STRUCTURE magazine June 2019

STRUCTURE JUNE 2019

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LeMessurier Calls on Tekla Structural Designer for Complex Projects Interoperability and Time Saving Tools

Tekla Structural Designer was developed specifically to maximize collaboration with other project parties, including technicians, fabricators and architects. Its unique functionality enables engineers to integrate the physical design model seamlessly with Tekla Structures or Autodesk Revit, and to round-trip without compromising vital design data.

“We’re able to import geometry from Revit, design in Tekla Structural Designer and export that information for import back into Revit. If an architect makes geometry updates or changes a slab edge, we’ll send those changes back into Tekla Structural Designer, rerun the analysis and design, and push updated design information back into Revit.”

Tekla Structural Design at Work: The Hub on Causeway

For over 55 years, LeMessurier has provided structural engineering services to architects, owners, contractors, developers and artists. Led by the example of legendary structural engineer and founder William LeMessurier, LeMessurier provides the expertise for some of the world’s most elegant and sophisticated designs while remaining true to the enduring laws of science and engineering.

Known for pushing the envelope of the latest technologies and even inventing new ones, LeMessurier engineers solutions responsive to their clients’ visions and reflective of their experience.

An early adopter of technology to improve their designs and workflow, LeMessurier put its own talent to work in the eighties to develop a software solution that did not exist commercially at the time. Their early application adopted the concept of Building Information Modeling (BIM) long before it emerged decades later.

While LeMessurier’s proprietary tool had evolved over three decades into a powerhouse of capability, the decision to evaluate commercial structural design tools was predicated on the looming effort required to modernize its software to leverage emerging platforms, support normalized data structure integration and keep up with code changes.

After a lengthy and thorough comparison of commercial tools that would “fill the shoes” and stack up to the company’s proprietary tool, LeMessurier chose Tekla Structural Designer for its rich capabilities that addressed all of their workflow needs. According to Derek Barnes, Associate at LeMessurier, ” Tekla Structural Designer offered the most features and the best integration of all the products we tested. They also offered us the ability to work closely with their development group to ensure we were getting the most out of the software.”

One Model for Structural Analysis & Design

From Schematic Design through Construction Documents, Tekla Structural Designer allows LeMessurier engineers to work from one single model for structural analysis and design, improving efficiency, workflow, and ultimately saving time. “Our engineers are working more efficiently because they don’t need to switch between multiple software packages for concrete and steel design. Tekla Structural Designer offers better integration of multiple materials than we have seen in any other product,” said Barnes.

LeMessurier engineers use Tekla Structural Designer to create physical, information-rich models that contain the intelligence they need to automate the design of significant portions of their structures and efficiently manage project changes.

TRANSFORMING THE WAY THE WORLD WORKS

“Tekla Structural Designer has streamlined our design process,” said Craig Blanchet, P.E., Vice President of LeMessurier. “Because some of our engineers are no longer doubling as software developers, it allows us to focus their talents on leveraging the features of the software to our advantage. Had we not chosen to adopt Tekla Structural Designer, we would have needed to bring on new staff to update and maintain our in-house software. So Tekla Structural Designer is not just saving us time on projects, it is also saving us overhead.

Efficient, Accurate Loading and Analysis

Tekla Structural Designer automatically generates an underlying and highly sophisticated analytical model from the physical model, allowing LeMessurier engineers to focus more on design than on analytical model management. Regardless of a model’s size or complexity, Tekla Structural Designer’s analytical engine accurately computes forces and displacements for use in design and the assessment of building performance.

“Tekla Structural Designer offers better integration of multiple materials than we have seen in any other product.”

Positioning a large scale mixed-use development next to an active arena, a below grade parking garage, and an interstate highway, and bridging it over two active subway tunnels makes planning, phasing and engineering paramount. Currently under construction, The Hub on Causeway Project will be the final piece in the puzzle that is the site of the original Boston Garden.

Despite being new to the software, LeMessurier decided to use Tekla Structural Designer for significant portions of the project. “Relying on a new program for such a big project was obviously a risk for us, but with the potential for time savings and other efficiencies, we jumped right in with Tekla Structural Designer. It forced us to get familiar the software very quickly.”

“Tekla Structural Designer allowed us to design the bulk of Phase 1 in a single model,” said Barnes. The project incorporates both concrete flat slabs and composite concrete and steel floor framing. “Tekla Structural Designer has the ability to calculate effective widths based on the physical model which is a big time saver,” said Barnes. “On this project, the integration with Revit, along with the composite steel design features enabled us to work more efficiently. Adding the ability to do concrete design in the same model was a bonus because we had both construction types in the same building.”

“Tekla Structural Designer helped this project run more efficiently, and in the end it was a positive experience,” said Blanchet.

“Tekla Structural Designer gives us multiple analysis sets to pull from, which gives us lots of control. Most programs don’t have the capability to do FE and grillage chase-down. For the design of beam supported concrete slabs, Tekla Structural Designer allows us to separate the slab stiffness from the beam stiffness, so if we choose to we can design the beams without considering the influence of the slab. In the same model we can use a separate analysis set to review the floor system with the beams and slab engaged,” said Barnes.

Barnes also shared similar benefits with concrete column design. “Tekla Structural Designer does grillage take-downs floor-by-floor, finds the reactions and applies them to the next floor. This allows us to view column results both for the 3-dimensional effects of the structure as a whole and from the more traditional floor-by-floor load take-down point of view.

Doing both has always required significant manual intervention, but Tekla Structural Designer puts it all in one place.” “We reduce the possibility for human error because with Tekla Structural Designer less user input is required,” said Barnes. “Tekla Structural Designer automatically computes many of the design parameters, such as column unbraced lengths. The assumptions made by the software are typically correct, but we can easily review and override them when necessary.”

“Tekla Structural Designer provided the best fit for our workflow compared to other commercially available software.”

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Contents

Cover Feature

JUN E 2019

34 SAN DIEGO CENTRAL COURTHOUSE

By Mark P. Sarkisian, S.E., Peter L. Lee, S.E., and Rupa Garai, S.E.

The structural team for the new San Diego Central Courthouse was tasked with addressing enhanced seismic performance objectives.

The solution included special moment frames with reduced beam section qualified connections.

30 INNOVATIVE CONCRETE CORE FOR TALL TOWERS

38 PIER 70 BUILDING 113/114 RENOVATION

By Joe Ferzli, P.E., S.E., and Jason Thome, P.E., S.E.

By Michael Gemmill, S.E., and Anthony Giammona, S.E.

The Kiara project, a 41-story luxury apartment tower over a seven-

A dilapidated former shipbuilding complex from the late 19th century

level podium with six levels of below-grade parking, posed several

has been revitalized as part of a massive adaptive reuse project. The

challenges. Concrete core wall geometry and steel fiber-reinforced

project team aimed to develop a seismic retrofit that would stabilize

concrete coupling beams were vital to deliver an efficient ductile core.

the perimeter URM walls, and provide enhanced seismic performance.

Columns and Departments 7

Editorial Reforming Structural Engineering Education

24

By Chris Letchford, D.Phil, CPEng

Structural Rehabilitation

Snow Thermal Factors for Structural Renovations

50

By John S. Lawler, Ph.D., P.E., and Elizabeth I. Nadelman, Ph.D.

By Michael O’Rourke, Ph.D., P.E.,

8

12

and Scott Russell, S.E., P.E.

Building Blocks Drag Trusses

By David A. Fusco, P.E., and Judian Duran, E.I.

28

Structural Design Post-Tensioning Design 42

Construction Issues

Recommended Details for Reinforced Concrete Construction – Part 1

Lessons Learned Concrete

Foundation Walls Subjected to Lateral Soil Loads

By George A. Merlo, P.E., and Anthony C. Merlo, P.E.

Professional Issues Sharing Claims Experience for Better Structural Engineering

Structural Sustainability

By John G. Tawresey, S.E.

Community Resilience through Mandatory Retrofit Ordinances

53

46

48

Code Updates Tall Mass

By Erica Fischer, P.E., Keith Porter, P.E., Ph.D.,

Timber Construction Types Included in 2021 IBC

and Elaina J. Sutley, Ph.D.

By Kenneth Bland, P.E.

Historic Structures Rider Bridge Failure, New York & Erie Railroad 1850

Seismic Design and Hazard Maps: Before and After

and Michael Mota, Ph.D., P.E., SECB

By David A. Fanella, Ph.D., S.E., P.E.,

20

Northridge – 25 Years Later

By Frank Griggs, Jr., D.Eng., P.E.

By Nicolas Luco, Ph.D.

By Bijan O. Aalami, S.E., Ph.D., C.Eng.

16

InSights Non-Proprietary Ultra-High-Performance Concrete

59

Spotlight Wiikiaami

By Ahraaz Qureishi, P.E.

66

Business Practices From Peer to Manager

By Jennifer Anderson

In Every Issue 4 57 60 62 64

Advertiser Index Resource Guide – Tall Buildings NCSEA News SEI Update CASE in Point

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, the Publisher, or the Editorial Board, Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

J U N E 2 019

5

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EDITORIAL Reforming Structural Engineering Education

By Chris Letchford, D.Phil(Oxf), F.IEAust, F.SEI, F.ASCE, CPEng RPEQ

T

he Committee for the Reform of Structural Engineering Education (CROSEE) was instigated by the SEI Board of Governors to review, reimagine, and reignite Structural Engineering Education for the 21st Century, or at least the next 20 years!

The committee of practitioners and academics has deliberated on the nature and future of structural engineering, particularly as we see the passionate work done by researchers and standards writers to codify the practice of the profession almost to extinction in the era of AI/ML (Artificial Intelligence/ Machine Learning).

While performance-based code approaches are opening the door to the real value of structural engineering, what is clear is that the profession, like many others, must be seen as a 4 (or 5) + 40-year continuum. Clearly what is taught (much), what is learned (less), and what is practical (?) should, therefore, inspire lifelong learning and that curiosity must be cultivated and developed in the most flexible minds – those of the young, and young-at-heart. For curiosity is the spring of eternal (lifelong) learning, and the source of inspiration and innovation, two characteristics seen publicly as more the domain of the architect rather than the structural engineer in the current context.

It is without a doubt that structural engineers bear substantial responsibility for life safety and property preservation, and that our first-world expectations lead to safe and often conservative designs. Public expectations for such ‘reliability,’ low risk, and high professional regard, are welcome but equally challenge the ability to innovate and perhaps recruit the most talented to the profession. Interestingly, computers and computational – modeling, analysis, fabrication, and construction management tools – now allow architects to create amazing structures, albeit with the seemingly silent partnership of the structural engineer.

Better communication of the crucial role of the structural engineer in these creations will help maintain and grow the profession.

Examining the historical education of the structural engineer, we see a progression from the apprenticeship model of the master builder, wholly within the profession, to the current version in which the academy and the profession are almost wholly separated by the student body with limited overlap. While the academy rightly focusses on instilling fundamental skills of material performance, structural behavior, loading, modeling, and analysis, and the professional skills in students in design and project management, it is clear that there needs to be a blurring of the seemingly distinct boundaries seen by most parties in this education process. Academics often believe that this is the first and maybe last chance to instill a fundamental understanding of structural behavior, while for students it is often hard to see the relevance of material studied, trapped in the semester cocoon of homework STRUCTURE magazine

and quiz cycle. Professionals are apt to focus on the business end of the profession and the importance of application. Boundary blurring is increasingly achieved, at least in one direction – profession to academe, via adjunct professors (full or part-time) who bring much needed ‘real world’ examples to the students, often wallowing in analytical abstraction. One question CROSEE has been examining, and which was the subject of a panel discussion at SEI Congress in Orlando this year, is the training required, desired, or perhaps mandated for adjuncts. What motivates professional experts in the field to ‘profess’ and teach the next generation of engineers?

What impediments or inducements arise or obstruct? How best to convey this knowledge in the more formal classroom setting to a new generation of students driven by and interested in technology?

On the other hand, as the academy focusses more on scientific rather than engineering endeavors, often driven by research funding availability, the challenge is how to make the formal engineering education better relate to engineering practice and also how to engage students in the wonders of modern engineering. This is particularly the case when many university faculty no longer ‘practice’ engineering, hold licensure or have even worked in the industry, having transitioned from one education arena to another. Perhaps academics need to be challenged with sabbaticals in industry to develop relationships that lead to collaborations and research opportunities on real problems of interest to the profession, and hopefully to the betterment of society as a whole rather than the individual. Indeed, civil engineering, of which structural engineering is a significant component, is about creating common wealth, the infrastructure that supports our integrated, complex, urban existence.

By promoting a porous boundary between the profession and the academy, one that sees experienced adjuncts teaching and junior faculty embedding in Industry, it is hard to see how students would not benefit as those ‘in the middle.’ Additionally, the vigor of modern engineering practice, challenged by the amazing structures now conceived, can only help attract students to our profession.

Please follow CROSEE here and at future SEI Structures Congresses as we continue to promote structural engineering approaches and initiatives to fulfill the Board of Governors charge. Join the discussion at https://bit.ly/2V9c5n4.■

Chris Letchford is Chair of Civil and Environmental Engineering at Rensselaer Polytechnic Institute and Chair of CROSEE.

J U N E 2 019

7

building BLOCKS

Drag Trusses

Practical Applications and Considerations in Light-Frame Construction

By David A. Fusco, P.E., LEED AP, and Judian Duran, E.I., LEED Green Associate

P

remanufactured, metal plate-connected, wood drag trusses can provide a pathed load-delivery mechanism

designed to assist and engage the lateral force-resisting system (LFRS) elements during high-wind or seismic events.

This article reviews common force transfer considerations in drag trusses and provides suggestions to design professionals for complying with ANSI/TPI 1-2014, National Design Standard for Metal Plate Connected Wood Truss Construction, the truss standard as referenced by the 2015 International Building Code (IBC) which addresses the design criteria for pre-engineered wood trusses. Practical ways to reduce potential failure mechanisms are also discussed.

Background and Applications

Figure 1. Drag truss parallel to the shear wall below.

Common Considerations on Plan Truss Layout The preferred truss layout aligns the drag truss above the vertical LFRS. From a connection standpoint, adding and locating a drag truss directly above the LFRS element (Figure 1) when the typical spacing does not place a truss above the LFRS is the most direct solution and functions equally well for light and heavy shear loads. An alternative load path and connection detail (Figure 2) can be used if the drag trusses and shear wall are not aligned vertically but works better for light shear loads and less well for heavy shear loads.

Ideally, the drag truss will span the length of and “sit” on the LFRS to develop the lateral and tension/compression overturning forces into the shear wall(s) below. Shear walls that exist under the drag truss, whether partial, sectioned, or full-length, might receive both lateral and gravity forces from the drag truss and can be simultaneously subjected to uplift. In order to ensure a uniform force transfer, the design professional is required, by sub-sections 2.3.2.4 (a), (b), (c) of the truss standard, to furnish and denote the location, orientation, and extent of each drag truss and the connected shear wall(s) below (Figure 3).

A typical truss profile from the truss manufacturer, per Section 2.3.5.5.(f ) (7) of the truss standard and IBC 2303.4.1.1(6), is required to show bearing and span conditions on the truss shop drawings. Shop drawings involving drag trusses should be reviewed and scrutinized by the design professional upon receipt.

A drag truss, sometimes referred to as a “collector” truss, is a singleor multi-ply pre-engineered truss designed to “drag,” distribute, and transfer shear loads generated within the plane of the diaphragm to the vertical LFRS elements.

Common design practice for trussed roof and floor diaphragms in light-frame construction relies on drag trusses to serve as boundary elements (boundary members and their connections) and provide the primary and/or auxiliary load-transfer mechanism configured to carry in-plane axial tension and/or compression forces to the shear walls below.

Intricate building layouts or complex building geometries result in diaphragm irregularities that can result in significant stress concentrations and increased demands at localized boundary element disruptions. Depending upon the increased diaphragm shear demand and building footprint, interior shear walls may need to be engaged. Framing a drag truss over an interior shear wall can help alleviate stress concentrations at diaphragm openings, re-entrant corners, offsets, or reduce the aspect ratio of the diaphragm.

Section 2.3.2.4 of ANSI/TPI 1-2014 sub-sections (a), (b), (c), and (d) require the design professional to provide distinct criteria on construction documents as they relate to drag trusses. As a result, the delegated truss manufacturer requires clear and well-documented plan callouts and details from the design professional. Several common considFigure 2. Trusses adjacent to the shear wall below. erations are summarized below.

8 STRUCTURE magazine

Truss Drag Force The drag force, as developed within the plane of the diaphragm sheathing and transferred into the truss, must be called

Detailing Considerations The effective strength and stiffness of horizontal wood diaphragms depend primarily on the mechanism of force transfer between adjacent wood structural panels.

In-plane unit diaphragm shears are usually limited by the localized load transfer nail capacity in the wood, rather than by the shear capacity of the panels. A drag truss acting as diaphragm support and boundary chord requires boundary nailing stipulation.

Drag trusses interior to a diaphragm will generally receive unit shears from

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two separate diaphragm spans. Both sequencing and pattern of the nailing must be evaluated.

The staggered nature of the sheathing generally dictates the sequencing. To avoid receiving varying edge and field nailing, special uniform nailing along the full extent of the truss top chord should be shown on both the framing notes and section cut detail.

The nailing pattern is generally dictated by the number of truss plies required to resist the shear demand along the top chord of the drag truss. A single-ply drag truss top and bottom chord may be adequate to handle the axial tension/compression force demanded but inadequate to accommodate the nailing. When this is the case, the design professional can specify to the truss manufacturer the required number of plies, the Figure 3. A typical illustration of truss shop drawings showing a drag truss over a shear wall. species, and the minimum size lumber required. The out on the framing plan or in the framing notes for the truss manu- design professional should also review the truss manufacturer’s nail facturer. The design professional should provide the magnitude of this spacing into the drag truss chord to verify that the allowable onforce, in units of force per unit length of the sectioned shear wall(s) center spacing is not exceeded. below (e.g., pounds per linear foot), and note whether it is factored For wind applications, uplift and lateral loads can occur simultaor nominal (unfactored), as stipulated by the applicable code under neously. Both uplift and lateral load connection capacities must be which the structure is designed. evaluated under combined loading and need to consider both loading The loading source – whether wind or seismic – also needs to be directions. Light gauge shear angle clips are typically used to transfer noted. The design professional must indicate if the specified loads the shear from the truss bottom chord to the top of the shear wall. are reduced (i.e., for ASD 0.6W or 0.7E).

