walls.

The existing core walls were vertically extended to gain lateral support at the new floors. The upper story of existing walls was demolished to lap existing with new reinforcement over the story height, providing flexural continuity. An analysis of structural response largely drove this solution. Due to better alignment between centers of mass and rigidity, extending the existing core walls to the new roof produced less torsional response than extending the new core expansion to the roof.

Architectural project goals drove the framing layout of the upper floors. Existing perimeter columns had a built-in mechanical chase that was rendered obsolete by the building redesign. Tipping recognized an opportunity to repurpose the chase as an infill concrete column that would support new wide flange columns, thereby minimizing impact to floor space at the existing stories. The structural engineering team laid out the framing to maximize the span from perimeter columns to core walls, reducing the number of interior columns and creating

Figure 2. Core Diagram – Lateral force resisting system consists of existing concrete core walls and a new core expansion. Vertical addition is supported by an extension of the existing core walls. Both new and existing micropiles resist overturning loads at the core base.

a more flexible space. As a result, only 8 of the remaining 24 existing interior columns were employed to support the building’s new floors.

Moreover, the new floor framing layout minimized the architectural impact of column strengthening at the existing floors and mitigated the high cost of foundation strengthening.

Existing interior columns supporting new floors received a six-inchthick shotcrete jacket to increase gross area and confinement. All other interior columns were wrapped in FRP to increase ductility.

Lastly, steel collars were added around the columns below the existing lightweight concrete slab to address the potential of slab punching around the columns in a seismic event (Figure 3). All columns with added load required foundation strengthening (Figure 4). Owing to the congested site, all perimeter footing retrofits required an offset pilecap with eccentric piles.

Analysis Validates Atypical, Lean Design

Figure 3. Column Retrofits – Existing interior columns are strengthened with a shotcrete jacket or CFRP wrap. All columns are topped with steel collars to catch the slab in the event of a punching shear failure.

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The vertical expansion of the building added square footage and new stories, which triggered a complete code upgrade to ensure that all new and existing structures satisfied current code requirements. While early schemes for seismic strengthening of the vertical addition were based on elastic analysis, Tipping relied on a performance-based approach, using nonlinear response history analysis (NLRHA) to justify the final structural design.

The structural engineering team determined that NLRHA delivered the optimal solution for 633 Folsom. The lateral system is atypical, comprising construction from three different eras.

Because both material and geometric nonlinearity influence holistic structural behavior, the design relied on careful modeling of all material properties and elements of the lateral force-resisting system. Expected material properties were accounted for in the NLRHA, whereas a response spectrum or equivalent lateral force analysis would have dictated that only nominal properties were considered. By taking advantage of expected material properties

Figure 4. Interior Foundation Retrofit – New micropiles extend through the existing foundation for additional vertical support in select locations.

using a performance-based approach, the designers could account for the existing structure’s inherent strength and contribution to the overall building response. This approach was key to minimizing the strengthening required for the existing structure and the new elements.

The lateral analysis was conducted with Perform 3D. Wall fiber sections were used to model concrete and shotcrete shear walls. Existing perimeter moment frames were modeled with flexural hinges at beam ends, while the columns were modeled to be elastic. Separate calculations confirmed that the columns did not exceed their nominal moment strength under the maximum force demands from the analysis. The designers accounted for soil-structure interaction at new and existing mat slabs by modeling a grid of beams spanning to soil springs and micropile elements. Rigid diaphragms were assumed at all floors, and explicit collector elements delivered seismic load to the core walls. All gravity columns were modeled as pinned-pinned columns with dead loads on them to capture additional P-delta shear during an earthquake.

Observing and considering the building’s specific response provided insight into where additional strengthening was necessary. Allowing for some yielding and applying displacement-based instead of strengthbased acceptance criteria helped minimize the impact on the existing building and the project budget. Further, the team modeled soilstructure interaction, softening the response and lowering the demand compared to a fixed-base building. Early phases of design called for the new core expansion to extend the full height of the building. The nonlinear analysis model validated the approach for new shear walls

to extend upwards from the existing core only, while also removing one wall from the initially proposed core expansion. Savings in cost and materials were also made at the foundation of the core expansion, reducing the initially planned number of micropiles at the new mat foundation by more than half, from 34 to 16.

Performance and Value Renovating and vertically expanding 633 Folsom demonstrates that undesirable, obsolete buildings can be significantly transformed, delivering a high degree of performance and value through thoughtful analysis and design. Public response to the transfigured structure has been overwhelmingly positive: the tech company Asana leased the entire building as its new headquarters before construction was complete. In addition, the project received a 2021 Structural Engineers Association of California (SEAOC) Excellence in Structural Engineering Award. Performance-based engineering and a collaborative process among the design and construction team reinvented the building, which again inspires its users and contributes to a vibrant community.■ All authors are with Tipping Structural Engineers in Berkeley, CA. Andrew Jimenez is a Project Manager (a.jimenez@tippingstructural.com).

Gina Beretta is an Associate (g.beretta@tippingstructural.com).

Marc Steyer is a Principal (m.steyer@tippingstructural.com).

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INSIGHTS 3-D Printed Concrete 1939 to Present

By Alexander Curth, M.Arch

I

n 1939, inventor William E. Urschel created the world’s first 3-D printed building behind a small warehouse in Valparaiso, Indiana. The following year he would file a series of patents for a “Wall Building Machine” (Figure 1). This simple yet ingenious machine would be used to fabricate multistory structures with integrated reinforcement and a self-supporting dome, all printed in concrete without formwork. In the late 30s, this process might have been described as layered, horizontal slip forming. With these early prototypes, Urschel matched much of the innovation we see today in Large Scale Additive Manufacturing (LSAM) 60 years before the first modern examples of construction 3-D printing were published by Behrokh Khoshnevis in the early 2000s (Khoshnevis 2004). Urschel explored geometric design freedom, reinforcement, variable extrusion, material compaction, and, most notably, created full-scale buildings, the very first of which is a still an occupied, working structure. A look at the details of Urschel’s Wall Building Machine (Figure 1) provides a critical lens for engineers and designers to view the rapidly growing industry adoption of 3-D printing technology.

Why 3-D Printing for Buildings?

Figure 1. The Wall Building Machine operates radially around a central axis, its footprint far smaller than that of the building it would create, much like a modern construction crane.

Courtesy of David Tippold – Urschel Laboratories, Inc.