Clip location, however, should not coincide with the metal plate, A typical truss profile by the truss manufacturer, per Section either the primary plate or the shear plate of the truss joint. Fastening 2.3.5.5.(f )(7) of the truss standard and IBC 2303.4.1.1(6), is required of the clip into a plate can potentially push the plate “teeth” out of to show superimposed drag force loads on the truss shop drawings. the truss bottom chord.

The reactions are generally expressed as “RL,” “R,” and “U.” RL is Details should include proper specification of nail sizes (pennythe maximum horizontal reaction in pounds per linear foot from weight, type, diameter, and length), nail spacing, nail-to-panel edge non-gravity loading or drag force loading, R is the maximum verti- tolerance, and account for the density-dependent capacity of the cal reaction from a gravity load case, and U is the maximum uplift framing member. Nail withdrawal capacity can be increased with reaction from a wind load case (Figure 3). properly fastened ring-shank nails and should be considered by the It should be noted that the truss designer is not responsible for design professional. Nailing patterns based on species, structural calculating drag force loads in the structure. However, it is the respon- sheathing thickness, and other specified requirements for diaphragms sibility of the truss designer to understand, review, and incorporate have also been tabulated and are included in the American Wood all applicable framing requirements and framing specifications into Council’s (AWC) Special Design Provisions for Wind and Seismic the design of the truss system. At the request of the design professional or local building official, the truss manufacturer must submit a preliminary truss submittal package to the design professional for review and approval before the manufacturing of the trusses.

J U N E 2 019

9

Figure 4. Overall plan view.

Figure 5. Local plan call-outs.

(SDPWS). The design professional should also specify any other tolerance-driven limits.

Bracing and Restraint Considerations Most wood light-frame construction in low-rise single- or multi-family dwellings have sloped roofs. Where wood structural paneling sheathes and shapes the diaphragm, it also mechanically restrains and braces pitched trusses to provide permanent lateral stability.

Similarly, Section 7.3.3.5 of the truss standard recognizes gypsum board installed directly to the bottom chord of the truss and fastened in accordance with ASTM C840 as an approved continuous lateral restraint and brace mechanism.

Top and bottom chords in compression can buckle laterally if not braced or appropriately reinforced. Truss-framed roof systems with high-pitched top chords and horizontal bottom chords have a smaller weak-axis moment of inertia than, say, a flat roof or floor joists, and under shear load will tend to deflect more in weak-axis bending.

The delegated truss manufacturer is responsible for the bracing that resists in-plane buckling of the individual components of the truss while subjected to compression forces. The design professional, however, is ultimately responsible for the design of the structure and, therefore, needs to confirm that the truss system bracing design has been addressed adequately.

The truss design drawings will indicate, per Section 2.3.5.5(m) and Section 2.3.5.5(o) of the truss standard, which web members require lateral restraint, the maximum axial forces in the truss members, as well as the type and location of the reinforcement. Commonly used types of lumber web reinforcement include scab bracing and T- and L-bracing. Also, the truss manufacturer will provide, within the truss design drawings, the fastening requirements from the reinforcement to the trusses.

Example The following example shows a common application typical to the nature of a drag truss. Consider a wood-framed, rectangular, singlestory family dwelling with a wind-driven uniform distributed loading of 200 plf acting along the diaphragm (Figure 4).

10 STRUCTURE magazine

The tributary-based unit diaphragm shear of 8 kips is to be transferred along the drag truss and must be called out on the framing plans. Shear walls of 10 feet and 8 feet are sectioned to accommodate for architectural features, which yields approximately 445 plf to be transferred into each shear wall below (Figure 5).

It should be noted that excessive drift can occur if the aspect ratio of the shear walls underneath the drag truss is not fully evaluated per AWC’s SDPWS. A plan call-out can be phrased as follows: delegated truss engineer to design drag truss for the bottom chord transfer load (ultimate) due to wind: W = 445 plf. For the condition presented in Figure 5, additional parameters on the plan include the bearing surface width, the location and extent of the truss, and shear walls below.

Summary This article provides a review of additional considerations required from the design professional for specifying common truss design variables to be used by the truss designer. Proper communication with the truss manufacturer is vital. It is the responsibility of the design professional to specify and detail the design intent and requirements in a clear and unambiguous manner on the construction documents.

For more information regarding the design and criteria of metal plate connected wood trusses, the reader is encouraged to read a free, read-only download of the standard ANSI/TPI 1-2014 on the Truss Plate Institute website, www.tpinst.org.â– The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

David A. Fusco is Vice-President Structural Engineer at Thornton Tomasetti, Inc. in Tampa, Florida. David is the Past President (2017-2018) of the Florida Structural Engineers Association Bay Area Chapter (FSEA-BAC) and now serves as Director for the Bay Area Chapter on the FSEA Board of Directors. (dfusco@thorntontomasetti.com) Judian Duran is an Engineer at Thornton Tomasetti, Inc. in Tampa, Florida.

Judian is a member of the Florida Structural Engineers Association Bay Area Chapter (FSEA-BAC) and Associate Member with the American Society of Civil Engineers West Coast Branch. (jduran@thorntontomasetti.com)

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structural DESIGN Post-Tensioning Design A Simple, Serviceable, and Safe Option

By Bijan O. Aalami, S.E., Ph.D., C.Eng.

T

here are three fundamentally different methods to design a post-tensioned con-

crete member: load balancing, rigorous, and straight. All three methods, when followed correctly, result in serviceable and safe members.

They differ substantially, however, both in the computational effort and potentially in the economy of the final design.

Figure 1. Contribution of post-tensioning in the reduction of the effect of dead load on the member.

The application of each method depends on the extent of the project, the expectations of the client, and the expertise of the design engineer.

Engineers who specialize in post-tensioning design and who are comfortable with large and complex projects benefit most from the rigorous method. The additional time and effort expended when using this method are offset by the economy in design.

increased application of post-tensioning in building construction.

The load-balancing method did away with the complexity of posttensioning design, which hindered its widespread adoption by structural engineers.

The load-balancing method relies on the common knowledge and daily practice of consulting engineers except for the recognition and explicit computation of hyperstatic moments from post-tensioning.

The computation and inclusion of the hyperstatic moments are necessary for the safety compliance of the member. The understanding and Design Options treatment of the hyperstatic effects is generally Most post-tensioned concrete buildings are an obstacle for many structural engineers. designed using the load-balancing method.

Figure 1 illustrates the concept in its basic While simple and intuitive, it requires the form. The tendon shown in part (a) is in the computation of hyperstatic (secondary) shape of a simple parabola in each span. Pulled moments – a somewhat unfamiliar concept to force P, the tendon exerts a uniform uplift for many engineers. Engineers who do not (PT ) shown in part (b). The uplift can be Figure 2. Hyperstatic reactions from post-tensioning. routinely design post-tensioning tend to considered to reduce the effect of the dead pass the design to those specialized in the load of the structure (DL), to (DL-PT) shown field. The rigorous method is detailed. It provides a closer picture in part (c) of the figure. of the member’s response to post-tensioning, as opposed to merely For deflection, stresses, and crack control, which are part of the servicemeeting the code-specified requirements of serviceability and ability check of the member, the effective downward force on the member safety. It applies where more reliable deformation, cracking, and is reduced by the PT force. This improves the service performance of post-cracking information of the member are sought. the member.

It is not uncommon that a designer may have to verify the adequacy The service load combination for “total load” is: of a post-tensioned member or design of a couple of the members of U = 1.00DL + 1.00PT + 1.00LL (Eqn. 1) the project with post-tensioning. In this case, the optimization of the At ultimate limit state (ULS), the member relies on the contribution post-tensioned members will have little impact on the overall economy of the tendons in resisting the demand forces. Hence the member of the project. For these applications, the straight method is preferred. shown in Figure 1c – with missing tendons – would not qualify.

The straight method relies on the knowledge and tools that structural Placing the tendon back in the member to provide resistance, the engineers commonly use, instead of the unique, rigorous method. configuration and load condition shown in Figure 1d applies.

Flexing of the member under the tendon force, coupled with the restraint of the supports to the member’s unrestricted flexing, result in Load Balancing the reactions R at the supports (Figure 2). The reactions are generated Basic load balancing, introduced by T. Y. Lin in the early 1960s when the member is not free to change its shape. Hence, “hyperstatic” and extended to non-prismatic members by the author, has led to reactions – also referred to as secondary reactions.

12 STRUCTURE magazine

The hyperstatic reactions (R) result in forces in the member, such as moments, that must be resisted by the post-tensioning tendons shown and possibly by adding mild-steel reinforcing bars.

For the safety of the structure at ULS, using ACI 318-14, Building Code Requirements for Structural Concrete and Commentary, the primary load combination is: U = 1.20DL + 1.60LL + 1.00HYP (Eqn. 2) Concrete and prestressing tendons provide the resistance to the force demand.

The shortfall in resistance, if any, is met by adding nonprestressed reinforcement.

In summary, (i) tendon is viewed removed to check the service condition; (ii) tendon is in place acting as reinforcement to check the safety condition.

prestressing, creep, and shrinkage provides the necessary information for the serviceability and safety load combinations of the member.

Figure 4 shows a typical finite element flat shell for slab design using the rigorous method. The element contains the prestressing tendon and nonprestressed reinforcement if any.

Other significant features to the rigorous method include: • For non-prismatic members, such as members with a change in thickness or step, the computation of balanced loads (forces from prestressing) in the load balancing method is complex and laborious. At each change in location of a member’s centroid, special treatment is required through the addition of local moments. In the rigorous method, the necessity of Rigorous Method special treatment at changes in the In the rigorous method, unlike the loadcross-sectional geometry of the balancing method, the prestressing steel member does not arise. is not considered as removed from the • The nonprestressed reinforcement member. The prestressing steel is treated in its actual size and position can Figure 3. Comparison between load balancing and rigorous the same as nonprestressed reinforcebe specified as part of the analysis modeling methods. The sub-images on the left represent model. For rigorous deflection ment, but with an initial stress. the Load Balancing Method, where prestressing is viewed calculation, it is not necessary to Figure 3 highlights the features of the as applied load. The sub-images on the right represent the substitute the reinforcement by its rigorous method and its comparison with Rigorous Method, where the prestressing steel is considered equivalent area in concrete to corthe load-balancing method. It is a partial as reinforcement with initial stress. rect the member’s stiffness. view of a post-tensioned member, sub• Cracked deflections can be computed with authenticity. The divided into finite elements. presence of prestressed and nonprestressed reinforcement in The sub-figures on the left (b, d, and f ) represent the load-balancing amount, location, and orientation included in the compumethod applied to a segment of the member. The prestressing tendon tational model (Figure 4 ) enables more realistic prediction is removed and is replaced by the forces that it exerted when in place of crack depth and the reduced stiffness necessary for the (b and d). At the application of load, and lapse of time, the member estimate of cracked deflection. deforms (f ). In a detailed analysis, the stresses in the deformed conThe necessity of computing hyperstatic actions from prestressing dition must be fine-tuned to account for losses in prestressing and change in dimensions of the segment from creep, shrinkage, and and including them in the ULS load combination remains as in the other stresses. load-balancing method.

The sub-figures on the right (c, e, and g) represent the rigorous method.

The “total” load combination for service condition is: Tendons are initialized with forces at stressing U = 1.00DL* + 1.00LL (Eqn. 3) (P). Nonprestressed basic reinforcement, if Note that, in this case, post-tensioning is any, is initialized with zero stress. an integral part of the structure similar to The solution shown in Figure 3g is the concrete. DL* includes the effects of postresult of the application of load, and lapse tensioning. For this reason, it does not appear of time, with due allowance for loss of stress explicitly in the load combination. in prestressing, creep, and shrinkage of conFor safety, ACI 318-14 gives the dead crete. The solution for the given time and load and hyperstatic effects of post-tenload reflects the instantaneous and longsioning, each with a different load factor. term effects of prestressing and concrete.

For this reason, the rigorous solution must The necessity of post-solution computation be broken into parts. The contribution of long-term losses and their allowance – as of post-tensioning must be extracted and is the case in the load-balancing method – Figure 4. Flat shell finite element showing the deducted from the solution to arrive at the does not arise. dead load values. inclusion of prestressing tendon and rebar within The breakdown of the solution at any the element. Tendon segment within the element U = 1.20DL + 1.60LL + 1.00HYP stage into contributions from dead, live, is initialized with prestressing force P. (Eqn. 4) continued on next page

J U N E 2 019

13

It is noteworthy that, in the rigorous method, the hyperstatic components of creep and shrinkage are of the same category as hyperstatic moments from prestressing. These are available as part of the solution. They impact the force demand for the load combination at ULS. ACI 318 does not require their inclusion in the load combination, however.

• Rigorous method o Tendons are retained in the member for service design. o Tendons are retained in the member for safety design.

• Straight method o Tendons are removed from the member for service design. o Tendons are kept out of the member Figure 5. Ultimate limit state (ULS) load diagram using for safety design. the straight method.

Three items enhance the performance and Straight Method improve the economy of a post-tensioned member. These are (i) uplift, In the straight method, the post-tensioning tendons are assumed (ii) precompression, and (iii) gain in tendon stress at ultimate limit removed from the member. The tendons are replaced by their uplift state, compared to service condition. effect (balanced load) on the member.

The first two design methods take advantage of all three enhancements.

The straight method accounts for the contribution of uplift from The straight method, in its simplest form, takes advantage of the prestressing in-service condition, identically to that of the load- tendon uplift only. Allowing for precompression at ULS improves balancing method. For the safety condition, however, the contribution the economy of this design option. The method does not benefit of post-tensioning is handled differently. from the gain in tendon stress at ULS, however.

The straight method does not require to compute and account for The straight method is safe and expeditious. It eliminates the effort the hyperstatic actions from prestressing as a separate design step. This of computing the hyperstatic actions from prestressing as a separate is advantageous for engineers who do not deal with post-tensioning design item and its explicit inclusion in design. regularly and may not be conversant with the computation of hyperAt ULS, prestressing steel is stretched beyond its service condition. static actions.

The stress gain from added tendon stretching, among other factors, The load combination for the “total” service condition is the same depends on whether the tendons are bonded or unbonded. Using as the other two design methods, namely: ACI-318-14, Section 20.3.2.4.1, the stress gain for bonded tendons U = 1.00DL + 1.00LL + 1.00PT (Eqn. 5) can be as much as 60 ksi (414 MPa) and for unbonded tendons (In this load combination, “total load” is considered. The coefficient LL up to 30 ksi (207 MPa). Assuming rebar at 60 ksi (414 MPa), the depends on the code case, but that of PT remains equal to 1.00.) gain in tendon stress translates to mild-steel reinforcement equal For ULS, post-tensioning is viewed as an externally applied load, similar to half the cross-sectional area of unbonded tendons and equal to to the service condition (Figure 5). Unlike the previous two methods, the the cross-sectional area of bonded tendons. This is the maximum post-tensioning tendons are not considered reinforcement for providing potential loss in economy of design when using the straight method. resistance to the applied load. The entire resistance to the “computed” Accounting for precompression from prestressing improves the design moment, if any, is provided by nonprestressed reinforcement, economy of design, however. along with compression from prestressing (Figure 5).

The code-mandated provision of minimum nonprestressed rebar in In this case, the load combination for the ULS is: prestressed members reduces the margin of an economic disadvantage U = 1.2DL + 1.6LL + 1.00PT (Eqn. 6) when using the straight method.

In the above load combination, PT is substituted for HYP (hyperWhere the building code greatly restricts the gain in tendon stress at static) when compared with the other two methods. The value ULS, the application of load-balancing and rigorous methods lose their of PT in this load combination is the same as that used for the advantage. As an example, the European code, Eurocode 2 (2004), service condition. limits the stress gain for unbonded tendons at ULS to 100 MPa (14.5 The preceding is complete and valid in arriving at a safe design. It ksi). This is a 9% gain over service condition. The allowed is not the most economical alternative, however. Its simplicity and meager gain in stress erodes the advantage of alternative expediency justify its application where the economy of the member design methods compared to the straight method.■ is not of critical concern.

Summary

The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

To summarize, the primary distinguishing features of the three design methods are: • Load-balancing method: o Post-tensioning tendons are considered removed from the member for service design. o Tendons are considered back in the member for safety design.

Bijan O. Aalami is Professor Emeritus, San Francisco State University, Principal, ADAPT Corporation; former Professor and Vice-Chancellor of Arya Mehr (Sharif) University. He has published extensively on design of post-tensioned structures and has held courses on post-tensioning in over 40 countries worldwide. (bijan@adaptsoft.com)

14 STRUCTURE magazine

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construction ISSUES Recommended Details for Reinforced Concrete Construction Part 1: Two-way Slabs

By David A. Fanella, Ph.D., S.E., P.E., F.ACI, F.ASCE, F.SEI, and Michael Mota, Ph.D., P.E., SECB, F.ACI, F.ASCE, F.SEI

This article is the first in a series on recommended reinforcement details for cast-in-place concrete construction.

T

wo-way slabs are generally defined as suspended slabs where the ratio of the long to the short side

of a slab panel is 2 or less. In two-way construction, load transfer is by bending in two directions. The main flexural reinforcement usually consists of two mats of reinforcing steel – a top mat and a bottom mat – that run predominately in the directions that are orthogonal and parallel to the rectangular grid of column lines. The bottom mat of reinforcement resists the positive bending moments at the critical sections in the span and is usually continuous over the entire slab area.

Figure 1. Reinforcing bars indicated on a plan view of the two-way slab system.

The top mat resists the negative bending moments at the critical sections adjacent to the supports in the column strips and middle strips. Guidelines and recommendations on the economical detailing of two-way reinforced concrete slabs are presented in this article.

General Detailing Recommendations

Figure 2. Minimum bars extensions for a two-way slab system.

16 STRUCTURE magazine

Two-way slabs can be detailed in different ways. One popular method is to call out the top and bottom amounts of reinforcement in the column strip (CS) and middle strip (MS) directly on a plan view of the floor or roof level, as shown in Figure 1 (in the figure, reinforcement is shown in one direction only for clarity; typically, the reinforcement in the perpendicular direction is also shown on the same plan).