The promise of 3-D printing lies in reducing cost by minimizing labor, construction time, and material by using a computer-controlled gantry or robot arm to deposit continuous layers of material. As a computationally-driven construction process, it is possible to achieve structurally sound, geometrically

complex, mass customizable designs without significant increases in project cost. The theoretical results of this process are beautifully efficient, low embodied carbon structures, which can be built in a matter of days anywhere on earth. However, while there are exciting research examples of spanning structures printed without formwork and click-together bridges (Curth 2021, Bhooshan 2022), the reality is that most completed 3-D printed buildings today are extruded vertical walls filled with conventional reinforced concrete and capped with timber frame construction. Looking to Urschel’s Wall Building Machine, one can recognize where the fundamental challenges have persisted and the exciting next steps in materials, reinforcement, and form for large-scale additive manufacturing.

Materials Figure 2. Molding elements were added to the extruder, creating interlocking channels between layers and emphasizing the distinct appearance of a printed structure. Courtesy of David Tippold – Urschel Laboratories, Inc.

Large-scale structures have now been printed in plastic, foam, and even earth; however, most are made from cementitious mixes.

Cementitious mixes present several clear

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Figure 3. Integrated wire reinforcement was deposited in much the same way as the modern entrained wire printing system. Courtesy of TU/Eindhoven.

benefits shared with other modes of concrete construction, such as cost and workability, while also presenting particular drawbacks such as cold joints between layers of printed material. Urschel’s system included interlocking channels between layers to mitigate cold joints (Figure 2).

Now, the issue of anisotropy between layers is mainly being worked on at the mix design level. Although, printing companies are beginning to add grooves to the surface of their extrusions to increase surface area and, in turn, friction between layers to reduce the mechanical disadvantages of cold joints, much like Urschel did in 1940.

The material used in the Wall Building Machine was not a specially formulated, quick-setting mortar mix like the 3-D printing materials used in most modern systems. Instead, it was a dry concrete mix with sizable aggregate used widely in the local construction industry of that time. The key difference between this system and modern systems is that Urschel’s machine included an automatic ramming mechanism (Figure 3) that compressed the concrete mix between spinning disks, consolidating and smoothing each layer of the print as the material was extruded.

After 20 years of research, large-scale 3-D printing companies are only beginning to find ways to move aggregate greater than 1⁄8 inch (4mm) through their extruders. Smoothing is still a continual challenge because only static, not spinning, smoothing mechanisms are employed. In addition, because the problem of pumping admixes that combine large aggregate, fast cure time, and proper workability has not been resolved, most modern prints are simply mortar, making a typically printed wall section both lower strength and more carbon-intensive than a traditionally poured concrete one.

As printing companies improve their pumping systems and mix designs, they may begin to consider Urschel’s approach of designing the extruder to use conventional building materials instead of designing materials to match an unconventional building system.

reinforcement requires greater consideration. Urschel’s patent drawings from 1941 show a mechanism for real-time deposition of embedded steel wire reinforcement (Figure 3), which is also the subject of several recent projects at TU Eindhoven (Mechtcherine 2021). Other research groups have tried similar approaches with rolls of steel mesh and even staple guns which follow the extruder, placing interlocking reinforcement between layers of material. Fiber-reinforced cement mixes have also been printed with satisfactory results. Most recently, researchers at Ghent University have produced modular spanning units with internal channels for post-tensioned cables (Vantyghem 2020).

In general, geometric flexibility allowed by 3-D printing makes it possible to detail rebar reinforcement in ways that meet conventional building codes, though rebar cold joint extensions are manually positioned and grouted into vertical cavities in printed wall structures instead of being cast in place. APIS Core's multistory structure completed in 2019 is the largest scale example to date of this hybrid approach (Figure 4).

Reinforcement Beyond compression-only structures, concrete requires reinforcement. Most 3-D printed wall structures are hollow, making the placement and grouting of vertical rebar elements straightforward post-printing. However, horizontal 48 STRUCTURE magazine

Figure 4. Under construction, a modern 3-D printed building features relatively conventional reinforcing and a radial printing system built by Apis Core. Courtesy of Apis Core.

Looking to the Future Beyond the deposition of building elements themselves, printing offers the possibility to create complex formwork at a low cost.

While Urschel only explored the construction of walls and domes (Figure 5), both positive and negative space can be considered in a continuous design process. Foundations, and floor slabs, typically two of the highest embodied carbon building elements, which typically require formwork for their construction, can be fabricated in place with printed geometry acting as a mold for concrete or rammed earth. In addition, forms for windows and doors can be printed in a reusable material like earth or plastic to be removed after the structure is complete, allowing for continuous printing (in Figure 1, Urschel’s hand-placed wooden framing to support printing above the door). A printer may also be used to create recyclable, customized scaffolding for workers or other machines within a hybrid construction of printed and human assembled elements.

With advances in physical and simulation tools for large-scale printing, it is now possible to produce structures far more efficient and functional than what Urschel created in 1939. The challenge continues to be adapting the advancing tools to the realities of materials, structural design, and building code.■ Alexander Curth is a PhD student in the Design and Computation Group at MIT. He works on developing tools for democratization and access in the world of architectural additive manufacturing (curth@mit.edu).

Figure 5. A domed structure is being constructed by one person and a Wall Building Machine, a feat yet to be accomplished with a modern printing system at this scale. Note the decreasing layer height as slope increases, a technique now used in modern 3-D printing systems to manage overhanging layers. Courtesy of David Tippold – Urschel Laboratories, Inc.

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structural FAILURES Failure of Equalization Basin at Water Treatment Plant

By Hal K. Cain, P.E., and Michael A. Amos, P.E.

C

ain and Associates Engineers and Constructors, Inc. of Mobile, Alabama, were engaged by a Law Firm from Knoxville, Tennessee, to investigate the cause and origin of the collapse of a large Wastewater Treatment Plant Equalization Basin in Gatlinburg, TN. On April 5, 2011, following a period of very heavy rains, the equalization basin at the wastewater treatment plant experienced a catastrophic failure when the east wall of the structure collapsed (Figures 1 and 2). As a result, over one million gallons of wastewater were released. In addition, two employees working in a flow control building located approximately 8 feet in front of the Figure 1. Overall view of the collapsed east wall of the Gatlinburg Wastewater Plant equalization east wall were killed in the incident. basin looking from south to north.