It is assumed that the bars are uniformly spaced throughout the strip. For design professionals who are not familiar with the economics of reinforced concrete construction, it may be tempting to blindly copy the results from the computer program output directly to the structural drawings without thinking about the consequences of multiple bar sizes and spacing. While this may be adequate in some cases, economy is generally achieved by using the largest size reinforcing bars that satisfy strength and maximum spacing requirements and to repeat bar size and spacing as often as possible. This is illustrated in Figure 1, where #5 bars are used in both the top (T) and bottom (B) layers in

Figure 4. Recommended detail for additional reinforcement due to moment transfer at a slab-column joint.

Figure 3. A bar schedule for a two-way slab system.

bottom reinforcement across the whole slab, and to indicate any additional reinforcement that is required at the critical sections in the column and middle strips on the structural drawings; the spacing of these additional bars should be a multiple of that provided for the main bars.

In two-way slab systems without beams, the amount of negative reinforcement at the columns may need to be increased above that required for the negative bending moment at the critical section to satisfy the requirements of ACI 318-14, Section 8.4.2, pertaining to moment transfer at the slab-column joint. This additional reinforcement needs to be clearly documented on the structural drawings. Where required, the additional bars are usually provided in the column strips directly over the column. An example of a detail at an edge column is given in Figure 4. The 4-foot 4-inch dimension in the figure is the effective slab width that resists the transfer moment at the joint.

the column and middle strips, and the number of bars at similar critical sections are repeated as often as possible.

In conjunction with the bar callouts in Figure 1, minimum extensions for the top and bottom reinforcing bars in the column and middle strips must be provided. Figure 2 shows the minimum bar extensions given in ACI 318-14, Building Code Requirements for Structural Concrete and Commentary, Figure 8.7.4.1.3a, for two-way slabs without interior columnline beams; included in this figure are the structural integrity requirements in ACI 318-14, Section 8.7.4.2. The bar cutoff points apply to two-way slab systems subjected to the effects from gravity loads only; for systems that are part of the lateral force resisting system, an analysis must be performed to determine the required bar lengths.

For simpler detailing, all the top bars in the column strip can be made the same length (0.30ln "or" 0.33ln) instead of having at least half the bars longer than the others. Similarly, it is common for all the bottom bars in the middle strip to be continuous instead of cutting about half of the bars off near the faces of the columns. In general, overall cost savings are achieved by repetition even though more material may be used.

In lieu of calling out the number and size of reinforcing bars directly on the plan, a bar schedule like the one shown in Figure 3 can be used.

Another way to call out the reinforcement is to specify uniform top and Figure 5. Placement of reinforcement at offset columns.

continued on next page

J U N E 2 019

17

usually smaller than that of the beam reinforcing bars, the slab bars are typically placed above the top bars in the beam and in the same plane as the beam stirrups (Figure 6).

Corner Reinforcement ACI 318-14, Section 8.7.3, addresses exterior corners of slabs that are supported by stiff elements such as walls and edge beams. These stiff elements restrain the slab and cause additional bending moments at the exterior corners. Corner reinforcement must be provided in the top and bottom of the slab to resist these bending moments. According to ACI 318-14, Section 8.7.3.1.3, reinforcement must be placed parallel to the diagonal in the top of the slab and perpendicular to the diagonal in the bottom of the slab (Figure 7). Reinforcement parallel to the slab edges is permitted to be used instead of the diagonal bars. This layout is preferred because of ease of constructability.

Figure 6. Placement of reinforcement at column-line beams.

Regardless of the method that is used to identify the reinforcing bars, it is important to clearly indicate which reinforcing bars are to be placed in the outer and inner layers. Reinforcement in the direction of the larger design moments is usually placed in the outermost layers. Identifying the inner and outer layers of reinforcement can be accomplished by a note or a detail on the structural drawings.

Offset Columns Where columns are offset in plan, the top and bottom reinforcing bars should be placed orthogonally, if possible, as shown in Figure 5.

This minimizes constructability issues compared to skewed bars.

Where skewed bars are used, they should be provided in a separate, bottom layer; the top bars should be placed orthogonally. Top bars in the middle strip should be centered on a line connecting the column center lines.

Two-way Slab Systems with Beams The location of the top reinforcement in the slab must be clearly indicated on the structural drawings where column-line beams are present. Because the minimum cover to the slab reinforcing bars is

Drip Grooves Drip grooves or drip edges along the edge of a slab soffit can cause issues related to the required cover to the longitudinal reinforcement.

These grooves are usually formed using a form chamfer strip or a one-inch piece of dimensional lumber nailed to the formwork deck near the slab edge. The two ways to maintain required concrete cover are shown in Figure 8: (a) offset the bars crossing the groove, and (b) relocate the transverse and longitudinal layers of reinforcement so that the affected reinforcing bar (shown as a circle in the figure) is moved away from the groove, thereby maintaining the required cover.

Additional information on economical detailing of two-way slab systems can be found in the CRSI publication, Design Guide for Economical Reinforced Concrete Structures.â– The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

David A. Fanella is Senior Director of Engineering at the Concrete Reinforcing Steel Institute. (dfanella@crsi.org) Michael Mota is Vice President of Engineering at the Concrete Reinforcing Steel Institute. (mmota@crsi.org)

Figure 7. Required reinforcement at slab corners supported by stiff edge members: a) Reinforcement parallel and perpendicular to the diagonal; b) Reinforcement parallel to the slab edges.

18 STRUCTURE magazine

Figure 8. Slab with drip groove at the edge of soffit: a) Cover maintained by offsetting the bottom longitudinal reinforcing bars; b) Cover maintained by relocating longitudinal reinforcing bars.

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structural SUSTAINABILITY

Community Resilience through Mandatory Retrofit Ordinances What is the Role of the Structural Engineer?

By Erica Fischer, P.E., Keith Porter, P.E., Ph.D., and Elaina J. Sutley, Ph.D.

D

isasters are occurring with increasing severity and frequency [NOAA 2018]. Communities are also shifting design objectives to consider building functionality. As the severity increases and the design objective shifts, design standards generally rise with higher strength levels and stricter detailing requirements. With each code revision that raises safety levels, and/or incorporates increasing disaster intensity, a higher percentage of the existing building stock is deemed insufficient to meet the current design standards. The result is extraordinary federal, state, and local government spending on both disaster recovery and disaster preparation. For example, the Cascadia Subduction Zone, a roughly 622-mile (1,000 km) megathrust fault off the coast of British Columbia, Washington, Oregon, and Northern California, was discovered in the late 1980s. This means that structures in Oregon and Washington designed and built before this time were not designed to resist the forces that could be imposed on a structure in a potential magnitude 9.0 earthquake due to the rupture of this fault. Many buildings in lower Manhattan, Miami, and the Florida Keys are not designed for the flooding experienced over the last ten years. The increased frequency of disasters prevents communities from fully recovering before they are subjected to the hazard again. At the time of this writing, the island of Puerto Rico has only begun to recover from Hurricane Maria, yet the island was forced to prepare for the next hurricane season.

Governments face the choice of either paying to recover from disasters after the fact or implementing pre-disaster mitigation to avoid damage and loss in the first place. Extensive studies performed by the National Institute of Building Sciences’ Multihazard Mitigation Council [MMC, 2017] have shown that, on average, every dollar spent to mitigate existing public-sector buildings before a disaster saves six dollars in property losses, additional living expenses, business interruption, indirect economic losses, and the value of avoided deaths, injuries, and instances of post-traumatic stress disorder. MMC found that it is more economical to spend money on pre-disaster mitigation and retrofit buildings for the new level of hazards rather than responding after a disaster. MMC also found that design of new buildings to exceed certain requirements of the 2015 I-codes can also prove cost-effective, saving on average $4 for every $1 in additional construction cost for new construction.

Pre-disaster mitigation and retrofits could reduce sheltering and relocation needs for households, maintain business continuity, and prevent disruption to education either through school building damage or emergency sheltering use of the facility. Pre-disaster mitigation also can be planned and undertaken at a predictable, practical rate, rather than being suddenly and urgently required because of an unpredictable disaster. Cost effective, controlled mitigation results in a more stable social, economic, and political fabric of a community.

Some communities adopt mandatory retrofit programs for classes of vulnerable buildings. Often these retrofits do not bring old buildings to current code levels, but instead aim to reduce the danger posed by well-known weaknesses in common, vulnerable building types. Other 20 STRUCTURE magazine

Destruction in Armatrice, Italy, following an earthquake in 2016.

than understanding structural vulnerabilities, it is important to ask “what is structural engineering’s role in mandatory retrofit ordinances?”

For example, many California communities have required mandatory strengthening of parapets and other elements of unreinforced masonry bearing-wall buildings, as opposed to the much costlier effort to demolish the buildings or to replace their structural systems to meet current code. Other examples of seismic vulnerabilities addressed by mandatory retrofit programs include soft-story conditions on apartment buildings, roof-to-wall connections on tilt-up concrete buildings, and strengthening of older reinforced concrete buildings.

Decisions about what well-known deficiencies to mitigate, and how to do so, often seem to be driven by the knowledge that the deficiency exists, that it threatens safety, and that certain mitigation measures can be affordably implemented. That is not to say that these mandatory ordinances are driven by rigorous cost-benefit analysis or other canonical decision processes. More often, they seem to show an intuitive, ad-hoc decision-making process.

The ad-hoc nature of past retrofit ordinances raises many questions.

At the 2018 ASCE Structures Congress in Ft. Worth, Texas, an expert panel discussed the challenges associated with developing mandatory retrofits, after which the panelists and session attendees participated in break-out discussions. During the discussions, participants examined several issues: • Who pays for the retrofit?

• How does a community ensure that vulnerable populations are not left behind and do not end up living in the most vulnerable buildings in a community?

• Who decides what buildings have mandatory retrofits and what performance level are these buildings retrofitted to?

• How can society consider all of the different stakeholders and their potential biases?

• What is the risk that is accepted by the public when we do not retrofit unsafe, structurally deficient buildings?

• How can costs and benefits be equitably shared by the various stakeholders?

• How can society incentivize mandatory retrofits?

• What is the role of the structural engineer in developing mandatory retrofit ordinances?

For brevity, we have chosen three of these questions to explore and provide discussion within this article.

What Risk is Acceptable to the Public?

What risk does the public prefer? Let us first narrow the question by considering the breadth of risk measures one could discuss. Public and proprietary catastrophe risk analyses commonly measure risk in terms of collapse, life-threatening structural damage, fatalities and nonfatal injuries, and several measures of monetary loss.

Which of these measures matter most to the public? First: who is the public? Davis [1991] and Davis and Porter [2016] argue that “the public should be understood as including all those anywhere whose lack of information, technical knowledge, ability, or conditions for adequate deliberation renders them more or less vulnerable to the power that engineers wield on behalf of client or employer. The public is a collection or aggregate rather than an organized body. Unlike an electorate or corporation, it has interests, but no decision procedure – no will of its own.” ASCE’s Code of Ethics supports this conclusion, in that the Code of Ethics clearly distinguishes the public from the individual engineer, the engineer’s employer, and the employer’s client, and holds paramount – above the rest – the health, safety, and welfare of the public, that is, over the interests of the other groups.

So, which risk metrics matter most to the public? The FEMA P-58 project may represent the first attempt to identify “those aspects of earthquake-related risk that are of most concern to… stakeholders.”

Its authors held a workshop in 2001 to decide which measures to focus on as they developed a second generation of performance-based earthquake engineering procedures (Applied Technology Council, 2002). The workshop discussion assumed that life safety was provided and, therefore, the risk metrics discussed were financial loss, business interruption time, and building re-occupancy.

Davis and Porter (2016) present a large (800-person) public-opinion survey of adults in California and the central United States. The survey focuses solely on the perceptions and preference of the public for the seismic performance of new buildings. The survey asks which of a narrower set of risk metrics mattered most to the respondents.

Respondents cared most about the total number of community casualties (deaths and injuries) in a large earthquake.

In answer to the question of the level of fatality risk that is actually placed on the public in a large metropolitan earthquake, ShakeOut and HayWired (among other earthquake scenarios) suggest fatalities could reach thousands and nonfatal injuries could reach hundreds of thousands. With regards to what the public prefers, Davis and Porter (2016) suggest that the majority of respondents think that new buildings ought to be at least occupiable after a large urban earthquake. Davis and Porter (2016) findings suggest that respondents would be willing to pay the likely additional cost to achieve that level of performance.

This performance level is in stark contrast with what the building code intends to provide, which is essentially life safety (via a low collapse probability), even if buildings cannot be quickly, or even economically, repaired. Thus, the level of risk accepted by the public when we do not retrofit unsafe, structurally deficient buildings is a level that is much higher than the public’s preference.

Who Benefits from Mandatory Retrofits and Can Costs Be Shared?

The benefits of retrofits can accrue to occupants, owners, lenders, local jurisdictions, and anybody who visits or does business directly or indirectly with them. Depending on the intended increase in performance level, retrofits save lives, avoid search-and-rescue costs, reduce or avoid repairs, reduce business interruption costs, reduce insurance costs and claims, preserve tax revenues, and reduce costs for emergency sheltering.

The less damage a disaster causes, the more likely there will be sufficient

resources available to recover from it, the quicker the recovery, and inherently the more resilient the community grows.

Two crucial caveats of retrofits, stemming from the same attribute, must be addressed: retrofits are expensive and, without coupling appropriate funding mechanisms, retrofits can exacerbate the intersection of physical and social vulnerabilities in a community that will ultimately lead to unequal disaster impacts and differential recovery rates across cross-sections of communities (Sutley 2018).

Mandatory retrofits increase employment opportunities in the construction industry, including material suppliers, construction trades, construction contractors, structural engineers and architects, and the banks and other lenders that finance the work. On the other hand, in many cases, mandatory retrofits are viewed as a burden to building owners, who work to oppose them before they can be implemented.

If not appropriately explained, even the general public can work against efforts to develop mandatory retrofit ordinances. In Portland, the public started a petition in response to the development of a mandatory ordinance to retrofit unreinforced masonry bearing-wall buildings. This petition (https://saveportlandbuildings.com) asks for more equitable solutions so that social disparities do not become exacerbated through a mandatory retrofit ordinance.

Communities have choices to make when examining resilience and hazard mitigation. There are many options. Two of them include (1) re-zoning to remove people and infrastructure out of the most hazardous geographic areas of the community (i.e., moving people out of floodplains and away from active faults), and (2) retrofitting buildings and other structures within the hazardous regions to better withstand the hazard. These decisions can be easier in some communities than others depending on financial resources, political engagement, and available land. Rezoning to reduce flood damage is more straightforward than rezoning to reduce damage from earthquake shaking.

The cities of Boulder, Colorado, and Nashville, Tennessee, purchased land in the floodplain from private owners after floods in those communities in 2013 and 2010, respectively. The land was converted to green space. In Nashville, it was managed by non-governmental organizations and non-profit corporations. In the case of seismic hazards, however, rezoning has limited application: one can rezone to avoid building across mapped active faults, as in the case of California’s Alquist-Priolo special studies zones, but not to avoid shaking.

When communities can practically address mitigation through zoning, mitigation is equitable, at least superficially, in that all groups, including vulnerable populations, are prevented from living in highrisk regions of a city. However, it may be that more flood-prone areas are those that cost less, in which case more impoverished populations may be moved and wealthier ones not. One can see this imbalance in equity as either more beneficial for the poorer populations (because their long-term risk is reduced to a greater extent) or more harmful (because they suffer the greater short-term disruption).

Another way to promote mitigation is through incentives: tax breaks and grants, for example. Both the California Earthquake Authority (www.earthquakebracebolt.com) and the City of Portland (https://bit.ly/2CToDtd) have provided mitigation grants to owners of single-family dwellings, although not to owners of multi-family units. When supporting funding mechanisms are not provided, tenants often absorb the cost through substantial increases in rent. Rent increases can lead to gentrification and hurt vulnerable populations, particularly low income and older adults and renters who cannot afford higher rent payments. Rent increases push low-income renters out, leaving them to seek what is left of affordable housing, often consisting of un-retrofitted buildings in their home city, or permanent relocation to other, more affordable cities. continued on next page

J U N E 2 019

21

Thus, while the benefit of mandatory retrofitting is realized community-wide through decreased damage, disruption, and casualties, and minor spikes in the construction industry, the initial costs also need to be realized equitably. New financial mechanisms are imperative. It is not only up to public entities. Private insurance companies and banks can share the cost with building owners. Insurance companies often provide premium incentives for increasing safety through retrofits, such as the incentive program offered by the California Earthquake Authority, thus offsetting to some degree the owner’s retrofit cost. Similarly, the National Institute of Building Sciences is working to develop a resilience mortgage with a lower interest rate for risk remediation.

What is the Role of the Structural Engineer?

The business of developing seismic retrofit laws and ordinances is difficult because nobody possesses both the technical knowledge and authority required to know what is possible and appropriate. Policy decisions are the domain of elected officials, who often set objectives and the broad outlines of methods in law, and then delegate implementation details to regulators. The problem is that most policymakers and most regulators lack engineering knowledge of disaster risk – they do not understand the size of the problem, whether there is a solution, what the options are, how much they cost, and how much good they can do. Engineers can estimate risk, identify mitigation options and the ways in which we can estimate risk, and calculate costs and benefits. However, engineers seem to have set the standard lower than the public’s preferences, as evidenced by the gap between the building code’s seismic performance objectives for new buildings (mostly life safety) and the public’s apparent preference for buildings to withstand earthquakes and remain occupiable. The public’s preferences are often balanced with or opposed by engineers’ clients (such as developers and real estate investors) who are not representatives of the public, because of cost. How can engineers contribute meaningfully to practical solutions? The next section reviews the role of structural engineers in two efforts to develop mandatory retrofit ordinances.

Examples of Retrofit Ordinances San Francisco Soft-Story Ordinance. In 2015, San Francisco implemented the first mandatory aspect of its Earthquake Safety Implementation Program (ESIP), requiring mandatory evaluation and retrofit of high-occupancy soft-story wood frame dwellings. That element of ESIP grew out of two important actions by structural engineers. In 2006, Pat Buscovich, a San Francisco professional engineer, gave an interview to a reporter at the local newspaper, the San Francisco Chronicle [Smith, 2006], recounting how soft-story wood frame buildings were known to represent a significant risk to San Francisco housing. Mr. Buscovich added that city officials and the engineering community had known about the problem at least since the 1989 Loma Prieta earthquake almost 20 years prior, but had done little to solve the problem. Engineers provided options and cost/benefit information. Representatives of the public identified the risk measures they cared about and, together, the engineers and public representatives made the policy recommendation. Finally, the San Francisco Board of Supervisors and the San Francisco Department of Building Inspection defined the scope, timeline, and implementation details of the mandatory ordinance. Each group contributed the best of their expertise, stayed in their own lane, and did not try to intrude in the domain of the others.