The equalization basin was a 123- by 63-foot reinforced concrete structure with five baffle walls that allowed waste- conditions indicated that the structure did not meet ACI – 318, water to channel through the system. Three baffle walls intersected the Building Code Requirements for Structural Concrete and Commentary, east wall, and two intersected the west wall. The structure was 30.5 feet and ACI 350, Code Requirements for Environmental Engineering high from the base mat to the top of cantilever walkways on the north, Concrete Structures. This was the consensus of multiple experts who south, and east sides. The structure was constructed using cast-in-place reviewed the failure. reinforced concrete set in multiple pours. The exterior walls were 18 The design drawings did not indicate any construction joints in the inches thick, and the baffle walls were 12 inches thick. The total con- basin other than between the base slab and the walls. Due to the size crete volume above the base mat was approximately 1,100 cubic yards. and complexity of the basin, it was not reasonable or practical that The structure, designed and constructed between 1994 and 1996, it be constructed in a single pour. The ACI 318 and ACI 350 codes was placed in service in 1997 and operated until the failure. require that construction joints be located and detailed. ACI 350 states that construction joints “should be located so as to least impair the strength of the structure, to provide logical separation between Design and Construction of the Basin segments of the structure, and to facilitate construction.” This is It is unclear exactly which design code was used for the original certainly intended to be a combined effort between the engineer and engineering basis. However, analysis of the structure and the as-built construction personnel, but it should be initiated in the project’s design

Figure 2. Front view showing the collapsed east wall of the Gatlinburg Wastewater Plant equalization basin.

50 STRUCTURE magazine

Figure 3. Cracks in the north wall caused by the excessive displacement of the wall. The added buttress is shown to the right of the photo.

phase. In this case, it was initiated in the field based on coordination between the construction superintendent and the engineer’s representative. Reinforcing bar couplers were used to connect the baffle walls to the east and west perimeter walls.

There does not appear to have been any consideration given in the initial design for corrosion or chemical attack.

Per ACI 318, Section 1.4.4, Concrete Sanitary Engineering Structures should be designed following ACI 350 to provide higher resistance to chemical attacks and to minimize the possibility of cracking and deflections. In addition, special emphasis is placed on the design and placement of joints. Per ACI 350R89, factors used for reinforcement in flexure are increased by 30%, and factors for reinforcement in direct tension are increased by 65%. ACI 350R-89 requires that joints be shown on the design drawings and that joints should be placed so as not to impair the strength of the structure. ACI 350 also addresses the importance of concrete quality.

The ACI 350 code specifically addresses exposure to corrosive liquids, and that allowable loads are based on the premise that reinforcing steel will be in contact with those liquids. This is one basis for this code and a major factor in justifying the use of ACI 350 and the additional costs that this would entail.

Figure 4. Cracks in the upper walkway along the north wall. This was caused by the excessive displacement of the wall.

in a pinned-pinned condition rather than a more rigid fixed-fixed condition. Further, the walkway that ran the entire length of the north wall and that extended 4.5 feet from the face of the wall appeared to be acting as a horizontal stiffener beam. This was most likely not the intent of this element; however, its presence most likely lessened the observed displacement caused by the high water levels.

Modifications were made to the north wall of the structure. They included the addition of an exterior buttress (Figure 5) and a recommendation to limit the water retention height in the tank to 26.5 feet. The buttress was installed, but there was no indication that level control systems were installed.

At the time of this initial failure and subsequent investigation, it appears that the review and analysis were limited to problems associated with the north wall.

In retrospect, a complete analysis should have been performed based upon the complexity of the problems and the obvious under-design of the structure.

Failure of Basin and Subsequent Investigation

On April 5, 2011, following very heavy rains, the east wall of the basin collapsed in a catastrophic failure resulting in two fatalities. At the time, it was reported Operational and that the basin was filled to a depth of approximately 26 feet. The basin was Maintenance History constructed without overflow structures, Shortly after the basin was placed into serand there was evidence on the outside vice, the north wall began to bow outward Figure 5. Concrete buttress that was installed along the north wall of of the west wall that the basin had overby approximately 4 inches during a very the basin shortly after the basin was placed into service. The wall is flowed before the failure (Figure 6 ). heavy rain period. At that time, the tank cracked in the area in front of the buttress.

Cain and Associates performed an level was reported to be approximately analysis of the original design. They 26 feet. This was well above the normal indicated that the design did not meet operating level of approximately 10 feet. the requirements of ACI 350 but also The resulting lateral displacement caused did not meet the requirements of ACI cracks to form in this wall and the perim318. Additionally, several other deficieneter walkway (Figures 3 and 4). cies were noted from the original design An engineering study was initiated regarding the detailing and construction to determine the cause of the bow and of the basin. design a fix. During this process, it was Finite element analysis was performed determined that the reinforcing steel on the basin, which determined that, design of the north wall did not meet the at full water depth, the north wall reinrequirements of the ACI 318 code and forcing steel was underdesigned by up the more stringent ACI 350 code. It was to 113.35% per ACI 318 and by up to speculated that the ACI 350 code might Figure 6. View shows the structure’s west wall and evidence of 177.35% per ACI 350. Similarly, the not have been used because it would overflow and wastewater to a height of 30 feet 5 inches in the basin. east wall reinforcing was underdesigned have significantly increased the project’s by 127.96% per ACI 318 and by up to overall cost. However, this was never determined to be the case.

196.34% per ACI 350. The authors consider the analysis using a The study did indicate that the deflection and the resulting cracks depth of 30.42 feet reasonably based on evidence that the basin had were likely caused by the ends of the 63-foot-long north wall acting overflowed in the past. Additionally, independent analysis by other J U LY 2022

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professionals determined that the limiting Conclusions depth of water in the basin would have been 21 feet per ACI 318 and 17 feet per ACI 350.

The authors’ opinion is that the structure’s Reinforcing bar couplers were used to confailure was primarily due to three significant nect the baffle walls to the east and west walls factors. Initially, the structure was not designed of the structure (Figure 7). The couplers were to the correct code, which allowed for signifisimilar to a Dayton D50 – #5, product code cant reductions in construction costs, although 77100. Cain and Associates calculated the it is unclear that this was the motivating factor. maximum tensile force in a pair of couplers The correct design code should have been attaching the baffle walls to the east wall ACI 350, and this, when properly followed, to be approximately 43 kips at the height would have provided a much more robust of approximately 9.5 feet above the basin structure than what was constructed. The floor. The load at yield for a pair of threaded second contributing factor was the fact that #5 bars is 27.12 kips based on a yield stress there was insufficient information provided on of 60 ksi and a cross-section area at the cut the design drawings regarding joint locations threads of 0.226 in2. The required ultimate and designs. These are required by code to be design strength should have been no less than shown on the drawings, but the absence of this 70.95 kips based on the calculated load of information forced decisions for joint locations 43 kips and the direct tension factor of 1.65 to be made in the field. This was facilitated by as defined in ACI 350. coordination between the contractor, owner, There was some evidence of corrosion of and engineering representative. The location the couplers, but this did not appear to be a and design of the joints between the baffle significant factor in the failure. Instead, the walls and the exterior perimeter walls were overstress of these members, cyclical loadinferior and significantly contributed to the ing, and resulting fatigue over the structure’s collapse. Lastly, there were issues with the basin Figure 7. Looking up at reinforcing bar couplers placed life were more likely contributing factors to that occurred during the operational period. at the east end of a baffle wall. The couplers are placed the failure (Figures 8 and 9). In addition, at 12-inch vertical and 7-inch horizontal centers.

However, these were not adequately addressed. the placement of the couplers was such that Horizontal displacement and cracking issues the threads could not be fully engaged at some locations due to the associated with the north wall were significant and should have been proximity of the adjacent hooked bar. a harbinger of future problems. Although modifications were made to The north wall was also analyzed from a crack control standpoint. To lessen the detrimental effects occurring in the north wall, it was also control deflections and cracking, both ACI 318 and ACI 350 place a noted at the time that the structure was inadequately designed to not limit on a crack control factor known as the Z factor. These limits are only ACI 350 but also ACI 318. Once this was determined, a more 145 kips/inch and 115 kips/inch, respectively. Analysis of the north thorough structural analysis of the constructed conditions wall support determined that the Z factors ranged between 209 kips/ should have been undertaken, which may have resulted in the inch and 362 kips/inch depending on location and water depth in removal of the basin from service.■ the basin. These fall outside the acceptable range and support the conditions observed during the initial failure of the basin in 1997.

Hal K. Cain is a Principal Engineer with Cain & Associates Engineers & It is more likely that the ultimate collapse of the east wall was caused Constructors, Inc. in Mobile, AL. by the tensile failure of the reinforcing bar couplers. Once the initial Michael A. Amos is a Licensed Professional Engineer in multiple states who failure occurred, a subsequent instantaneous chain reaction failure previously worked with Cain and Associates as a forensic consultant. was caused by the transfer of loads to adjacent couplers.

Figure 8. Reinforcing bar couplers and location where reinforcing failed in direct tension.

52 STRUCTURE magazine

Figure 9. Reinforcing bar coupler that failed in direct tension.

historic STRUCTURES The Schoharie Creek Disaster

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng, P.E., P.L.S.

T

he New York State Thruway bridge across the Schoharie Creek collapsed on April 5, 1987, during a flood, killing 10 people.

Unlike many of the failures discussed previously, this was in the lifetime of many readers.

The Schoharie Creek has its headwater at the foot of Indian Head Mountain in the Catskill Mountains and flows 93 miles in a general northerly direction, meeting the Mohawk River at Fort Hunter. It has a watershed area of approximately 1,000 square miles and, while normally a small creek, it has a history of major flooding. Over the years, the creek was crossed by many bridges washed away by floods and the Erie Canal. A review of the impact of the Creek on the original and the 1842 enlarged Canal illustrates the destructive power of flooding. The original Erie crossed the creek at Ft. Hunter on a pond impounded by a low-level dam in 1822.

The mules would be loaded on the canal boat winched to the other The Schoharie Creek Bridge collapse. side. Later, a small bridge was built for the mules to walk, pulling the canal boat across the creek. Unfortunately, that dam was flooded out The design was approved in January 1952 and was very similar to in 1832. When the enlarged canal was built, the creek was crossed other Thruway bridges, consisting of two steel girders supporting with an aqueduct around 1842. A major flood destroyed another dam cantilevered crossbeams that supported steel stringers on which an built downstream from the Aqueduct in 1864. The Aqueduct was, 8-inch-thick reinforced concrete deck was placed. This gave a total in turn, damaged by significant flooding in 1869, in 1879, and still width of deck of 112 feet 5 inches. The foundations consisted of spread later in 1894. The Gilboa Dam was built in 1927 to create a reservoir footings on top of which was placed a concrete plinth. On top of the to hold water for New York City and provide some flood control plinth were two concrete columns supporting a connecting concrete downstream. It is clear from the historical record that the Creek was beam. On February 11, 1953, the construction contract was awarded a major threat to any structure built over or in its path. to B. Perini and Sons, Inc. (Now Perini Corporation) of Boston.

Moving upstream from its intersection with the Mohawk River It is of interest that the then-current AASHTO standards called for are the remains of the Aqueduct, two steel bridges carrying Route a careful study of local conditions, including flow (discharge) and 5S, followed by the Thruway Bridge. Farther upstream, the famous frequency, the performance of other bridges in the vicinity, and other Blenheim Bridge crossed the creek. It was built in 1855 by Nichols information pertinent to the design of the bridge and likely to affect Powers and was the longest span covered bridge in the country for the safety of the structure. After the collapse, in response to written many years until it was washed away in 2011 by tropical storm Irene. questions from the National Transportation Safety Board (NTSB), It has since been rebuilt. the bridge designer, Pavlo, stated that he did not study the history The New York State (aka Thomas E. Dewey) of Schoharie Creek before preparing the Thruway was built from New York City to final design. Madigan-Hyland Consulting Buffalo in the mid-1950s and is part of the Engineers, who developed the preliminary Interstate Highway System with Numbers plans, design plans, specifications, and I-87 from New York to Albany and I-90 from quantity/cost estimates, conducted a limited Inspectors noticed that Albany to the Pennsylvania State Line. The hydraulic review as indicated by its hydrauthe expansion bearings firms Madigan-Hyland (M-H) and Pavlo lic sheet. However, the sheet did not call were out-of-plumb, roadway Engineers were chosen to design the Schoharie for comments, nor were comments added Creek bridge and many others on the road. concerning the creek’s flood history or the approach slabs had settled, They chose two options as to the length of performance of structures along the creek roadway drainage was poor, the bridge. The first was for a 600-foot-long during previous floods, even though some of and the supporting material bridge and the second 540 feet. The state chose the information was readily available. Given the shorter length that would require filling the well-documented history of flooding of for west embankment dry in some of the creek channel. The final length the Schoharie Creek, this was possibly the stone pavement of spans were 100, 110, 120, 110, and 100 fatal error leading to the bridge’s failure. was deficient. feet. The bridge was designed following the The bridge opened to partial traffic in the 1949 edition of the American Association of summer of 1954 and was fully opened in State Highway Officials (AASHTO) Standard October. Shortly after it opened, cracking Specifications for Highway Bridges. of the plinth was noted and required repair.

54 STRUCTURE magazine

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Almost as a test of the bridge, a 100-year flood of 76,500 cfs roared down the creek in 1955, resulting in no visible damage, but it is thought that the floodwaters may have undercut (scoured) the foundation. “There were several other problems that occurred shortly after the completion of the bridge.