Los Angeles Resilience by Design. The 2008 ShakeOut scenario [Jones et al. 2008] highlighted several problem areas with existing 22 STRUCTURE magazine

buildings. As part of ShakeOut, Krishnan and Muto (2008) showed that a large southern California earthquake could realistically cause the collapse of several older high-rise steel-frame buildings. Their study was peer-reviewed by several highly regarded structural engineers.

Similarly, Taciroglu and Khalili-Tehrani (2008) reminded ShakeOut participants of older nonductile concrete buildings that could collapse as well. Other studies addressed problems with telecommunications, oil and gas pipelines, fire following earthquake, water supply, railways, hospitals, and other topics.

Partly in reaction to ShakeOut, Los Angeles Mayor Eric Garcetti solicited help from USGS seismologist and local earthquake celebrity, Lucy Jones. Within a 1-year timeline, a mayoral task force that Jones led held more than 100 meetings with numerous engineering and other stakeholder groups. Of the many problems Los Angeles could address, the task force selected four problems to solve: pre-1980 non-ductile reinforced concrete buildings; pre-1980 soft-first-story buildings; water system infrastructure (including its impact on firefighting capability); and telecommunications infrastructure. The task force recommended detailed programs to mitigate risk in all four areas. In 2015, the City of Los Angeles passed Ordinance 183893, requiring the retrofit of pre-1978 wood-frame soft-story buildings and non-ductile concrete buildings. It also adopted seismic standards for new cellphone towers that require new freestanding cellphone towers to be built to the same seismic standards as public safety facilities, i.e., with an earthquake importance factor of 1.5.

The Los Angeles Department of Water and Power has begun to implement an infrastructure resilience program, including the installation of new pipe at a critical tunnel and targeted pipe replacement to construct a resilient backbone grid of earthquake-resistant pipe.

Conclusion In both examples above, structural engineers provided advice on the nature of the problem, the degree of risk, available mitigation options, and costs and benefits. This information allowed the public and their elected officials to make decisions that balanced public protection and economy. Communities looking to implement retrofit ordinances can look to these locales, the processes they followed, and the selection of participants with particular expertise as examples to follow. Structural engineers should continue to understand existing physical vulnerabilities and be capable of communicating the accepted risk of not retrofitting to their clients.■ The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

Erica Fischer is an Assistant Professor at Oregon State University in the School of Civil and Construction Engineering. Dr. Fischer’s research interests revolve around the resilience and robustness of structural systems affected by natural and man-made hazards. She has been a member of postearthquake reconnaissance team missions including Haiti (2010), Napa (2014), Italy (2016), and Mexico (2017). (erica.fischer@oregonstate.edu) Keith Porter is a Research Professor in the Department of Civil, Environmental, and Architectural Engineering at the University of Colorado Boulder, and principal of the risk consulting company SPA Risk LLC. He helped lead the ShakeOut Scenario and performed much of the engineering calculations underlying the San Francisco Community Action Plan for Seismic Safety’s soft-story element. (kporter@sparisk.com) Elaina J. Sutley is an Assistant Professor in Structural Engineering at the University of Kansas. Sutley’s research has an emphasis on wood buildings and housing. She actively develops interdisciplinary approaches to assess mitigation, predict losses, and model recovery. (enjsutley@ku.edu)

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structural REHABILITATION Snow Thermal Factors for Structural Renovations

By Michael O’Rourke, Ph.D., P.E., and Scott Russell, S.E., P.E.

T

he thermal factor, Ct, in the American Society of Civil Engineers’ Minimum Design Loads for Buildings and

Other Structures, ASCE 7-16, is intended to account for expected changes in roof snow loads due to heat flow through the roof. As one might expect, for poorly insulated structures with large amounts of thermal energy available to melt roof

Figure 1. Cathedral ceiling model considered by O’Rourke et al. (2010).

snow, the Ct factor is low (Ct = 0.85 for certain greenhouses) while, for very well insulated structures, the Ct factor is high (Ct = 1.30 for freezer buildings). For more common structures, the Ct factors in ASCE 7-16’s Table 7.3-2 are 1.1 and 1.2 for cold roofs and unheated structures, respectively. In relation to Table 7.3-2, cold roofs are those with an air passageway (typically inflow at the eave and outflow at a ridge vent) between the insulation layer below and the roofing material layer above. Finally, for the remaining heated structures (i.e., all structures except as indicated above), Ct = 1.0.

Figure 2. Warm-unvented attic mode.

These thermal factor values are based upon observations and engineering judgment. In relation to the observation, there are two available databases of simultaneous ground snow and roof snow load measurements. The first is a result of a United States Army Corps of Engineers’ Cold Regions Research and Engineering Laboratory (CRREL) project in the early 1980s while the second is a more extensive measurement program by the Agriculture University of Norway. Unfortunately, the Norwegian study involved unheated structures only and, hence, does not provide useful comparative information on the thermal factor. In the CRREL project, the structures were characterized as either heated or unheated. The ratio of roof-to-ground snow load for the heated structures averaged 0.54, while the average for the unheated structures was 0.67.

Hence, the CRREL observations are reasonably consistent with the thermal factor of 1.2 for unheated structures in comparison to Ct = 1.0 for heated structures, since 0.67/0.54 = 1.24. The Ct = 1.3 value proscribed in ASCE 7 for freezer buildings was an outgrowth of observation in a Structural Engineers Association of Washington (SEAW) report on snow-related structural collapse in the Pacific Northwest during the winter of 1997-1998. Roof snow loads on freezer buildings, absent drifting or sliding, were observed to be larger than the corresponding ground snow load. Since the roof and ground in the Pacific Northeast were subject to nominally the same snowfall from above, the difference has been attributed to more

ground snow melting over time due to the “warm” earth below, than roof snow melting due to the freezer space below.

The ASCE 7 thermal factors seem reasonably consistent with the limited available measurements. Until recently, these factors have not generated many comments or complaints from practitioners. However, federal agencies are now requiring (or “are recommending”) increased amounts of roof insulation for certain structural retrofit situations.

The laudable goal of such new requirements is improvements in a building’s energy efficiency. However, requiring increases in roof insulation raises questions regarding possible corresponding increases in roof snow loads. That is, a building that was properly classified as a heated structure, with Ct = 1.0, may morph into the equivalent of an unheated structure with Ct = 1.2 due to increases in roof insulation.

Clearly, the available measurement-based databases which do not provide the roof insulation level (i.e., R or U, see later in this article for definitions) cannot be used to address the increased insulation question. However, for a class of buildings, analytical estimates of expected eave ice dam size can be used to investigate the influence of increased roof insulation upon the thermal factor Ct. O’Rourke et al.

(2010) (hereafter referred to as the OGT paper) considered a cathedral ceiling structure, sketched in Figure 1, with no air passageway located in the insulation and roofing material layers between the heated interior air space below and the roof snow layer above. As such, the building would be considered as having a warm roof with Ct = 1.0.

24 STRUCTURE magazine

For a building with an attic, the Ct = 1.0 warm roof classification would also apply, if the attic is unvented and there is no air passageway between the roof insulation layer and the roofing materials layer, as sketched in Figure 2. In terms of the interior to exterior heat flow through the roof, the cathedral ceiling in Figure 1 behaves the same as the warm attic in Figure 2 if, as assumed herein, the warm interior air temperature (Figure 1) is the same as the warm attic air temperature (Figure 2). Herein, for simplicity, both roofs sketched in Figures 1 and 2 will be referred to as “warm-unvented attic” structures.

Note that if the cathedral ceiling structure sketched in Figure 1 had an air passage space between the roof insulation layer below and the roofing material layer above, it would correctly be classified as a cold roof with Ct = 1.1. Similarly, if the structure sketched in Figure 2 had a vented attic with the insulation layer in the attic floor, it also would be considered as having a cold roof with Ct = 1.1. Herein, again for simplicity, both these Ct = 1.1 structures will be referred to as “coldvented attic” structures.

In relation to the melting of roof snow, determining the heat flow through the warm-unvented attic structures is straight forward. All the thermal energy flowing upwards goes through the insulation layer and then the roof snow layer. However, for the cold-vented attic structure, some of the thermal energy is carried by air flow out through the vents (air passageways) while the rest flows up through the snow layer. The melting of roof snow on cold-vented attic buildings is due solely to the portion of the thermal energy which flows through the rooftop snow layer. Since the OGT paper considered only warm, unvented attic buildings, the retrofitted roof Ct values suggested herein are only for warm, unvented attic buildings sketched in Figures 1 and 2, that is buildings initially classified as Ct = 1.0.

a function of the indoor temperature Ti, the roof R-value Rroof (a measure of resistance to heat transfer), the roof U-value (a measure of heat transfer = 1/R), the roof slope, and the 50-year ground snow load Pg. The tabulated values are for a roof with a horizontal eave-toridge distance of 80 feet. It is a simple matter to convert the horizontal extent values into the corresponding reduction in roof snow load due to the aforementioned eave ice dam, again with a 20-year MRI. These estimated reductions in the 20-year MRI roof snow load due to thermal effects are presented in Table 1.

Notice that wetting of interior surfaces due to eave ice dams was considered to be a serviceability issue. As such, a 20-year MRI value for the horizontal extent of an ice and snow guard product was thought to be appropriate. However, for structural loads due to snow (a strength issue), a 50-year MRI value is desired. This “mismatch” of return periods is addressed below by calculating the Ct for a 20-year ground and roof load, and assuming the Ct would normally be the same for the case of a 50-year ground and roof loads.

As one would expect, the thermal losses in Table 1 are larger for an indoor temperature of 75°F than for an indoor temperature of 65°F. Similarly, the losses are larger for a roof R-value, Rroof, of 20 in comparison to those for Rroof = 50.

The losses for a location with (Pg)50 = 10 psf are zero for both Ti = 65°F and Ti = 75°F and Rroof between 20 and 50. For such locations, the thermal resistance of the roof snow layer is quite small and the thermal resistance of the insulation layer is comparatively large. As a result, the location of the melting point (temperature = 32°F) is in the insulation layer. No meltwater is generated at the base of the roof snow layer, and there is no eave ice dam formation.

As noted above, the thermal losses in Table 1 are for a 20-year MRI winter. For consistency, they need to be compared to the 20-year MRI

Recommended Thermal Factors The purpose of the OGT paper was to estimate the physical size of eave ice dams so that informed provisions for the horizontal extent of various “snow and ice guard” products could be determined. Such eave ice dams form when the outdoor temperature is below freezing and the bottom of the roof snow layer is 32°F. For these conditions, meltwater at the bottom of the roof snow layer forms, some of it flowing downslope to the eave and refreezing into the eave ice dam.

As such, the size of the eave ice dam can be used to back-calculate the amount of roof snow “lost” to roof-related thermal effects.

Note that these losses are due to thermal energy flowing from the interior of a heated, unvented attic to the exterior during times when the exterior temperature is below freezing. Such ice-dam-related thermal losses are unrelated to the loss of roof snow due to solar radiation effects, or above freezing exterior temperature.

It should be mentioned that solar radiation and above freezing exterior temperature also reduce the ground snow load as well as the roof snow load atop unheated buildings.

Table 8 in the 2010 OGT paper presents the horizontal extent of the 20-year Mean Recurrence Interval (MRI) eave ice dam, as

Table 1. Roof snow reduction (psf) due to thermal effects, 20-year MRI.

(Pg)50 Ti (°F)

Rroof U-Value (ft2·h·°F/BTU)

65

75

10 psf

20 psf

30 psf

40 psf

50 psf

20

0.050

0

0.70

1.81

2.92

4.02

30

0.033

0

0.24

0.69

1.13

1.57

40

0.025

0

0.05

0.35

0.55

0.75

50

0.020

0

0.01

0.09

0.17

0.25

20

0.050

0

1.36

3.23

5.10

6.96

30

0.033

0

0.46

1.28

2.10

2.90

40

0.025

0

0.22

0.64

1.06

1.48

50

0.020

0

0.06

0.22

0.39

0.96

Table 2. Recommended ASCE 7 Thermal Factor, Ct, for a warm, unvented attic structure as a function of the 50-year MRI ground snow load, (Pg)50, and the roof R-value, Rroof.

(Pg)50 Rroof (ft2·h·°F/BTU)

U-Value

10 psf

20 psf

30 psf

40 psf

50 psf

20

0.050

1.20

1.11

1.05

1.01

1.00

30

0.033

1.20

1.17

1.14

1.13

1.12

40

0.025

1.20

1.19

1.17

1.16

1.16

50

0.020

1.20

1.20

1.19

1.19

1.19

J U N E 2 019

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roof snow load, which requires the 20-year MRI ground snow load.

The commentary of ASCE 7-16 (specifically Table C7.2-3) provides factors for converting the 50-year MRI ground snow loads into other return periods. Interpolated from the tabulated factors, (Pg)20 = 0.77 (Pg)50 (Eqn. 1) The 20-year roof snow load for an unheated building (Ct = 1.2) becomes (Pr)20, unheated = 0.7CeCsIs(1.2)(Pg)20 (Eqn. 2) While the 20-year roof snow load for a heated building with a yet to be determined Ct is (Pr)20, heated = 0.7CeCsIs(Ct)(Pg)20 (Eqn. 3) However, the difference is simply the thermal losses in Table 1, or (Pr)20, heated = (Pr)20, unheated – Table 1 Value (Eqn. 4)

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Hence, the thermal factor for a warm, unvented attic building based upon a 20-year MRI becomes Table 1 Value Ct = 1.2  (Eqn. 5) 0.7CeCsIs (Pg )20 As noted above, it is assumed that the thermal factor for a 20-year MRI ground snow and a 20-year MRI roof snow load (as given in Equation 5) also applies to the desired case of 50-year MRI ground and roof loads.

The Ct value from Equation 5 was determined for all the cells in Table 1, for Ce = Cs = Is = 1.0. The values for Ti = 65°F and 75°F were then averaged. The resulting “calculated” thermal factors are presented in Table 2 (page 25).

Note, for the ground snow regions considered, ranging from 10 psf (e.g., Nashville, TN) to 50 psf (e.g., Minneapolis MN), an Rroof = 50 roof provides enough insulation that it behaves like an unheated structure. Also, for a location with (Pg)50 = 10 psf, (such as Nashville) a roof R-value of 20 or more provides enough roof insulation that it behaves like a Ct = 1.2 unheated structure. Conversely, for a location with (Pg)50 = 50 psf (such as Minneapolis), an Rroof = 20 roof behaves like an ordinary Ct = 1.0 heated structure.

Conclusion

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Over time, building energy efficiency, in general, and roof insulation requirements in particular, have been increasing.

Specifically, very well insulated roofs for heated buildings may now behave thermally like an unheated building.

Hence, the current simplified categories of “heated” or “unheated” for the ASCE 7 thermal factors are no longer adequate.

Based upon prior eave ice dam research on roof snow losses due to thermal effects, the authors have developed recommended thermal factor values as a function of the ground snow load and the amount of roof insulation. The recommended Ct values are for heated buildings with unvented roofs.

They are expected to be particularly useful in situations where an existing building retrofit includes increased roof insulation.■ The online version of this article contains references. Please visit www.STRUCTUREmag.org.

Michael O’Rourke is a Professor of Civil Engineering at Rensselaer Polytechnic Institute, Troy, NY. (orourm@rpi.edu) Scott Russell is the Manager of Group Technical Services for the Vulcraft/ Verco Group at Nucor Steel. (scott.russell@nucor.com)

NORTHRIDGE

25 YEARS LATER

Seismic Design and Hazard Maps: Before and After

By Nicolas Luco, Ph.D.

T

he 1994 Northridge earthquake generated world-record ground motions. At the time, the horizontal peak ground acceleration of 1.8 g measured by a seismometer in Tarzana was the largest ever. The same is true of the peak ground velocity of 148 cm/s measured in Granada Hills. Both measurements were within approximately 15 km of the source of the earthquake; they were also near most of the damage described in other articles of this series.

Consequently, the near-source design forces from the seismic zone maps in the Uniform Building Code (UBC) were increased. From the 1994 to 1997 editions, acceleration- and velocity-related nearsource factors were introduced. The factors increased the design forces in Zone 4, already the highest seismic zone, by a multiplier as large as 2.0. More enduringly, generational changes were made to the seismic design maps in the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures. The NEHRP maps were – and continue to be – adopted into the International Building Code (IBC), which supplanted the UBC and other model building codes. As described below, the changes to the NEHRP maps took advantage of another post-Northridge change: the modern generation of U.S. Geological Survey (USGS) National Seismic Hazard Maps.

Changes to Seismic Design Maps in Model Building Codes Before the Northridge earthquake, the NEHRP Provisions (1994 and preceding editions) provided maps of effective peak acceleration, Aa, and effective peak velocity-related acceleration, Av. Such maps were also used by two of the three model building codes of the time, the National Building Code (1993 edition) and the Standard Building Code (1994 edition). Even the seismic zone map in the UBC (1994 edition), shown in Figure 1, was derived from an Av map.

All of these maps were based – with some modifications, updates, and simplifications – on the Aa and Av maps first introduced in the 1978 Tentative Provisions for the Development of Seismic Regulations for Buildings, also known as ATC 3-06. In turn, these seismic design maps were loosely based – with truncations, modifications, and approximations – on a USGS peak ground acceleration (PGA) fully probabilistic hazard map published in 1976. For more information on pre-Northridge seismic design and hazard maps, a good reference is USGS Spectral Response Maps and their relationship with Seismic Design Forces in Building Codes.

In the 1997 edition of the NEHRP Provisions, the seismic design maps changed to the short-period and 1-second spectral response acceleration parameters, SS and S1. These parameters are more closely related to the seismic response of structures. The new maps became more directly based on the USGS National Seismic Hazard Maps.

These changes resulted from Project ’97, a collaboration between the developers of the NEHRP Provisions (i.e., the Building Seismic 28 STRUCTURE magazine

Figure 1. Seismic zone map of the 1994 Uniform Building Code (redrawn from the original by Kenneth Rukstales of the USGS).