Inspectors noticed that the expansion bear- Plan of the bridge showing piers and span lengths; spans 3 and 4 failed. ings were out-of-plumb, roadway approach slabs had settled, roadway drainage was poor, and the supporting • Berms, built in 1963, directed floodwaters under the bridge. material for west embankment dry stone pavement was deficient.”

• An embankment west of the creek channel increased flood All of the problems mentioned and other minor problems were corvelocities. rected by the fall of 1957.

• The Mohawk River dam downstream was set for winter condiThe spring of 1987 was very wet for upstate New York and New tions and was 3 meters (10 feet) lower than in the 1955 flood, England. Over six inches of rain had fallen on April 3rd and 4th, and, increasing the hydraulic gradient. coupled with snowmelt, water came roaring down the creek on April 5 The NTSB, which was very critical of state employees and their in what was determined to be a 50-year flood. Pier No. 3 was washed consultants, reported in part, out around 10:45 AM, and spans 3 and 4 dropped into the creek. With Correspondence between M-H and the DPW [Department of debris in the channel, the flow was directed to Pier No. 2, which also Public Works] relating to hydraulics usually addressed the length was washed out, and span 2 dropped into the creek. At the time of of the bridge and the elevation of the backwater, but not the the collapse, there was one car and a tractor-trailer on the bridge, and frequency and magnitude of previous floods or their effects on before traffic could be stopped, three more cars ended up in the creek. other structures over the Schoharie Creek. [they didn’t] menA total of 10 people lost their lives in the disaster. The body of the tenth tion [any] of the three floods that exceeded 50,000 cubic feet victim was found two years later downstream in the Mohawk River. per second (cfs), which occurred during the first half of the 20th As is always the case, people and politicians want to know what caused century, let alone an analysis of their importance to the design the failure and loss of life. The Thruway Authority hired the firms of and construction of the bridge.

Wiss, Janney, Elstner Associated of Northbrook, Illinois, and Mueser [Madigan- Hyland’s] failure to review the available history limRutledge Consulting Engineers of New York City to investigate. In ited its appreciation for the potential for scour at this bridge site. addition, the NTSB conducted its own study of the failure. A fourth If M-H had visited some of the other structures along the creek, investigation was undertaken by Thornton-Tomasetti, PC, of New such as the Aqueduct 3,000 feet north of the bridge, it probYork for the New York State Disaster Preparedness Commission. ably could have observed scour near the piers, and this may have They all agreed that scour undermined Piers No. 2 and 3, leading heightened its concern for scour. to the disaster as a hole approximately 9 feet deep and 25 to 30 feet The NTSB then went on for five paragraphs criticizing the elevation long had formed, and the spread footing simply slid into the hole. of the base of the spread footings as being too high based on boring Some conclusions were, data. They then discussed the scour situation again and the use of • The shallow footings used, bearing on soil, could have been piles as a means of minimizing scour effects, writing, undermined. Therefore the depth of the footings were not Based on this collapse, as well as on an improved understanding enough to take them below the probable limit of scour. of hydraulics and an improved ability to predict scour, the Safety • The foundation of Pier 3 was bearing on erodable soil. Layers of gravel, sand, and silt, interbedded with folded and tilted till, Structural design prowess allowed high-velocity floodwaters to penetrate the bearing stratum. meets architectural vision.

Riprap protection, inspection, and maintenance were inadequate.

In addition, it was found that the riprap stone was smaller than required and that sheet piling around each pier, intended to be left in place, had been removed.

Thornton-Tomasetti also found, • The flood was greater than that anticipated by the designers and followed the 1955 flood and others that had disturbed the riprap.

• A curve in the river upstream of the bridge directed a highervelocity flow toward Pier 3.

• Drift material, caught against the piers, directed water downward at Seattle | Tacoma | Portland | pcs-structural.com the base of Pier 3. © Jeff Amram

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Bridge with three spans out.

Board believes that AASHTO should modify its requirements for the depth of the footings… and require that the depth be based on estimates of the maximum potential depth of scour at the bridge site, rather than on the existing streambed elevation.

Scour piles may have provided the stability needed for the substructure to withstand the scour of the 1987 flood since the maximum scour depth observed at the bridge site was 9 feet below the bottom of the footing of pier 3…However, since the Safety Board cannot be certain how deep the scour hole at pier 3 may have become had spans 3 and 4 not fallen when they did, it is not possible to conclude that piles driven in accordance with the AASHTO recommendations for pile depths would have prevented the bridge collapse. Certainly, if piles were driven deeply enough, the piers would not have lost their support. Therefore, the Safety Board concludes that had the Schoharie Creek Bridge been designed with piles to protect against scour, the collapse might not have occurred, depending on how deeply the piles were driven below the footings…

They concluded in this section, “Thus, the Safety Board believes that section 4.3.1.2 should be modified to require that the depth of piles exceed the predicted maximum potential depth of scour.” They then got into the importance of riprap and the fact that all inspections downplayed the importance of riprap and concluded that state inspectors, and their superiors, did not seem to know whether the footings were on piles or not.

When the Schoharie Creek Bridge was designed and built, riprap was a recognized means of protecting scour, and riprap was specified in the contract. Without piles, the integrity of the bridge foundation depended entirely on the maintenance of riprap for protection against scour.

The NTSB concluded in this section “that had the piers been protected by riprap at the time of the April 1987 flood as they were during the 1955 flood, the bridge probably would not have collapsed.”

They then discussed the inspection and criticized everyone who was or should have been involved with the in-depth inspection. Some of their statements are as follows. (The online version of this article contains expanded comments from the NTSB report.)

The inspections in the New York State Thruway Authority (NYSTA) Albany division were accomplished not by engineers but by personnel whose primary responsibilities were in bridge maintenance. The Albany assistant division engineer (bridges) was not a professional engineer...

However, in his 1986 inspection of the bridge and previous inspections, the Albany assistant division engineer (bridges) failed 56 STRUCTURE magazine

to evaluate the condition of the riprap at the piers properly, and he failed to take the dropline readings necessary to evaluate the conditions in the streambed…The fact that he overlooked these two tasks indicated that he either did not think they were important or did not understand their importance. In addition, the engineer’s supervisors, who should have reviewed his reports, apparently did not review his reports or failed to recognize the seriousness of the omissions and therefore did not attempt to correct the situation.

In 1979, an engineering firm conducted bridge inspections for the New York State Department of Transportation (NYSDOT) to comply with the National Bridge Inspection Standards (NBIS) inventory requirements for off-system bridges...The measurements and photographs from the inspection clearly indicated that riprap was not piled at an even level around the plinth.

This information should have alerted a person knowledgeable in river mechanics and structures that riprap had moved, posing a danger to the structure…

The Safety Board believes that the sketches showed that a significant amount of riprap had moved away from the upstream ends of the piers in 1979 and, especially since there were no piles, the engineering firm should have, in accordance with its agreement with the NYSDOT, immediately called the NYSDOT project manager to alert him…

The Safety Board is concerned that bridges similar to the Schoharie Creek Bridge may not be receiving proper riprap maintenance because there is no proper guidance as to when to replace riprap…The Safety Board is thus convinced that specific guidance must be provided to bridge inspectors.

The Safety Board believes that research is needed to determine the size and amount of riprap needed for scour protection and the degree of depletion that may occur before replacement is necessary…

In conclusion, the failure of the Schoharie Creek Bridge can be traced to bad design, bad construction, bad inspection, lax supervision, etc. However, it did have a positive effect in instituting periodic underwater investigations to determine the extent, if any, of bridge scour and bringing a greater awareness of the potential problem to the designers of bridges over rivers, streams, and creeks. In February 1991, the FHWA issued Hydraulic Engineering Circular, HEC, No. 18, Evaluating Scour at Bridges. Ten lawsuits against the Thruway Authority were filed from relatives of those who died in the collapse. They were settled for a total of approximately $4,000,000.

The Authority then sued the bridge contractor and designer, but the statute of limitations had expired on the contractor. They later settled with the designers for $600,000.

This is the last in the series on bridge failures. Recent failures such as the Minneapolis I-35, Pittsburg (Frick Park), and Florida International University pedestrian bridges are still being studied and debated. A new series on 19th Century bridges across the Mississippi River will be published in future issues.■ Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He is now an Independent Consulting Engineer (fgriggsjr@twc.com).

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SPOTLIGHT Meeting Seismic Demands

Successful Use of the Latest Advances in the Field

S

tanford School of Medicine’s new Center for Academic Medicine (CAM) provides consolidated workspaces for clinical faculty, computational researchers, departmental administration, and leadership for several departments. The new four-story 170,000-square-foot building is built above a three-level subgrade parking structure, concealing vehicles and eliminating the heat islands of surface parking. It is an innovative workplace that includes collaborative conferencing facilities, quiet office space, a host of amenities, and ample access to natural light and views. Situated adjacent to Stanford’s historic arboretum, CAM is a new threshold to the campus, the School of Medicine, and an extension of the arboretum experience.

The U-shape provides building occupants direct access to daylight, views, and the arboretum itself. It also maximizes the benefits of passive solar and ventilation design on the site.

To make this configuration a possibility under high seismic demands, a FEMA P-58 simulation of steel special moment resisting frames versus buckling restrained braced frames was performed to inform a structural system decision that considered first costs, strength, drift ratios, and annualized losses. These studies resulted in the selection of buckling restrained braced frames for the constrained layout controlled by the cantilevering U-shaped plan.

The seismic force-resisting system was also designed to meet Stanford’s seismic guidelines with performance objectives beyond basic Code requirements. Turning a challenge into a design strategy, the braced frames were located within the building to delineate workspaces and define the boundaries of amenity spaces, integrating elements with the program that may otherwise have been obstacles.

Parts of the gravity system have several elements intended to blur the lines between indoors and nature outdoors. These elements include a feature cantilevered multi-story volume over the ground floor lobby, two pedestrian bridges stretching between the narrow parallel wings of the building, an exterior corridor, and cantilevered terraces at the office wing extremities. One bridge at Level 2 is suspended from the structure above using inverted moment frames, providing lateral stability without the need for braces interrupting the horizontal lines of the design language. The other bridge is supported atop STRUCTURE magazine

HOK was an Outstanding Award Winner for the Stanford Center for Academic Medicine Project in the 2021 Annual NCSEA Excellence in Structural Engineering Awards Program in the Category – New Buildings $80M to $200M.

moment frames from below, traversing the open end of the U-plan and overlooking the landscaped plaza. The 2-story tall columns that support the bridge were designed with a yielding shear link at mid-height that ensures the elastic behavior of the pairs of columns.

To further accentuate the indoor-outdoor experience, the exterior corridor is hung from the roof to allow for a column-free area at the courtyard. The structural engineering for this project demonstrates excellence in integrating high-performing, innovative structural systems within the overall theme of the building.

The multi-story volume and terrace at Level 3, which appear to float over the ground floor lobby, are suspended from four hanging columns connected to four cantilevered steel trusses at the roof. This permits the glazed, cube-shaped ground floor lobby to be relatively free of structural elements, making it appear more transparent and light. To achieve this, the cantilever-supporting roof trusses are elevated from the nominal roof elevation such that their bottom chord is in the same plane as the typical roof framing. Given the critical nature of the trusses in resisting gravity loads, the trusses, their components, and their connections were designed to remain elastic at a Maximum Considered Earthquake level of force. A parametric optimization analysis was performed to optimize the location of the cantilevered truss diagonals to minimize tonnage.

Additionally, to assist in the erection of the roof trusses, the engineering team developed

a proposed erection sequence with anticipated deflections at significant construction milestones. Rather than fabricating a camber directly into each member of the four roof trusses, the support columns closest to the free end of the roof trusses were fabricated approximately 1 inch taller than their nominal height. This raised the tips of the roof trusses upwards, permitting them to deflect to the intended elevation once superimposed loads were applied. This simplification assisted in coordinating changes to the sequence proposed by the erector. Surveyed deflections from the field met the simulated prediction, resulting in the ease of installing the finish materials and cladding.

The structural challenges of a U-shaped plan flanked on all sides by cantilevers and interconnected by slender pedestrian connectors were met with strategies that used the latest advances in the field to meet seismic demands that well surpassed code-required performance. Parametric design was utilized to reduce gravity system tonnage beyond that possible with conventional structural workflows. The FEMA P-58 simulation allowed quantification of the ideal lateral system strategy for multiple project objectives, and the resulting elements helped delineate zones in the programmatic layout. The result is a structure that presents the best of passive environmental design strategies, biophilic elements, and a resilient structure that blends seamlessly into the design language and landscape.■ J U LY 2022

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Leadership Retreat Success!

With nearly 50 attendees representing 33 different Structural Engineers Associations (SEAs), 16 individuals representing all the national committees, and full in-person support of the Board of Directors, NCSEA successfully completed its first-ever SEA Leadership Retreat. Held virtually in 2020 and 2021, the time was right to hold a face-to-face meeting allowing what an in-person event does best – relieve isolation while encouraging collaboration.