Safety Council, BSSC, with funding from the Federal Emergency Management Agency, FEMA) and the USGS. Project ’97 also changed the seismic design maps from a nominal hazard level of 10% probability of exceedance in 50 years to 2%-in-50-year ground motions factored by two-thirds. However, this change was largely driven by 19th-century earthquakes in the central and eastern United States.

Like pre-1997 seismic design maps, the Project ’97 maps continued to truncate the probabilistic USGS National Seismic Hazard Maps. The pre-1997 maps were truncated at a PGA of 0.4 g. Above the roughly corresponding spectral response accelerations (i.e., wherever SS>1.5 g and S1>0.6 g), the Project ’97 maps were capped with newly defined deterministic ground motions. In some cases, these capped ground motions of the 1997 NEHRP Provisions were even larger than those corresponding to the near-source factors of the 1997 UBC. All of the Project ’97 changes described above have persisted through the seismic design maps used today, with some additional modifications. For more information on Project ’97, a good reference is a journal publication titled Development of Maximum Considered Earthquake Ground Motion Maps. The 1-second (S1) map from Project ’97 is shown in Figure 2. In addition to the 1997 NEHRP Provisions, the Project ’97 maps were later adopted into the 1998 ASCE 7 standard (Minimum Design Loads and Associated Criteria for Buildings and Other Structures) and the 2000 edition of the IBC. Ten and twenty years later, Project ’07 and Project ’17 resulted in additional changes reflected in the current (2015 NEHRP Provisions) and proposed nextgeneration (2020 NEHRP Provisions) seismic design maps.

Changes to USGS National Seismic Hazard Maps The USGS has developed National Seismic Hazard Maps since 1976, after publication of the Probabilistic Seismic Hazard Analysis (PSHA) approach. However, the modern generation of the USGS maps began in 1996, after the Northridge earthquake. The modern maps have benefitted from engagement of the earthquake science and engineering communities through workshops and public comment periods.

For example, from 1994 to 1995, the USGS convened six regional

workshops for the 1996 National Seismic Hazard Maps. The workshops resulted in substantial modifications to the methodologies used.

After posting interim maps on the Web in 1995, additional comments on the methods led to further modifications. A similar process has been followed for the 2002, 2008, 2014, and 2018 updates of the USGS maps.

Also persisting through the more recent USGS updates, several scientific advancements were made for the 1996 National Seismic Hazard Maps. First, about 500 faults in the western United States were added, including the Northridge fault. Their earthquake occurrence frequencies (required for probabilistic hazard mapping) were estimated from fault slip rates or trenching studies. Second, spatially smoothed historical earthquake locations supplemented the broader zones of seismicity modeled in previous USGS maps.

Such smoothed seismicity is used to capture the hazard from undiscovered sources of earthquakes, like the Northridge blind thrust fault that was revealed by the 1994 earthquake. Third, alternative models of seismic hazard were included in a logic tree formalism.

This is now common practice in order to represent the modeling uncertainty inherent in seismic hazard mapping. For more information on the 1996 USGS National Seismic Hazard Maps, see its documentation and/or Appendix B of the commentary of the 1997 NEHRP Provisions.

Other changes to the USGS National Seismic Hazard Maps that were initiated by the 1994 Northridge earthquake came to fruition after the 1996 maps. For example, the 2002 USGS maps applied five new models of ground motion attenuation that made use of data from the Northridge earthquake; the 1996 maps used three such models that predated the 1994 earthquake. From two of the new ground motion models, so-called hanging wall terms for thrust (like Northridge) or reverse earthquakes were used in the 2002 maps.

At one of the USGS workshops convened for the 2002 update, the community suggested including the phenomenon of near-source earthquake rupture directivity observed in the Northridge earthquake. Due to the computational requirements of doing so, and the dearth of published literature on approaches, directivity effects were not explicitly included in the 2002 maps. Even now, explicit inclusion of directivity remains a goal for future USGS updates.

Summary and Other Ongoing Changes

amplified ground motions in deep sedimentary basins at relatively long spectral response periods. Correspondingly, Project ’17 has recommended use of USGS ground motions for periods longer than 1 second (in addition to shorter periods). These and other ongoing changes are continuously advancing model building codes. The results are structural designs more commensurate with the seismic hazard, which benefit society by balancing constructions costs and seismic risks. Even so, more data from future earthquakes are needed (e.g., on rupture directivity), and more changes to seismic design and hazard maps are to come.■ The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

Nicolas Luco is a Research Structural Engineer with the USGS in Golden, Colorado. Since 2004, he has served as a liaison between the USGS National Seismic Hazard Mapping Project and the BSSC Provisions Update Committee, among other building code committees. (nluco@usgs.gov)

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Seismic design maps in model building codes, and the fully probabilistic USGS National Seismic Hazard Maps upon which they are based, have come a long way since ATC 3-06 (1978) and the 1976 USGS maps. Earthquakes like the 1994 Northridge event have taught new lessons and prompted changes from previously learned lessons. In particular, the Northridge earthquake generated nearsource ground motion data and other observations that improved the USGS maps. Post-Northridge seismic design maps in the NEHRP Provisions (and ASCE 7 and IBC) were also improved, by more directly basing them on the USGS updates. More recently, the 2018 USGS update has incorporated another effect observed from Northridge:

Figure 2. Seismic design map of the 1997 NEHRP Provisions providing spectral response accelerations, in units of %g, at a period of 1 second (redrawn from the original by Kenneth Rukstales of the USGS).

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J U N E 2 019

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INNOVATIVE CONCRETE CORE FOR TALL TOWERS

By Joe Ferzli, P.E., S.E., and Jason Thome, P.E., S.E.

KIARA

is a 41-story, 435-foot-tall luxury apartment tower over a seven-level podium with six levels of belowgrade parking in Seattle’s South Lake Union neighborhood, near Amazon’s global headquarters.

This modern, sleek tower features over 460 residential units, 15,600 square feet of retail space, and two rooftop decks – one at the top of the podium with a freestanding outdoor pavilion and the other at Level 41 with lush landscaping and an indoor sky lounge. The project is targeting LEED Silver. Construction started in 2016 and was completed in the summer of 2018 (Figure 1).

Figure 2. Kiara lateral system.

The tower is composed of two alternating curved vertical elements that slenderize and break down the massing into an elegant, layered composition of alternating glass colors. The top of the tower is terraced, and the rooftop provides an additional garden and owner’s lounge where residents can see stunning views of Lake Union, the Space Needle, and Elliot Bay.

Project Challenges The project posed several challenges. The unbalanced loading was about 2.5 times larger than the lateral loads due to seismic and wind. The site dropped 20 feet from East to West. The concrete core was limited in depth in the east-west direction to maximize the architectural program efficiency on the residential floor. Cary Kopczynski & Company, Inc. P.S. (CKC), structural engineers for the project, developed a ductile core lateral system with multiple 30 STRUCTURE magazine

elements of ductility and seismic energy dissipating mechanisms.

This was accomplished by incorporating steel fiber-reinforced concrete (SFRC) in the shear wall coupling beams on all four sides. SFRC significantly reduced reinforcing bar quantity and improved constructability, leading to a four-day cycle at the tower floors. SFRC eliminated the need for all diagonal bars, which are typically extremely congested and very difficult to install. Performance-based design (PBD) provided a means for the design of the SFRC coupling beams.

Ductile Core System Coupled core systems are a combination of wall piers that are solid or connected with coupling beams with various span to depth ratios.

Figure 2 illustrates the Kiara central

Figure 1. Kiara Tower.

Figure 3. Seismic drift comparison.

core wall system extending 48 stories above the mat foundation with six levels of basement.

The geometry of the lateral system has a significant impact on the overall behavior of the building. In a concrete core system, the seismic energy dissipating mechanisms are hinge regions located at the base of the wall piers and the ends of the coupling beams. A lateral system with low energy dissipation leads to higher shear, flexural, and diaphragm demands.

CKC performed a series of sensitivity studies to evaluate core shear and interstory drift, with and without additional openings and coupling beams being introduced into solid shear walls. It was observed that the shear demands in the system could be significantly reduced, both in the shear walls and at the transfer diaphragms, without excessive impact to the seismic drift of the building, by the addition of coupling beams at targeted locations on all sides of the core. Figure 3 compares the Kiara tower inter-story drift over the building height with and without coupling beams in the short direction of the core. As shown, the two plots are similar in that there is a minimal increase in seismic drift by adding the coupling beams at the center of the wall.

Coupling beams are detailed to accommodate large inelastic deformations while maintaining adequate strength. The internal forces generated in the system, such as core wall moment and shear, can be significantly reduced by increasing the system ductility.

Figure 4 illustrates two core wall studies for a 450-foot (138 m) tower and the 435-foot (134 m) Kiara tower for A and B cases, respectively, using a PBD approach. Both core wall systems are located in a highseismic region with special reinforced concrete shear wall lateral systems.

Nonlinear models were generated using PERFORM-3D software (by CSI) and were subjected to at least seven pairs of scaled maximum considered earthquake (MCE) level ground motions.

The effect of the coupling beams on the shear core demand is exemplified in the Core Wall A configuration study. The core is two-celled and utilizes SFRC coupling beams. All Figure 5. Core Wall A shear results.

coupling beams have a span-to-depth ratio of 3.0. A parametric study was completed where the number of coupling beams was varied in the short direction of the core and compared to solid wall piers. The results were quite dramatic (Figure 5).

With only one coupled wall and two solid walls, the peak core shear forces were approximately 22,000 kips (97,860 KN) at the dynamic base. When all three walls in the same direction Figure 4. Core Wall where coupled with SFRC coupling beams, the configurations. peak core shear forces reduced to 12,500 kips (55,602 KN), dropping the shear demand at the base of the core wall system by 45%.

The Kiara core, Core Wall B in Figure 4, is the perfect example of increasing ductility in core wall systems by introducing coupling beams. To mitigate high shear loads at the base of the structure due to seismic and unbalanced soil loading, as well as to decrease the required amount of shear reinforcing in the shear walls and transfer diaphragm, a series of coupling beams were added on all four sides of the central core. While some of these openings were required for the architectural program, several openings were added to introduce distributed ductility and additional energy dissipating mechanisms through coupling beams in the lateral force-resisting system.

The lateral force due to unbalanced soil loading was 2.5 times larger than the lateral seismic force at the base of the building. The central core resisted a large portion of the unbalanced soil lateral force. The introduction of coupling beams in the walls resisting the unbalanced soil load reduced the elastic shear force that would have been imposed on the core by dissipating the energy through the plastic hinging mechanisms at the end of the coupling beams. Thus, the building response improves during a seismic event and mitigates the damage to the vertical load carrying lateral elements such as the shear wall piers.

Steel Fiber-Reinforced Concrete The code-prescribed coupling beam options of conventional horizontal reinforcing or diagonally reinforced coupling beams, largely based on research conducted in New Zealand in the late 1960s and early 1970s, are appropriate in many systems but can have limitations.

For example, diagonally reinforced coupling beams have excellent hysteretic properties and drift capacities but are often quite congested and challenging to place in the field. The ends of the sloping diagonal bars must extend into the adjacent boundary elements of the shear walls, causing conflicts with the heavy vertical and transverse bars.

The cost saving was mostly related to labor time savings. The reinforcing material savings were offset it by the additional cost for the fiber concrete mix. However, the labor time to place the fiber coupling beams was significantly reduced compared to a conventional diagonal beam.

Additionally, when the project requires the geometry of the coupling beams to become relatively slender (e.g. a span-to-depth ratio on the order of 3.0), the effectiveness of the diagonal bars decreases with the resulting shallow angle of inclination. continued on next page

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One option is to utilize conventionally the concrete, allowing the fibers to detailed coupling beams. This type separate and disperse throughout the of detailing can accommodate larger mix. After workability was confirmed span-to-depth ratios, but both the at the site, a bucket was used to place drift capacity and the amount of coupling beam concrete. seismic energy able to be dissipated are reduced.

Constructability SFRC coupling beams have been reported to be a very effective Kiara is an excellent example alternative to conventionally reinforced of collaborative work between concrete beams while ensuring an researchers, design consultants, equivalent or higher level of strength and contractors to achieve success and ductility. A collaborative effort through teamwork on all aspects of Figure 6. SFRC coupling beam hysteresis loops calibration. between structural engineers and this project. The owner, architect, researchers has shown a potentially engineer, and contractor worked economical application of SFRC for closely from the early design phase to reinforced concrete buildings. This create a one-of-a-kind building that decade-long effort culminated with not only expresses the architectural a recent Charles Pankow Foundation intent, but also meets the project report, Evaluation of Seismic Behavior budget and schedule. of Coupling Beams with Various Types Weber Thompson is the architect of Steel Fiber Reinforced Concrete. responsible for Kiara’s architectural These fibers allow the diagonal design. Holland Partners and Holland reinforcement to be eliminated, Figure 7. Kiara SFRC coupling beam.

Construction were the owners and the transverse reinforcement to be general contractor responsible for the reduced, and generate value by increasing the speed of construction. overall coordination and day-to-day oversight of the project. Conco Kiara tower was designed using 386 SFRC coupling beams was the concrete subcontractor responsible for the structural frame throughout the central core. SFRC coupling beam nonlinear properties construction. Close communication between Holland Construction, were calibrated to replicate the hysteretic results from testing and Conco, and CKC was the key to the successful implementation of modeled using PERFORM-3D. Figure 6 illustrates the hysteresis Kiara’s unique core design. Early buildability meetings were scheduled loops for the SFRC test specimen and calibrated performance-3D between Conco and CKC to streamline the reinforcing detailing of component model. the core wall and to ensure compatibility with the formwork. Also, Since an SFRC coupling beam design procedure is not yet prescribed mockups of the SFRC coupling beams were done to finetune the in ACI 318, Building Code Requirements for Structural Concrete concrete mix design and placing procedure. Making constructability and Commentary, a series of nonlinear analyses were essential to central to the structural design and detailing created synergy between demonstrate acceptable performance. A geometry and concrete the design and construction team. The construction team went above mix were selected to fall within the range of tested specimens. The and beyond to quickly implement the constructability plans, leading flexural strength was determined at two critical locations: one at each to a four-day cycle at the tower levels. extreme end of the beam where the SFRC contribution to flexural strength is neglected, and one at the termination of the added u-bars Core Innovation where a flexural strength increase of approximately 15% from the SFRC material was considered. The shear strength was determined Concrete core wall geometry and SFRC coupling beams were vital by assuming the SFRC resists up to 3√fć, limited to 60% of the to deliver an efficient ductile core and added to the success of the total shear strength. The beam ends where plastic hinging may occur Kiara project. By breaking up many of the solid wall piers in the are detailed to be fully confined, while the interior of the beam has central shear wall core and using highly ductile coupling beams, transverse reinforcement proportioned per shear demands. the overall lateral force demands and reinforcement quantities were By introducing SFRC coupling beams in the core wall system, the significantly reduced. SFRC provides the structural engineering project energy dissipating mechanisms were enhanced to reduce the profession with a valuable tool for improving the constructability total internal forces in the lateral system. Even though similar core of reinforced concrete buildings in high seismic regions. The use wall response could be achieved using diagonally reinforced coupling of SFRC in Kiara resulted in a coupling beam design that beams, SFRC coupling beams provided a significant reinforcing eased reinforcing congestion, facilitated faster construction, quantity reduction of 30 to 40% and expedited the reinforcement and reduced rebar tonnage.■ placement schedule. Figure 7 illustrates the reinforcing in a typical Kiara tower coupling beam before placing steel fiber concrete (shown Joe Ferzli is a Senior Principal at Cary Kopczynski & Company, Inc. P.S. to the right) and the ability to accept beam penetrations.

(CKC), a structural engineering firm with offices in Seattle, San Francisco, and Chicago. He serves on the ACI Washington Chapter Board of Directors and The concrete for Kiara coupling beams contained Dramix© steel fibers he is a member of the ASCE 7-22 Concrete Subcommittee. (joef@ckcps.com) manufactured by Bekaert, with a fiber dosage of 200 lb/yd3 (120 kg/ m3) of concrete. The fibers are 0.015-inch (0.38mm) diameter by 1.18 Jason Thome is a Senior Associate at Cary Kopczynski & Company, Inc. inches (30mm) cold-drawn steel wire with hooked ends for anchorage.

P.S. (CKC). He is a current member of ACI Committee 374, PerformanceFibers were delivered to the producer in subsets of thirty. The subsets Based Seismic Design of Concrete Buildings. (jasont@ckcps.com) were bonded with water-soluble glue that dissolved when mixed into 32 STRUCTURE magazine

The new San Diego Central Courthouse is a bold and iconic civic landmark that replaces the seismically vulnerable existing courthouse facility.

Figure 1. View of new San Diego Central Courthouse at the main entrance. Courtesy of Bruce Damonte.

San Diego Central

COURTHOUSE Superior Court of California

By Mark P. Sarkisian, S.E., Peter L. Lee, S.E., and Rupa Garai, S.E.

The new Superior Court of California, San Diego Central Courthouse consolidates San Diego County’s criminal trial, family, probate, and civil courts into a 704,000-square-foot downtown facility integrated with the neighboring hall of justice and county jail facilities. The new courthouse, comprising a full city block, includes 71 courtrooms and consists of a 24-story 396-foot-tall tower and 4-story podium clad in glass and precast concrete with two below grade basement levels (Figure 1).

The courthouse replaces the adjacent existing courthouse facility built over an active earthquake fault zone. Designed for the Judicial Council of California (JCC) by the Skidmore, Owings & Merrill LLP (SOM) – Architecture, Structures, Interiors, and Graphics team working closely with the client and a group of trusted consultants – a complex set of site, programming, and cost constraint challenges were solved. The design achieves a high level of integration and efficiency while meeting client enhanced-seismic performance objectives. The facility construction was led by the Rudolph & Sletten (R&S) construction manager and contractor team which began in March 2014 and was completed in June 2017.