The Retreat, held over two days, was focused on two objectives: strengthen the national committees’ relationships with local SEA committees and help all NCSEA volunteers increase their skill set regarding volunteer leadership. Retreat presentations included: Improving Your SEAs Committees Effective Meeting Planning & Programming Tips & Tricks to Effective Association Operations A special highlight was a presentation and workshop by Amanda Kaiser, a member engagement specialist from Kaiser Insights LLC. With live input from the Retreat attendees, Amanda created a Membership Engagement Handbook specifically for the SEAs. Also, an update of the We SEE Above and Beyond brand and marketing campaign was shared.

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SEI Update Learning / Networking

The 2022 SEI Standards Series Review ASCE 7-22, changes from ASCE 7-16, the Digital Products/Hazard Tool, and join the discussion with the expert standard developers. 1.5 PDHs per session.

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New from ASCE Library The Engineer's Project Delivery Method Primer Uniform Definitions and Case Studies Douglas D. Gransberg, Ph.D., P.E.; Michael C. Loulakis, Esq.; and Ghada M. Gad, Ph.D. Focuses on the design-bid-build project delivery method and variations for contracting. www.ascelibrary.org

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ASCE Infographics

View ASCE Infographics and resources providing practicing civil engineers with guidance on advancing the UN Sustainable Development Goals as well as the Infrastructure Investment and Jobs Act (IIJA). Topics include: • Infrastructure Resilience • Bridge Asset Management • Dams and Levees • Structural Health Monitoring ascelibrary.org/infographics

News of the Structural Engineering Institute of ASCE Students and Young Professionals

Congratulations to SEI Futures Fund Student Scholarship Recipients Award is to Attend the Electrical Transmission and Substation Structures Conference Adetona Adediran, S.M.ASCE, University of South Alabama Kehinda Joseph Alawode, A.M.ASCE, Florida International University Ahmed Ali, S.M.ASCE, University of Cincinnati Ariana Cabral Felix, S.M.ASCE, University of Cincinnati Xinlong Du, S.M.ASCE, Northeastern University Laura Gray, S.M.ASCE, Saint Louis University William Hughes, S.M.ASCE, University of Connecticut Miaomiao Li, S.M.ASCE, Michigan Technological University Sarvadaman Pachade, S.M.ASCE, University of Western Ontario Babak Salarieh, Ph.D., A.M.ASCE, University of Alabama in Huntsville Ashray Saxena, S.M.ASCE, University of Texas at Austin Ahmed Shahin, S.M.ASCE, American University of Sharjah Ali Shokrgozar, S.M.ASCE, Idaho State University Xinyue Wang, S.M.ASCE, Lehigh University Andualem Yadeta, C.Eng, S.M.ASCE, Delhi Technological University

Advancing the Profession

Congratulations to:

Recent SEI Fellows recognized at Structures Congress.

Learn more and apply at www.asce.org/SEIMembership.

SEI members recognized as 2022 ASCE Distinguished Members: • • • •

James L. Beck, Ph.D., F.EMI, Dist.M.ASCE Michel Bruneau, Ph.D., P.Eng, F.SEI, Dist.M.ASCE Halil Ceylan, Ph.D., C.Eng, Dist.M.ASCE Therese P. McAllister, Ph.D., P.E., F.SEI, Dist.M.ASCE See full list at www.asce.org/career-growth/awards-and-honors/distinguished-members.

futureworldvision.org

Errata

SEI Standards Supplements and Errata including ASCE 7. See www.asce.org/SEI. To submit errata, contact sei@asce.org.

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CASE in Point Events in Structural Engineering ACEC Coalitions Summer Meeting August 3-4, 2022, Salt Lake City, Utah The Summer Meeting offers the opportunity to connect with peers in your field, gain education, and participate in the executive committee meetings that shape CASE’s participation.

This meeting features a reception showcasing an overview of the seismic upgrade and renovation of the Salt Lake Temple. The meeting is open to all CASE members.

The 2022 Meeting includes: • CASE Roundtable addressing trending industry topics.

• In-depth presentation on the Salt Lake Temple seismic upgrade and construction project, which includes new additions, revamped grounds, seismic upgrades and steel additions, along with an updated mechanical and electrical system. The presentation will give an overview of the project and its purpose, and a review of the base isolation system.

It will discuss the unique design and construction efforts for this extraordinary project.

Learning Objectives: 1) Learn how base isolation systems protect buildings.

2) Learn ways to minimize building movement during load transfer.

3) Learn the importance of proper shoring to meet project needs.

This project will utilize massive concrete footings and transfer girders to support the temple on the base isolators. The presentation will discuss the measures taken to meet the ACI code requirements for mass concrete, bending, shear, and torsion.

• Education sessions will focus on Green Energy trends and how engineers can identify projects and opportunities for their firms.

Register Now! Visit www.acec.org and click on “Events” then “Coalitions.”

ACEC Fall Conference October 16 -19, 2022, Colorado Springs, Colorado Each year, ACEC sponsors two major national meetings: the Annual Convention and the Fall Conference. National meetings provide attendees an opportunity to obtain information about issues that affect the industry through informative education, networking, and exhibits.

The Coalition of Structural Engineers (CASE) hosts a roundtable at each meeting to discuss issues specific to the field of structural engineering. Hope to see you there!

Announcing 2022 CASE Scholarship Winner Congratulations to the 2022 CASE Scholarship winner, Taylor Drahota from the University of Nebraska-Lincoln! View information about the scholarship and winners at acec.org under “Awards.”

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News of the Coalition of American Structural Engineers CASE Tools and Resources Did You Know?

CASE has tools and practice guidelines to help firms deal with a wide variety of business scenarios that structural engineering firms face daily. Whether your firm needs to establish a new Quality Assurance Program, update its risk management program, keep track of the skills engineers are learning at each level of experience, or need a sample contract document – CASE has the tools you need!

Check out some of the CASE Contract Documents developed by the Contracts Committee…

CASE #1 – An Agreement for the Provision of Limited Professional Services.

This is a sample agreement for small projects or investigations of limited scope and time duration. It contains the essentials of a good agreement including scope of services, fee arrangement, and terms and conditions.

CASE #2 – An Agreement Between Client and Structural Engineer of Record for Professional Services.

This agreement form may be used when the client, e.g., owner, contractor developer, etc., wishes to directly retain the Structural Engineer of Record. The contract contains an easy-to-understand matrix of services that will simplify the “what’s included and what’s not” questions in negotiations with a prospective client. This agreement may also be used with a client who is an architect when the architect-owner agreement is not an AIA agreement.

CASE #9 – An Agreement Between Structural Engineer of Record and Consulting Design Professional for Service.

The Structural Engineer of Record, when serving in the role of Prime Design Professional or as a Consultant, may find it necessary to retain the services of a sub-consultant or architect. This agreement provides a form that outlines the services and requirements in a matrix so that the services of the sub-consultant may be readily defined and understood.

You can purchase these and other Risk Management Tools at www.acec.org/bookstore.

If you are a member of CASE, this tool and all publications are free to you.

NCSEA and SEI members receive a discount on publications. Use discount code – NCSEASEI2022 when you check out.

Get to Know the CASE Committees The CASE Coalition has several committees that meet regularly to develop documents that guide engineers in their business practice. The work of these committees is an important part of what coalitions do and are one of the biggest values of CASE membership. The Contracts Committee is responsible for developing and maintaining contracts to assist practicing engineers with risk management.

This month, the Contracts Committee is preparing for the Summer Coalitions Meeting in Salt Lake City, UT, August 3rd and 4th, and working with ACEC National on outreach.

We are currently seeking two to four new members to join the Contracts Committee. Do you know someone in your firm that is looking for ways to expand and strengthen their business skillset, gain experience serving on a committee, sharpen their leadership skills, and travel to interesting places? Please consider applying for a position on the committee. Committee member commitments include a monthly virtual meeting, a few hours a month working on relevant documents, and travel to the Coalitions winter and summer meetings!

To apply, your firm should: • Be a current member of ACEC • Be a member of the Coalition of American Structural Engineers (CASE); or be willing to join the Coalition • Be able to attend the groups’ regular face-to-face meetings each year: August, February (hotel, travel partially reimbursable) • Be available to engage with the committees via email and video/conference call • Have some specific experience and/or expertise to contribute to the group J U LY 2022

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SPOTLIGHT SR99 Tunnel

Meeting Stringent Seismic and Highway Design Criteria

T

he Washington State Department of to match the previous segment. Refinements during and post-earthquake events. The cyclic Transportation State Route 99 tunnel is in the detailing of the segments, the size and testing requirement was specially developed North America’s largest-diameter bored tunnel. orientation of bolt pockets, gasket placement, to address the ‘memory effect’ concern of With a stacked roadway design, the single- and grouting port locations made for a repeti- elastomeric gasket material. bore tunnel sets a new standard for The tunnel’s seismic pertunnel and infrastructure design, formance is enhanced by especially in dense urban areas where custom joint design to allow surface disturbance and impacts on relative movement between communities, including residences the tunnel liner and interior and businesses, are at a premium. structures. The unique tunnel The completed tunnel created a interior structure design safe north-south route bypassing included a series of 650-foottraffic congestion in Seattle’s downlong moment frame systems town core and improved mobility of walls and slabs that rest between neighborhoods. In addition, on continuous corbels, can it allowed the removal of a seismically expand and contract lonvulnerable viaduct that reduced trafgitudinally independent of fic noise and opened access to the the tunnel rings, and accomSeattle waterfront. modate relative transverse The General Contractor, STP, deformation between the inteHNTB Corporation was an Outstanding Award Winner for the State selected a 57.5-foot tunnel boring rior structure and tunnel ring.

Route 99 Alaskan Way Viaduct Replacement Program Project in the 2021 machine (TBM) and the tunnel This project’s tunnel lining Annual NCSEA Excellence in Structural Engineering Awards Program in the configuration, which enhanced and underground structures Category – New Buildings or Transportation Structures. vehicular safety and traffic operawere designed to achieve a tions by providing a wide eight-foot shoulder tive factory line-style installation of the rings. 100-year design service life. The concrete and two 11-foot travel lanes for a 32-foot wide When erected behind the TBM, the rings mix was designed to deliver dense and relaroadway plus a 15.5-foot vertical clearance for formed a permanent watertight structure that tively impervious concrete to overcome the both upper and lower roadways. supported roadways. Behind the completed effects of moderately aggressive soils, the use The tunnel travels beneath more than 150 tunnel rings, STP used an innovative system to of chlorides in roadway deicing chemical, and buildings in downtown Seattle, under utili- efficiently construct the interior roadway struc- carbon dioxide and moisture of the tunnel ties and other structures requiring protection tures. HNTB’s unique structural design fully operating environment (which lowers the during construction. To implement a tunnel- accommodated constructability allowing STP pH level in concrete and accelerates the rate ing system that provided ground support and to achieve continuous construction progress. of reinforcement corrosion). To this effect, advanced the TBM in its planned alignment, The large double-deck tunnel required spe- low water/cementitious material ratio and structural engineers at HNTB designed a cial seismic design to sustain minimal damage, non-reactive aggregates were used; Portland single-pass segmental liner system. The remain operational during moderate earth- cement was partially replaced with fly ash combination of the proper operation of the quakes (108-year return period), and prevent and micro silica. Rebar covers were deterTBM, placement of the precast rings, and collapse by limiting inelastic deformation to mined through a comprehensive concrete subsequent grouting efforts to fill voids as acceptable levels should a rare earthquake durability study and models calibrated for the tunnel advanced controlled the settlement (2,500-year return period) occur. A unique 100-year design life. A waterproofing system below the design threshold values resulting in challenge was the water tightness of the tunnel encapsulated the cut-and-cover structure and no damage to existing facilities. lining’s radial and circumferential joints under expansion joints of approach structures. All The use of precast concrete segmental liner, seismic events. Joint openings caused by interior tunnel superstructures, joints, and coupled with innovations in its detailing and tunnel deformation during a critical seismic bearings are fully accessible for inspection and construction, allowed the team to meet the event and static loadings were magnified by maintenance and detailed for future periodic design criteria, build to specified construction required construction tolerances for gasket replacement. tolerances and facilitate rapid tunnel advance. design; to ensure gasket satisfactory perforThe SR 99 Tunnel is designed as one of A universal ring made up of 10 tapered seg- mance ‘staggering effect’ of the adjacent rings the safest tunnels worldwide. The structural ments was used; its placement pattern allowed was utilized by means of shear cones and bolts. design meets both stringent seismic and adjustments at the TBM heading to keep the SR 99 also was the first project in the world highway design criteria. As a result, tunnel tunnel within the desired tolerances. Ring to require cyclic testing of the tunnel lining systems will function during and after a design segments were installed inside the TBM gasket to confirm the gasket material is resil- earthquake to control a fire, evacuate shield with vacuum erectors and bolted up ient under repetitive cyclic loading and that smoke, allow people to exit safely, by operators with compressed air hand tools gasket water tightness is never compromised and maintain structural integrity.■ 66 STRUCTURE magazine

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