Site Selection and Context The assessment program, conducted in accordance with the California Trial Court Facilities Act of 2002 (CA Sen. Bill 1732), found that the existing San Diego County Courthouse/Old Jail building structures (built 1957-1962) were exposed to significant 34 STRUCTURE magazine

seismic risks. Seismic fault studies indicated that a projection of the known San Diego Fault runs through the northern and central portions of the existing County Courthouse/Old Jail site located in the central downtown San Diego district (Figure 2). For this reason, a new facility site location was investigated to fall outside the limiting 50-foot setback from a potential earthquake fault. A site was selected one block west of the existing facilities where, based on previous seismic fault studies, no active or potentially active faults were known to occur. Project-specific fault rupture boring and fault trenching investigations later confirmed that the site is not underlain by an active or potentially active fault. The new courthouse design accommodates a future underground tunnel planned to connect the courthouse with the county jail facility. Additionally, at the third level of the new courthouse, a pedestrian bridge connects with the existing Hall of Justice to the south. This bridge is supported by one tapered bridge bent column located within the sidewalk setback. The pedestrian bridge features an 85-foot cantilevered orthotropic steel deck structure to avoid placing new loads on the existing Hall of Justice.

Enhanced Seismic Design Recognizing that the new San Diego Central Courthouse lies in a region of high seismicity, and in close proximity to active and potentially active downtown San Diego earthquake fault zones, the 24-story above-grade and two levels below grade structure

was designed to meet “enhanced” building components including seismic performance objectives of both acceleration-sensitive and the Judicial Council of California deformation-sensitive elements.

(JCC), California Trial Court These components include exterior Facility Standards (CTCFS, 2011) wall elements, interior partition and the 2010 California Building walls, circulation and courtroom Code (CBC). In compliance with finishes, as well as anchorage requirements of CBC Table 1604.5, of suspended ceilings, lighting a risk Occupancy Category III fixtures, fire sprinkler/protection was assigned to the design of the lines, HVAC/MEP distribution courthouse in recognition of and equipment, elevator guide “Buildings and other structures rails/bracing, and egress stair that represent a substantial hazard elements. Potential damage to to human life in the event of failure, building contents including desktop … whose primary occupancy is electronics, office workstations, public assembly with an occupant lateral filing cabinets, and shelving load greater than 300.” Therefore, is also reduced. In summary, the an increased Importance Factor, I structure of the courthouse was = 1.25 per ASCE 7-05 Table 11.5designed to withstand the seismic 1, was used in the seismic design forces of a rare earthquake with less calculations. Additionally, the risk damage or disruption than a similar Occupancy Category III criteria Figure 2. Site selection relative to existing fault lines. building designed to code-minimum required a more restrictive seismic standards. story drift limit of 1.5% of story height.

In accordance with CTCFS (2011) Chapter 12, Criteria for Rare Response-History Analysis and Design Loads – Earthquake, the “normal” seismic performance of all new JCC facilities is intended to be above average to buildings designed by In accordance with the prescriptive requirements of ASCE 7-05 Chapters minimum prescriptive building code provisions. Due to its regional 12, 16, and 18, a 2-stage approach was taken in the structural analysis importance, close proximity to active earthquake faults, and public modeling and design of the SMF + VDD seismic force resisting system. assembly occupant loads, the San Diego Central Courthouse was Response-history analysis using a suite of seven site-specific ground designated to be designed for an “enhanced” seismic performance motions was used at both the Design Earthquake (DE) (475-yr) and objective. Enhanced performance is intended to limit damage and MCE (2475-yr) earthquake hazard levels using sets of 7-pairs of ground potential loss of use. motion orthogonal time history acceleration records representing fault The courthouse superstructure seismic force resisting system consists normal (FN) and fault parallel (FP) components. In Stage 1, a linear of distributed and redundant ductile steel “special moment frames” elastic SMF (damping, ζ=5%) with nonlinear VDD model is used (SMF) with “reduced beam section” (RBS) qualified connections in and, in Stage 2, a nonlinear SMF (damping, ζ=2.5%) with nonlinear both the transverse (E-W) and longitudinal (N-S) building directions VDD, both modeled using fast nonlinear analysis method (CSI ETABS (Figures 3 and 4, page 36). The SMF wide flange steel beams dissipate Nonlinear, V9.7.4). Stage 1 analyses included 725 modes to capture energy during strong earthquake ground shaking, while large SMF both VDD nonlinear link elements and vertical masses; Stage 2 analyses steel cruciform wide flange and built-up box columns provide included a total of 3000 modes capturing additional nonlinear link stability and strength to transfer vertical gravity and lateral loads. definition (FEMA 356) of the 2056 RBS steel connections. Four A total of 106 viscous damping devices (VDD) with extender envelope FN and FP ground motion orientation analyses, for each DE braces are interconnected with the SMF columns and distributed and MCE earthquake level, were undertaken. along the height of the structure in the continued on next page transverse (E-W) direction from Level 6 to the roof at Level 25. Configured in 2-bays on each of three column grid lines, the damping devices act to reduce earthquake-induced building story forces, drifts, and accelerations to provide enhanced seismic performance under peak demand of a CBC Maximum Considered Earthquake (MCE) magnitude with a 2,475-year average return period.

By controlling peak story drift demands, a reduction of inelastic deformations on the superstructure steel SMF connections is achieved. The enhanced performance also results in a reduction in damage to nonstructural Figure 3. Typical tower steel gravity composite and lateral SMF + VDD framing plan.

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and 358 acceptance criteria, with expected RBS behavior undergoing two cycles at both 4% and 5% drift levels (Figure 6 ).

Viscous Damping Device Energy Dissipation Designing tall, slender structures with steel SMF renders architectural versatility and high ductility but presents a unique challenge in controlling story drifts due to its low elastic stiffness, especially for the typical 16-foot story height. The application of advanced supplemental fluid viscous damping device (VDD) technologies to these tall SMF structures is known to effectively reduce potential damage to the structure, non-structural elements, and building contents, as well as loss of use under expected moderate to major earthquake events. The new Figure 4. Transverse building section with frame VDD. Figure 5. Transverse E-W direction DE story drift. courthouse VDD energy dissipating Additionally, the analyses are repeated to represent both lower bound system reduces seismically induced building story shears, story and upper bound damper device properties (+/-15%) at each of the drifts, floor accelerations, and inelastic rotational demands on the DE and MCE earthquake hazard levels as required by ASCE 7-05. steel SMF beam-column joints. The VDDs were further utilized to Peak transverse DE drifts are shown in Figure 5. The direct analysis resist the low-amplitude, velocity-excited portion of wind loads with method introduced in AISC 360-05 was utilized; seismic design effective linear damping exceeding 11%. This was substantiated by requirements of AISC 341-10 and AISC 358-10 were also satisfied. wind tunnel modeling, analytical methods, and a damper prototype testing program in conformance with ASCE 7 (Figure 7).

Special Moment Frame Qualification Testing

Strong Motion Instrumentation System

Motivated by both a significant reduction in steel tonnage cost savings and improved enhanced seismic performance, testing of three The new San Diego Central Courthouse has incorporated the full-scale specimens consisting of steel SMF with RBS connection installation of a strong motion instrumentation system to collect and to a large box column was undertaken during the construction process data obtained during strong earthquake ground shaking. The document design phase. The project specific qualification testing system was designed, specified and, procured during the base building allowed the use of larger box columns exceeding the AISC 358-10 design, construction document, and construction administration phases box column width and depth limitation of 24 inches. Typical SMF under the direction of the JCC and general contractor. Installation and beams in the project consist of W24 and W36 rolled shapes with commissioning support, and long-term maintenance of the system, is a maximum size of W36x302. The SMF columns included both provided by the California Strong Motion Instrumentation Program bi-axial W30 cruciform columns and bi-axial built-up box columns, (CSMIP), California Geological Survey, Department of Conservation, typically 33-inch-square, ranging from 20- to 36-inch dimension. as part of the statewide network of instrumented buildings. A total of The three tests were conducted at UC San Diego testing laboratory 28 accelerometers, 24 installed in the main building superstructure in December 2012, January 2013 and April 2013, respectively. All (19 horizontal and 5 vertical sensors) and 4 installed in the connecting three test specimens consisted of same size box column, 24by 36-inch with 2-inch plates and RBS beam W36x302 section. Specimens #1 and #2 exhibited only marginally acceptable brittle fracture failure modes. Post-test investigations and analysis revealed several conditions that contributed to the nonductile fractures. Detailing improvements, incorporated in Specimen #3 testing, successfully demonstrated overall ductility in Figure 6. SMF Specimen #3 test results (UCSD). conformance with AISC 341 36 STRUCTURE magazine

cantilevered pedestrian bridge to the adjacent HOJ at Level 3 (3 horizontal and 1 vertical sensors) were provided. All accelerometers are interconnected to computer controlled digital recording equipment at Level B2 with cabling for a common start, GPS timing, and synchronization. At the tower roof level, a GPS antenna and junction box are installed to support the system. The system is triggered by the sensors located at Level B2 at a threshold of 0.5% g, and de-triggered by the sensors located at Levels 23 and 25 at approximately the same acceleration (0.5-1.0% g).

Conclusion As structural engineers, it is not often that we are challenged to pursue design performance objectives that require extra effort, initiative, innovation, peer review, cost constraints, collaboration, and construction oversight – with the full support of client and user groups. This was the case for the design of the new San Diego Central Courthouse working with the Judicial Council of California and the San Diego Superior Court team’s leadership and well-established goals and objectives. In the design of the new courthouse, while meeting a range of complex design challenges – architectural integration, efficiency, physical security, sustainability, and cost control – a high priority was to achieve an “enhanced” seismic performance objective intended to limit damage and potential loss of use.■ The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

All authors are with the San Francisco office of Skidmore, Owings & Merrill LLP. Mark P. Sarkisian is a Partner. (mark.sarkisian@som.com) Peter L. Lee is an Associate Director. (peter.lee@som.com) Rupa Garai is an Associate Director. (rupa.garai@som.com)

Figure 7. Viscous Damping Devices (VDD).

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Pier 70 Building 113/114 RENOVATION San Francisco, CA

By Michael Gemmill, S.E., and Anthony Giammona, S.E.

Figure 1. View of Building 113/114 from the Northwest.

S

ituated along the historic San Francisco waterfront on San Francisco Port Authority property, a dilapidated former shipbuilding complex from the late 19th and early 20th century has been revitalized as part of a mas-

sive adaptive reuse project (Figure 1). The Pier 70 complex was the location of Union Iron Works, a premier West Coast shipbuilding facility through World Wars I and II until it was shuttered several decades ago. The anchor of the complex is Building 113/114, a two-block-long brick and steel building.

It was initially constructed as two buildings in 1885 and 1886, and subsequently joined by a reinforced concrete and steel “connector building” in 1914 (Figure 2). The nearly 500-foot-long, over 60-foot-tall, 82,000-square-foot building was built to serve as Union Iron Works’ machine, blacksmith, and boiler shops.

Orton Development, Inc., Marcy Wong Donn Logan Architects, Mark Hulbert Preservation Architecture, Nibbi Brothers General Contractors, and Nabih Youssef Structural Engineers teamed up to bring the historic building back to life.

Building History The building was primarily constructed of tall unreinforced masonry (URM) perimeter walls with steel trusses spanning between the perimeter walls and interior cast iron columns.

The long-span steel trusses used for the gable roof create a dramatic interior volume (Figure 3).

Heavy-industrial structural steel crane rails and associated support steel is located throughout the facility, offering a glimpse back to the former shipbuilding days.

38 STRUCTURE magazine

Figure 2. View of connector building.

Through over a century of service as an active ship-building facility, innumerable small and large alterations resulted in limited structural conditions that could be considered “typical.” As a result, an extensive field investigation was necessary before design and construction commenced. Heavy involvement by the structural engineers was also required during the construction phase to deal with the numerous atypical conditions that were uncovered.

Despite construction methods that are considered brittle and fragile by today’s standards, and an incomplete lateral-load-resisting system, historical records indicate that the building performed relatively well in the 1906 Great San Francisco Earthquake and the 1989 Loma Prieta earthquake. In fact, part of the modern retrofit work included reviewing historical reports, photos, and engineering drawings for relatively minor repairs performed after the 1906 event.

Damage from the 1906 earthquake was limited primarily to the outof-plane failure of two sections of brick walls on the east and west sides of the building. Fortunately, these two sections of wall were not significant load-bearing elements,

so the stability of the structure was maintained. The two sections of wall were subsequently replaced with corrugated metal cladding, and the ship-building activities resumed. It is believed that the large steel and cast iron gravity frame that was intended to support the crane in the building acted as a secondary lateral-resisting system to brace the top of the URM perimeter walls.

Building 113/114 is the focal point of the Pier 70-20th Street historic corridor, which includes many of the most historically significant buildings along San Francisco’s waterfront and is the easternmost segment of the rapidly developing Dogpatch neighborhood. At the end of its prior life, Building 113/114 was abandoned and red-tagged due to the damaged and noticeably leaning perimeter URM walls. Years of service in a marine environment, low-quality historic mortar, and poor soil conditions combined to take a toll on the perimeter walls, causing them to crack and visibly lean upwards of 18 inches out of plumb. A seismic retrofit – which was mandated by the Port of San Francisco building code because of the URM construction – and the adaptive reuse brings an abandoned treasure back into service.

Figure 3. Interior view of Eastern wing of Building 113.

Seismic Retrofit Scheme Collaboratively, the developer, general contractor, architects, and structural engineer aimed to develop a seismic retrofit that would stabilize the perimeter URM walls, provide enhanced seismic performance, and provide additional square footage (where possible). The resulting scheme utilizes two new levels of perimeter framing to stabilize and brace the perimeter walls (Figure 4). The first level of new framing braces the walls at mid-height and creates a lower mezzanine that significantly adds to the square footage of the building. The second level of new framing is an unoccupied upper mezzanine composed of horizontal steel diaphragm trusses used to brace the top of the walls and supplement the deficient existing roof diaphragm.

This approach also allowed the existing roof diaphragm to Figure 4. View of buckling restrained braced frames. remain largely as-is, thus preserving the historic character and avoiding substantial strengthening to brace the walls out of plane. A horizontal steel truss was selected to recreate the look concentrically braced frame (SCBF) system because of improved of an existing heavy timber horizontal truss that was utilized at the seismic performance. The new gravity-and-lateral-framing system same location to brace the original crane rails (Figure 5, page 40). essentially creates a “building-within-a-building” and uses the roof By locating the mezzanines around the perimeter, the spacious inte- and perimeter URM walls as cladding-only, thus completely relieving rior volume was preserved along the building’s central spine. Both the seismic demands of the existing structure. The existing, leaning new mezzanines were utilized as diaphragms to laterally support the URM walls were ultimately not straightened, so the lateral braces massive unreinforced masonry walls out of plane. Thru-bolts were and diaphragms were designed to resist the horizontal forces due used to tie the diaphragms to the walls and incorporated vintage plate to the permanent lean in the walls. washers to better align with other thru-bolts that had been added over Since an entirely new lateral-load-resisting system was created, the life of the building. The mezzanine slab edge is pulled back from the seismic system was designed based on the requirements of windows to minimize the visibility from the exterior. the 2013 California Building Code (CBC) for new buildings. A A complex new steel frame was added within the existing building to linear dynamic analysis was performed using SAP 2000 software provide a gravity system to support the new mezzanines and a lateral to design the seismic system. The geotechnical engineer, Langan system for both the new mezzanines and the remainder of the existing Engineers, developed a site-specific response spectrum. A global, building. Strengthening and reusing the existing columns was studied capacity-based design approach was incorporated by designing the but, because they are constructed of brittle cast iron, this approach diaphragms, connections, beams, columns, and foundations based was ultimately abandoned in lieu of an approach that provides new on the capacity of the braces. This allows the braces to act as the columns strategically located to minimize their visibility. fuse in the seismic-force-resisting system while protecting other Buckling-restrained braced (BRB) frames provide lateral resis- elements of the system. tance for the new and existing elements. This system was chosen Integrating the new structural system into the existing structural over a moment frame to limit seismic drifts and over a special system was a key design goal and one of the most significant challenges J U N E 2 019

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within the overall building. These spaces had been constructed at various times and served as offices and storehouses. They were constructed with materials that varied even more than the overall building. Each small building retrofit required multiple structural interventions, including new shear walls, strong-backs, and connections to the primary steel structure. Saving these spaces from wholesale demolition and turning them into unique modern office spaces that greet visitors at the main entrance contribute to the character and historic feel of the overall building.

Upgrade of Perimeter Walls

Figure 5. View of historic crane.

of the project. The historic fabric of the structure had to be maintained while balancing structural safety. Extensive coordination between the design architect, historic architect, general contractor, and structural engineer was required to locate nearly every new structural element within the existing building. Because no two parts of the building are symmetric, field verification for nearly every new above-grade element and exploratory excavation for every below-ground element was required.

Foundations and Bridges The foundation design was particularly challenging, as the building straddles the historic shoreline such that one half of the building is founded on rock and the other half is underlain by up to approximately 20 feet of fill that is susceptible to liquefaction. The fill was obtained by partial demolition of the adjacent Irish Hill, so named for the large population of Irish shipbuilding workers that inhabited the hill in the 1800s and early 1900s. A foundation system that included a combination of shallow footings over bedrock and a system of micro-piles joined by grade beams allowed new columns to be placed in the architecturally desirable locations, typically adjacent to existing columns. The new foundations are adjacent to, or even spanning over, existing foundations. Deep foundations were typically required at braced-frame locations because the frames resisted the lateral load of the heavy perimeter walls but supported minimal dead load of the mezzanines. Micro-piles, which are small diameter drilled and grouted piles, were selected because of their large uplift capacity, a minimal amount of spoils (the site soil is believed to be contaminated), and because the drill rig could be driven into difficult areas to access locations within the existing building.

New long-span steel bridges were incorporated to offer a striking addition to the central spine, provide for circulation within the mezzanines, and add square footage to the building. The bridges, designed with tapered end-plate girders to mimic the existing crane girders of the machine shop that remain as historic elements, also serve a key role in seismically linking the disparate mezzanines throughout the building. This allows the building to behave more uniformly in a seismic event.

Interior Spaces In addition to the global Building 113/114 retrofit, there were multiple small-scale retrofits of the existing wood framed ‘buildings’ contained

40 STRUCTURE magazine

Although the existing perimeter walls are now considered as cladding to the new steel frame, their severe deterioration still required a substantial amount of refurbishment. An extensive survey of the existing thick perimeter URM walls, typically constructed of three wythes of brick, was conducted to determine their current state and to develop a methodology to upgrade them. Despite their deteriorated condition and visible lean, preserving their historic character and finish was critical to the success and personality of the project.

Various tactics were used on a case-by-case basis including repointing mortar joints inside and out, installing ties to join wythes together, retrofitting foundations, and, most importantly, installing anchor bolts to tie the walls to the new concrete and steel framed mezzanines.

Some areas had deteriorated to such a degree that wholesale rebuilding with salvaged historic bricks was required. The ability of the walls to arch out of plane under seismic loading was evaluated based on h/t (height/thickness) and compared to limits contained in the URM retrofit building code. Where limits were exceeded, HSS strongback members were added to span between floors and brace the walls.

Conclusion The adaptive reuse of this project brings new life to an otherwise abandoned building. The new structural system was integrated into the existing building in a way that preserved the historic character, with the goal of allowing another century of use in this blossoming neighborhood on the San Francisco waterfront. Building 113/114 is currently complete and houses two technology tenants. The connector building is open to the public and links 20th Street to an internal public piazza, which was repurposed and intended to be used for community events.

The Design Team has also collaborated to seismically retrofit Buildings 14, 101, 102, 104, 115 and 116 within the same Pier 70 complex, creating approximately 300,000 square feet of mixed-use development.

A second phase of the development, spearheaded by developer Forest City, will involve retrofitting several other buildings while constructing several million square feet of new construction to complete the nearly 70-acre Pier 70 site. Building 113/114 is a significant component of the revitalized district, and its completion serves as a major milestone for the overall development.â– Photos courtesy of Dave Zahrobsky.

Michael Gemmill is the Managing Principal of the San Francisco office of Nabih Youssef Structural Engineers. (mgemmill@nyase.com) Anthony Giammona is a Vice President in the San Francisco office of Nabih Youssef Structural Engineers. (agiammona@nyase.com)

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Concrete Foundation Walls Subjected to Lateral Soil Loads

By George A. Merlo, P.E., and Anthony C. Merlo, P.E.

A

s structural engineers engaged in forensic investigations of failed structures, the authors have encountered concrete foundation walls at residential and retail facilities that have failed as the result of lateral soil pressure. In many instances, the walls are primarily reinforced with vertically oriented rebar with nominal horizontal steel, implying that the top of the wall is supported laterally by the framed floor system.

The 2012 International Residential Code (IRC) recognizes the distinction of a wall supported laterally at the top and one that is not.

Section R404.1.2.2 provides prescriptive requirements for horizontal and vertical reinforcement for concrete walls laterally supported at the top and bottom. Minimum reinforcement requirements are contained in tables R404.1.2(1) through R404.1.2(8). Section R404.1.2.2.2 addresses walls that are not laterally supported at the top and refers to section R404.1.3 which requires that the walls be designed in accordance with accepted engineering practice. Subsequently, the question arises as to what constitutes lateral support at the top of the wall?

It is the purpose of this article to outline the procedures for adequately designing concrete foundation walls subjected to lateral soil pressure.

Included are charts and graphs to assist in the design of structural systems with and without the wall laterally supported at the top.

Background Typical residential construction consists of a 2x wood sill plate bolted to the top of the foundation. Where floor joists are parallel to the wall without continuous periodic blocking between joists, no lateral support is provided at the top of the wall. For joists perpendicular to the

Figure 2. Flat plate subjected to equivalent fluid pressure (EFP).

42 STRUCTURE magazine

wall, lateral forces at the top of the wall must first be transferred from the wall to the sill plate via anchor bolts, from the sill plate into the floor joists via toe nails, Figure 1. Example of typical wall construction. and subsequently into the floor diaphragm.

An example of typical basement wall construction laterally supported at the top is illustrated in Figure 1 and will be evaluated as noted in Figure 2. Regardless of the type of construction utilized at the top of the wall, the methodology presented would apply to any construction detail at the top of the wall. The detail must be capable of transferring the wall lateral pressure into the floor diaphragm.

While numerous solutions exist for rectangular plates subjected to equivalent hydrostatic pressure (Timoshenko, et.al., and Portland Cement Association Document ST 63), the authors have found that the use of a commercially available finite element program provides the results necessary to determine the reactions at the top of the foundation and the resultant bending moments.

A comparison was made of the results of the FEM method and those contained in the Portland Cement Association (PCA) document to verify the validity of the FEM computer model. To evaluate the effects of various elements, three sizes were considered: 12- x 12-inch,

Figure 3. Plate simply supported on four sides.

Table 1. FEM method results compared to PAC-ST 63.

Element Size

12” x 12” 6”x 6” 3” x 3”

PCA

Error

.935

.922

.64%

-1.73

-1.77

-1.86

7%

.758

.769

.717

5%

Horizontal Moment Mid Span, ft-k/ft

.912

.928

Horizontal Moment Fixed at Support, ft-k/ft

-1.59

Vertical Moment Mid Span, ft-k/ft

.741

6- x 6-inch and 3- x 3-inch. The plate considered is 16 feet long by 8 feet high, 8 inches thick and free at the top, pinned at the base, and fixed at the vertical edges. The resulting moments are as shown in Table 1.

Utilization of an element size of 6 inches x 6 inches results in a difference of approximately 5 to 7% when compared to the PCA results. It is recommended that a 6- x 6-inch element size adequately defines the results.

Graphs were developed for various wall boundary conditions based on an element size of 6 inches x 6 inches to facilitate the design of the walls subjected to lateral soil pressure. Figure 3 represents a plate simply supported on all four sides subjected to an equivalent fluid pressure (EFP). Figure 3 provides the reaction at the top of the wall to evaluate if the connection of the top of the wall is adequate to transfer the reaction into the floor diaphragm.

For walls with an aspect ratio of horizontal span to wall height of two and greater, one can determine the horizontal reaction at the top of the wall by treating it as a simply supported beam. Aspect ratios less than two can be analyzed utilizing Figure 3. With aspect ratios greater than three, the vertical bending moments can be determined by treating the wall as a simply supported beam spanning from top to bottom. Less than three, one can utilize Figure 3 to determine the vertical bending moments.

If it is determined that the detail at the top of the wall cannot be economically designed to accommodate the reactive forces, then the wall must be treated as a plate with the top edge free.

Figure 4 (page 44) represents a plate free at the top and simply supported on the remaining three sides. Figure 5 and 6 (page 44) represent a plate simply supported at the base, top edge free, and vertical sides fixed.

It should be recognized that typical foundations do not consist of a single flat plate but several plates forming a box section.

Consequently, the conditions at the vertical edges cannot be represented as being pinned or totally fixed. If the length of the walls at the supports are equal and perpendicular to the wall in question, the supports can be fixed provided the wall pressure is equal on all sides.

In some cases, however, the designer may choose to conservatively assume simply supported conditions at the vertical edges to calculate the horizontal and vertical bending moments of the mid-span of the wall. Table 2 represents the effect of simply supported versus fixed supports for various wall geometries. In this case, an EFP of 40 pcf was utilized in calculating the results. As can be observed, for an approximate aspect ratio of 4 and greater, the difference in resulting bending moments is approximately 10% or less when treating the end conditions as simply supported versus fixed.

Case Study Assume an 8-inch concrete foundation wall, 8 feet high by 24 feet long, supported at the top of the wall with 5⁄8-inch anchor bolts spaced at 4 feet on-center, subjected to an equivalent fluid pressure (EFP) equal to 40 pcf.

Treating the wall as simply supported on all four sides from Figure 3, the top edge reaction equals 430 lbs/ft. With the anchor bolts spaced at 4 feet on-center, the shear force per anchor bolt equals 1720 lbs. Referring to the 2018 National Design Specification for Wood

Table 2. Bending moment comparison.

b/a

b, ft

Plate Simply Supported Three Sides With Top Edge Free

Km

Moment ft-lbs/ft

% Difference

Moment ft/lbs/ft

Km

Plate Simply Supported at Base, Top Edge Free, Sides Fixed

1

8

Horizontal Bending Moment Mid Span, ft lbs/ft

0.0327

670

48.8%

343

0.0168

Horizontal Bending Moment Mid Span, ft lbs/ft

Vertical Bending Moment Mid Span, ft lbs/ft

0.0236

483

37.3%

303

0.0148

Vertical Bending Moment Mid Span, ft lbs/ft

Horizontal Bending Moment Mid Span, ft lbs/ft

0.0754

1544

39.8%

929

0.0454

Horizontal Bending Moment Mid Span, ft lbs/ft

Vertical Bending Moment Mid Span, ft lbs/ft

0.0481

985

22.5%

764

0.0373

Vertical Bending Moment Mid Span, ft lbs/ft

Horizontal Bending Moment Mid Soan, ft lbs/ft

0.09076

1859

20.7%

1475

0.0720

Horizontal Bending Moment Mid Span, ft lbs/ft

Vertical Bending Moment Mid Span, ft lbs/ft

0.0589

1206

13.4%

1044

0.0510

Vertical Bending Moment Mid Span, ft lbs/ft

Horizontal Bending Moment Mid Span, ft lbs/ft

0.09532

1952

9.6%

1765

0.0862

Horizontal Bending Moment Mid Span, ft lbs/ft

Vertical Bending Moment Mid Span, ft lbs/ft

0.0627

1284

7.5%

1188

0.0980

Vertical Bending Moment Mid Span, ft lbs/ft

Horizontal Bending Moment Mid Span, ft lbs/ft

0.0966

1978

4.1%

1896

0.0926

Horizontal Bending Moment Mid Span, ft lbs/ft

Vertical Bending Moment Mid Span, ft lbs/ft

0.064

1311

3.8%

1262

0.0616

Vertical Bending Moment Mid Span, ft lbs/ft

2

3

4

5

16

24

32

40

J U N E 2 019

43

Figure 4. Plate simply supported on three sides with top edge free.

Figure 5. Plate simply supported at base, top edge free, sides fixed.

Construction, page 99, Table 12E, the allowable shear force is on the order of 400 pounds, less than the applied load.

The solution at this point could consist of decreasing the bolt spacing to approximately 1 foot on-center or designing the wall as a flat plate supported on three sides with the top edge free. In most cases, the contractor will balk at installing anchor bolts at 1 foot on-center which leaves the designer with the option of designing the wall reinforcing by treating it as a flat plate free at the top and supported on the three remaining sides.

To determine the horizontal and vertical steel, one could conservatively assume free at the top and pinned at the remaining three sides. The resulting horizontal and vertical bending moments can be calculated using Figure 4 and are equal to 1.86 and 1.21 ft-kips/ft, respectively.

Assuming “vertical edges fixed” results in a reduction of approximately 21% and 13% for the horizontal and vertical bending moments,

respectively, based on Figure 5. Horizontal bending moments at the vertical edges can be conservatively determined utilizing Figure 6, resulting in 3.77 ft-kips/ft.

Table 3 shows the required reinforcing steel.

Table 4 shows a comparison of the reinforcing for an 8- by 24-foot wall simply supported on four sides based on the prescriptive method in the 2012 IRC and the charts developed via the FEM method.

Concrete strength is 4 ksi and steel yield is 60 ksi, with reinforcing in the middle of the wall.

The IRC is based on the premise that the top of the wall provides adequate support such that only beam action is considered spanning from top to bottom. As noted earlier, in some cases an inadequate connection potentially exists at the top of the wall. As a result, one must first determine if the connection at the top of the wall can provide adequate support and then design the wall accordingly. The reinforcing vertical steel for the prescriptive IRC method is based upon treating the wall as a simply supported beam from top to bottom; whereas, the FEM method considers plate action. Safe designs require having justifiable assumptions that adequately account for real-world behavior. The use of these design aids can help the practicing engineer meet code requirements by providing a rational basis for answering the question “what constitutes lateral support at the top of the wall?”■ The online version of this article contains references.

Please visit www.STRUCTUREmag.org.

George A. Merlo and Anthony C. Merlo are the owners of Merlo Consulting Engineers, LLC in Englewood, Colorado, and serve as forensic engineers. (info@merlo4n6.com)

Figure 6. Plate simply supported at base, top edge free, sides fixed.

Table 3. Required reinforcing steel (case study).

8 feet high by 24 feet long

Free at Top and Pinned at Remaining Three Sides

Free at Top, Pinned at Bottom, and Fixed at Vertical Supports

Horizontal Steel Mid Span

#4 @ 14” o.c. (.17 in²/ft.)

#4 @ 18” o.c. (.13 in²/ft.)

Same as Mid Span

#4 @ 6” o.c. (.40 in²/ft.)

#4 @ 18” o.c. (.13 in²/ft.)

#4 @ 24” o.c. (.10 in²/ft.)

Horizontal Steel Vertical Supports Vertical Steel Table 4. Reinforcing comparison.

Method Horizontal Steel Mid Span Vertical Steel

44 STRUCTURE magazine

IRC

FEM Method

Not Specified

None Required

#6 @ 36” o.c. (.145 in²/ft.)

#4 @ 18” o.c. (.13 in²/ft.)

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professional ISSUES Sharing Claims Experience for Better Structural Engineering

By John G. Tawresey, S.E., F.TMS, F.SEI, Dist. M.ASCE

S

tructural Engineering is a great profession.

Every day structural engineers engage in something new, making visions into reality, visions often created by talented people many of whom are architects. Their contribution to society, while mostly hidden from the general public, cannot be discounted. Seldom do their structures fail. It is a great profession.

However, among all engineering disciplines, the structural engineer in private practice bears the highest cost, as a percentage of fees, for professional negligence insurance.

A 2001 study of 17,000 claims against A&E consultants resulted in the Table of 2011 study results below.

Insurance underwriters use two metrics to assess risk exposure, frequency, and severity. Frequency is how often we are sued, and severity is how much it costs each time. The Table (Factor) indicates a higher severity. Since architects’ fees are generally five times that of structural engineers, architects’ number of claims should be closer to 60 percent as a percentage of claims studied. Right?

The conclusion is that structural engineers experience more frequent and more severe claims. Why? Are structural engineers part of the problem? Are there things that can be done to reduce the chance and severity of a claim?

Here are some suggestions, with six simple rules to follow: From the same study, 13 percent of the claims, when finally resolved, are a result of the contract. The rest, 87 percent, were claims for professional negligence. Recall the generally accepted definition of professional negligence; the failure to use such care as a reasonably prudent person would use under similar circumstances in the same geographic region.

During a deposition in one of the author’s cases, an opposing expert was asked, “Will you testify at the time of trial that XXXX practiced below the standard of care?” The answer, “I don’t know what that means.” The author’s firm was sure they would win the case. They did not. What does the standard definition of professional negligence actually mean? How do you put it into everyday practice? When this definition is communicated to staff, what are they supposed to do?

Let us try a non-legal definition that may be more useful: to provide structural engineering services in accordance with expectations. At least this definition provides some direction and 46 STRUCTURE magazine

includes the expectations of the client, the owner, the contractor, and society.

To meet expectations, structural engineers need to know the expectations of those they serve and, to a large degree, they are in a position to control these expectations and reduce claims.

The control begins with the decision to accept participation in the project. There are internal concerns related to the ability to perform the work and external concerns like the project’s inherent technical risk and the expectations of all those involved, including expected schedule and budget. Many firms have checklists to evaluate participation. Checklists are also available through the American Council of Engineering Companies’ Coalition of American Structural Engineers (CASE).

External concerns can be controlled with an appropriate scope of services and contract.

Internal concerns are easier to control. If the internal concerns are not met (availability and capability of staff) then pass on the project, Rule 1.

A crucial part of controlling external expectations is the written scope of services, within a contract or otherwise communicated. Be diligent when writing the scope of services.

Some structural engineers forget that the project scope of services defines what they are required to do instead of all the things they are capable of doing, or worse, all the things they would like to be capable of doing.

The scope of services needs to be carefully written, Rule 2.

Add a corollary to the definition of standard of care: the standard of care means the level of engineering quality. The level of quality, including the amount of detail in the contract documents, depends on many factors, one of which is the expectations of the client. But, sometimes a structural engineer’s client, even a well-established architect, does not perceive or communicate the owner’s expectations.

For owners without design and construction experience (like many municipalities, school districts, churches), the architect often does not realize a need for full or expanded construction services. For example, if the project has a brick facade, employing a brick veneer on steel stud system, then the owner’s expectation of the building life requires investigation. If the building is expected to last more than 100 years, stainless steel ties are required in addition to other special detailing of the exterior, and the stud design should be fully defined in the contract documents.

The owner of a 100-year building does not expect the veneer to crack, even if the cracks are cosmetic.

Being detached from the owner’s expectations by clients could be one of the reasons structural engineers’ claims are more frequent and severe. Investigation into owners’ expectations is required, Rule 3.

Besides owner expectations, the level of quality depends on the type of project. For example, when the project is a wood frame condominium in a high seismic area, the inspection of the installation and placement of hold-downs is a required level of quality, not to mention a straight forward way to show the hold-down locations on plans so that even a dyslexic contractor knows where they need to be placed. For condominium developers, if the project does not include full construction and inspection services, then pass on the project, Rule 4.

As the design and construction proceed, the level of quality, usually occurring with changes in the scope of services, will adjust.

Appropriate adjustments are critical to the prevention of claims. When a change in scope occurs, structural engineers need to consider and communicate any resultant changes in quality.

Often, a situation can occur where a cost reduction proposal results in a design change and an

Table of 2001 study.

No. of Claims as a Percent of Claims Studied

Cost as a Percent of Claims Studied

Factor

Architects

48%

46%

.96

Civil/Survey

29%

26%

.90

Structural

12%

18%

1.50

Mechanical

9%

9.0%

1.00

Electrical

2%

1%

.50

Discipline

the 2016 NCSEA Winter Forum, to just adding to a redacted database identifying source and nature of claims.

If you would like to share a claim with the rest of the profession, please contact the author and arrangements will be made.

A final rule for this writing – If structural engineers want to help the profession avoid claims, add the following phrase, which has always been accepted without resistance, to the settlement agreement’s no-disclosure clause: “except for educational purposes,” Rule 7.■

John G. Tawresey is retired CFO of KPFF Consulting Engineers in Seattle, WA. He is a past president of The Masonry Society, past editor of the Masonry Society Journal, past president of the Structural Engineers Risk Management Council (SERMC), past president of the Structural Engineering Institute of ASCE, current member of the TMS 402/602 Main Committee, and is a member of the National Technical Programs Committee for SEI. He is an adjunct professor at the University of Washington. (johntaw@aol.com)

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associated change in the level of quality. Others typically make these decisions, and structural engineers often do not adequately record the event, record the reason for the change, or identify the decision maker. When scope and quality are changed, make a record, Rule 5.

On many projects, the level of quality necessary does not match the fees for the project.

The reasons are many. For example, accept for a moment that a structural engineer is involved in a project where the fee is less than the level of quality the project requires.

A typical response is to cut back on the hours applied to the project, reducing the design cost, and inadvertently lowering the quality of the contract documents and services.

The structural engineer has thus created a mismatch between external expectations and what they produce. A low fee is no defense. Do not change the level of quality (standard of care) on a project because of a low fee, Rule 6.

This is a small sample of lessons learned from the author’s 40 years of actual claims experience. What structural engineers do is technically complex. However, the dayto-day practice, the many decisions made daily, and their relationships with others in the construction industry is also complex. There are no easy answers to avoiding being sued for negligence; structural engineers can only reduce the probability. The most important thing is to become more educated about the source and nature of claims, and the best way to do this is by sharing claim stories.

Unfortunately, today the sharing of mistakes seldom occurs. Some complain that the problem is with lawyers and insurance companies. They run the show, drive the messages, and, as professionals, structural engineers act like sheep, just pawns in a much larger economic game set up by those who have different interests. Insurance companies, brokers/agencies, defense counsels, and resulting defense settlement agreements are blocking feedback to practicing structural engineers on mistakes engineers are making. However, it is not the attorney’s or insurance company’s fault. They are behaving perfectly rationally within their own interests, and structural engineers cannot expect them to change.

The problem is us; our profession needs to change.

The ASCE Committee on Claims Reduction and Management (CCRM) was formed to fill the information-sharing gap. Different levels of sharing claims information are being defined, ranging from full-disclosure, like the claim presentations at the last seven Structures Congresses and

J U N E 2 019

47

code UPDATES Groundbreaking: Tall Mass Timber Construction Types Included in 2021 IBC Historic Action by ICC Follows Ad Hoc Committee Recommendations

By Kenneth Bland, P.E.

T

he 2021 International Building Code (IBC) will introduce three new types of construction for fire-resistance-rated mass timber structures, the first significant addition to the types of construction in many years. Although still considered combustible construction, the structural frames of these buildings are designed for integrity in the unlikely event of fire exposure.

The Governmental Members of the International Code Council (ICC) approved a package of 14 proposals to recognize these new types of construction and related provisions. ICC’s rigorous code development process has led to the recognition of a strong, low-carbon alternative to traditional materials in the building and construction industry.

These changes expand the use of mass timber for larger and taller wood buildings up to 18 stories – a move welcomed by architects, engineers, and building developers.

The new construction types are designated as: • Type IV-A – Maximum 18 stories, with non-combustible protection such as gypsum wallboard on all mass timber elements and providing 2- and 3-hour fire resistance.

• Type IV-B – Maximum 12 stories, limited exposed mass timber is permitted and providing 2-hour fire resistance.

• Type IV-C – Maximum 9 stories, mass timber designed for 2-hour fire resistance.

The approval concludes several years of scientific research and testing, verifying that mass timber meets the performance standards called for by the most widely adopted U.S. building code.

ICC Code Development Process In late 2015, at the request of the American Wood Council, the ICC Board of Directors formed the Ad Hoc Committee on Tall Wood Buildings (AHC-TWB) to explore the science of tall wood buildings. The AHC-TWB was appointed in early 2016 and was led by Stephen DiGiovanni, P.E., a fire department protection engineer for Clark County, Nevada. Committee members consisted of code officials, fire officials, stakeholders, and other interested parties. If deemed appropriate after studying mass timber, the committee would act to develop and submit 48 STRUCTURE magazine

code-change proposals in the ICC process for the 2021 edition of the International codes.

The AHC-TWB determined fire testing was necessary to validate and verify that the performance level of passive fire protection intended by the IBC was retained. Five large-scale fire tests were developed to simulate characteristics of the three new construction types proposed.

Using cross-laminated timber (CLT) and glued laminated timber, a two-story building was constructed to resemble a fully furnished, onebedroom apartment on each level. Additionally, various configurations of exposed mass timber walls and ceilings, in addition to automatic sprinkler system effectiveness, were evaluated.

Corridors and an interior stairwell were instrumented to assess tenability conditions.

• Test 1: a mass timber structure with all interior surfaces fully protected with 2-layers of gypsum wallboard was subjected to a large furnishings-and-contents fire. The test was terminated after three hours without significant charring on the protected wood surfaces of the structure.

• Test 2: approximately 30 percent of the CLT ceiling area in the living room and bedroom were left exposed. The test was terminated after four hours, providing additional time to determine if there would be continued burning from the exposed CLT without intervention to extinguish the fire. Notably, once the fire consumed the furnishings and contents, the exposed CLT essentially self-extinguished due to the formation of char that protected the underlying wood and a gradual reduction in the room temperature.

• Test 3: parallel CLT walls were left exposed, one in the living room and one in the bedroom. Similar to Test 2, once the fire consumed the furnishings and contents, the exposed mass timber surfaces essentially self-extinguished.

• Tests 4 and 5: examined the effects of sprinkler protection. For both tests, all mass timber surfaces in the living room and bedroom were left exposed. Test 4 demonstrated that, under normal operating conditions, a single sprinkler easily controlled the fire. For Test 5, the

fire was allowed to grow in the compartment for 23 minutes before water was supplied to the sprinklers, which quickly controlled the fire.

The fires in Tests 1 – 3 were left to free-burn and reached a maximum peak heat release rate of 23 megawatts. The tests demonstrated that, if the required sprinkler system failed to operate, the resulting fire would eventually decay to a size easily controlled with limited intervention and without propagating to the next compartment.

After two years of study, the AHC-TWB submitted their proposals for consideration during the ICC 2018 Group A code development cycle. ICC’s code development process involves several public opportunities for new code proposals to be deliberated. First, proposals are considered by an ICC Code Development Committee. All 14 of the AHC-TWB proposals were recommended for approval or approval as modified, after considerable testimony. The proposals were deliberated a second time at the Public Comment Hearings where, once again, they received overwhelming support. Final approval occurred during ICC’s online voting process, cdpACCESS. Official voting results were announced in February 2019, and each of the 14 tall mass timber proposals was approved.

Moving Forward ICC’s 2021 model code development cycle continues throughout 2019 with the consideration of three additional proposals which address inspection and structural code requirements. The 2021 IBC, containing the complete package of tall wood proposals, is expected to be released in late 2020, along with the full set of 2021 I-codes.

The International Code Council develops and publishes a family of model codes suitable for local, regional, or statewide adoption. Once a governmental entity takes action to enact the code as law, all construction must be designed, constructed, and inspected for compliance.

For more information on the mass timber code changes, visit www.awc.org/tallmasstimber.■ Kenneth Bland is the Vice President of Codes & Regulations for the American Wood Council.

J U N E 2 019

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INSIGHTS Non-Proprietary Ultra-High-Performance Concrete For Structural Precast Applications

By John S. Lawler, Ph.D., P.E., and Elizabeth I. Nadelman, Ph.D.

U

ltra-high-performance concrete (UHPC) is an emerging class of concrete characterized by ultra-high compressive strength, significant post-cracking ductility, and exceptional durability compared to conventional concrete. UHPC is usually distinguished from conventional or high-performance concrete based on its compressive strength, which can range from a minimum of 17,000 psi to more than 30,000 psi. In addition, it also often exhibits post-cracking tensile strengths over 1,000 psi, a more than ten-fold decrease in chloride diffusion rates compared to conventional concrete, and essentially no susceptibility to deterioration by freezing and thawing (Russell & Graybeal, 2013). Most UHPC is also self-consolidating.

Because of these exceptional performance characteristics, UHPC has the potential to provide numerous benefits to the structural precast concrete industry, from enabling structural designs with smaller cross-sections, greater prestressing and reduced conventional reinforcing, to allowing the production of longer-span elements with fewer intermediate supports, thereby providing more usable space between bridge piers and building columns and accelerating the construction process. Despite its many advantages, the use of UHPC in structural precast concrete applications in the United States remains relatively limited. A primary reason for this is the perception that UHPC is an expensive, proprietary material that is too

complicated or too costly to produce.

However, ongoing research efforts by state and federal transportation agencies, university researchers, and private industry have sought to make UHPC more accessible, through the development of non-proprietary mixtures made from locally available materials. This article discusses some of the recent advances in the development and implementation of non-proprietary UHPC in the structural precast concrete industry.

Producing UHPC in a precaster’s mixer.

Developing a Non-Proprietary UHPC Mixture

UHPC’s exceptional performance is rooted in its constituent materials and their proportions. Materials for UHPC are selected not only for their reactive properties, but also for their ability to pack together to create a much denser matrix than typical for conventional concrete. Materials commonly used in UHPC include cement, silica fume, and a third powder material of intermediate size, such as ground silica, fly ash, slag, or silica fume. Fine sand – often finer than conventional concrete sand – fills out the rest of the matrix, with no coarse aggregates used. A high-range water reducing admixture (HRWRA), usually based on polycarboxylate technology, provides UHPC with its characteristic self-consolidating consistency and allows the mixture to remain flowable even at a water-binder ratio (w/b) of less than 0.20. Finally, high-strength steel fibers, added at dosages of 1 to 3 percent, by volume, provide the UHPC with its characteristic high tensile strength and post-cracking ductility.

Several concrete precasters have demonstrated that UHPC mixtures can also be produced using many of the materials already used for conventional concrete 328-foot (100-meter) Single Span BATU 6 UHPC Bridge Perak, Malaysia, production. In general, a produced at Dura Technology.

50 STRUCTURE magazine

typical structural precast plant may already have many of the core components of a UHPC mixture (that is, Portland cement, silica fume, supplementary cementitious materials, and an efficient HRWRA) in place and only need to acquire suitably fine sand and high-strength steel fiber for UHPC production. There are two conventional approaches for developing non-proprietary UHPC mixtures with locally available materials. The first is through iterative trial batching, where small batches of material are produced with various constituent materials and proportions, and the materials and proportions producing the most favorable characteristics are selected for production. The second is through optimization of theoretical “particle packing” models, which can be used to identify the particular dry constituent materials and proportions that will achieve the most favorable particle packing; this is usually accompanied by limited trial batching to identify the appropriate w/b, HRWR type, and HRWR dosage for the final mixture.

Performance of trial batches may be evaluated in terms of initial workability (flow spread), compressive strength, and flexural or tensile performance; evaluation of other characteristics, such as chloride penetration resistance or freeze-thaw durability may also be considered.

Producing Precast UHPC Elements Production of precast UHPC elements can be accomplished using most of the equipment already used for conventional precast production. Just about any type of mixer can produce UHPC – even a mixing truck – given enough mixing time. However,

efficient production is best facilitated by a high energy, high shear mixer, such as a planetary pan or horizontal twin-shaft mixer.

Unlike conventional concrete, UHPC is usually batched by first dry-mixing all of the powder materials and sand, then gradually adding the mixing water and HRWR and mixing until the material becomes fluid. This can take anywhere from 5 to 15 minutes, depending on the efficiency of the mixer and the HRWR used. Once a suitably flowable consistency is achieved, fibers are typically gradually added to prevent clumping, then the material is discharged from the mixer and placed into the forms.

Methods used to place UHPC into formwork are generally similar to methods used for self-consolidating concrete. However, consideration must be given to the influence of placement process on fiber alignment, which has a strong influence on mechanical performance (for example, tensile strength may be lower perpendicular to the direction of flow if fibers become preferentially aligned by the flow process). UHPC placements should also avoid the formation of cold joints or interfaces across which fibers do not bridge. If possible, UHPC should be placed in a single, continuous pour from a single

location, and internal vibration should be avoided to reduce the potential for fiber segregation. UHPC is susceptible to early-age drying and plastic shrinkage cracking; therefore, the element’s surface should be finished and covered immediately after placement. While heat can be added to accelerate initial hydration and facilitate strip out, a secondary post-cure thermal heat treatment (via Evaluation of the flexural performance of UHPC with ASTM C1609. high-temperature steam) may and building industries may be also be applied to the element to provide realized with the use of structural high early strength, or to “lock-in” voluprecast UHPC.■ metric changes due to shrinkage and creep.

As UHPC becomes more accessible to John S. Lawler is an Associate Principal at structural precasters, additional provisions Wiss, Janney, Elstner Associates, Inc. (WJE) in for the design and production of structural Northbrook, IL. His interests include concrete precast UHPC will become more readily materials evaluation and research, especially available. A project funded by PCI is curthe implementation of UHPC for precast rently underway with the primary goal of applications. (jlawler@wje.com) developing guidelines to be used by designElizabeth I. Nadelman is a Materials ers and producers of long-span, precast, Engineer at WJE. Her practice areas include pretensioned UHPC elements produced concrete materials evaluations, testing, and from non-proprietary UHPC mixtures. troubleshooting, with a focus on UHPC. Through these and other supporting efforts, (enadelman@wje.com) significant benefits to the transportation ADVERTISEMENT–For Advertiser Information, visit STRUCTUREmag.org

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historic STRUCTURES Rider Bridge Failure, New York & Erie Railroad 1850

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng., P.E., P.L.S.

T

his is the first in a series of articles on bridge failures in the United States. While failures are always newsworthy, especially if there is a loss of life, they are also lessons for bridge engineers to learn from. Failures, while traumatic, can be good teachers. Over the coming issues, this series will describe 20 failures, some of which you may have heard of. The earliest failure was in 1850 and the most recent in 1987.

Improved iron truss bridge.

This first article describes a failure of the Rider Bridge, a 66-foot-span cast and wrought iron bridge that Railroad Bridge, the committee would say that, in their opinion, it is crossed the Westcolang Creek, a short distance north of Lackawaxen, of great importance that railroad bridges should be constructed of a Pennsylvania. The bridge had opened about two years earlier and was less destructible material than wood, and that every effort to obtain similar to several of the type built earlier on the New York & Harlem that result is entitled to encouragement. Mr. Rider, in his plan, has Railroad. The use of iron in bridges in the United States started when aimed to effect this object, by a combination of wrought and cast iron, the Dunlap’s Creek Bridge, an original suspension bridge by James whereby with a limited amount of material the adequate strength is Finley in 1801, was replaced with a cast iron arch bridge spanning to be obtained. The committee think favorably of the combination, 100 feet over the same creek, built in 1839 by Richard Delafield. It and that for bridges of not too long span, believe that this plan will was not until Squire Whipple started to build cast and wrought iron be found useful; and they, therefore, recommend a favorable notice bridges across the Erie Canal with his 1841-patented design that iron on the part of the Institute.” It awarded him the Gold Medal for his bridge building gained momentum. Some of the first ones were by design. Several engineers, including Alan Campbell of the New York Richard Osborne who built a cast and wrought iron truss railroad & Harlem Railroad, praised the design. Campbell wrote, “Several road bridge at Manayunk for the Philadelphia & Reading Railroad in and street bridges over the railroad have also been built on this plan on 1845 and James Millholland who built a 54-foot-span wrought iron the Island of New York; these are light, durable, and cheap structures. plate girder bridge on the Baltimore & Ohio Railroad in 1846 at From my knowledge of Rider’s bridge, obtained from experience on Bolton Station. the road, I commend it to favorable notice of railroad companies.”

On November 26, 1845, Nathaniel Rider received a patent for a On the other hand, Squire Whipple was very critical of Rider’s cast and wrought iron bridge and built several for the New York & bridge, writing a lengthy article in the November 27, 1847, issue of Harlem Railroad, the first a 40-foot span, carrying streets over the the American Railroad Journal, tracks and tracks over streets in Manhattan in the next year. They It appears then, that a portion of the wrought iron in the bridge in were generally short spans, less than 60 feet. question is liable to a stress of more than 26,000 lbs. to the square inch, He entered his design at the 1846 Annual Fair of the American for a dead load on one track, of 1000 lbs. to the foot run, and the bridge Institute in New York City. A board of prominent engineers, Horatio endures the daily and rapid transit of the trains of the Harlem railroad.

Allen, John B. Jervis, and John D. Ward, wrote, “Of Mr. Rider's Iron If, then, bridges be built on the plans I have given in my work on bridges, which are estimated to sustain trains of twice the above weight, of 2000 lbs. to the foot run, with a stress of only 10,000 lbs. to the square inch of wrought iron on any part instead of 26,000 lbs. as above or less that ¼ of the stress for the same load, will not the chances of failure be reduced almost beyond the range of possibility, as far as wrought iron is concerned.

I have not time now to pursue this subject, nor is it necessary to my purpose. I only wished to point to the experimental lesson afforded by the bridge here spoken of.

If I have committed any errors in calculation or otherwise, I shall be very thankful to anyone who will point them out. I am certainly far from expecting to promote my own interest by continuing in error myself or by leading others into error on this subject.

Rider Patent with trussing underneath. continued on next page

J U N E 2 019

53

It should be noted that Whipple had an ordinary rate, but the engine had but written his book on bridges, in 1846 just got fairly off the solid track when the and 1847, in which he had presented, engineer heard a crack and felt something for the first time, a method to analyze giving way, whereupon he put on all the determinate trusses. It is unlikely that steam possible and succeeded in getting Rider knew of this method and that he the engine, which was a very powerful used rule-of-thumb methods, possibly one, upon the other side; but just as he based upon some experience, in bridgecleared the bridge, it went down with a work for his design. crash, carrying the tender with it; the Rider improved his bridge without cars following, tumbled into the abyss the under deck trussing in an advertiseone after another.” The report continued, ment printed in several journals of the “no one can imagine what caused the time. The American Railroad Journal, bridge to give way. It was inspected only the leading engineering journal of the last week, and pronounced perfectly safe. time, wrote, There are but two other iron bridges on There is, perhaps, no department of railthe road, and these are small ones. The road engineering, which requires, and railroad company has already set hands to is so susceptible of improvement, as the build wooden supports to them.” Three construction of bridges, and if the invenmen were killed in the wreck, in addition tion of Mr. Rider meets the exigency – as to many sheep, cattle, and hogs. we have reason to believe it does – it What followed was a case of an should be universally adopted on all of overreaction on the part of the president, our great lines of railroad.

Benjamin Loder, and the Board of From an examination of the cut of Mr. Directors of the railroad. The president Rider’s bridge, engineers will discover was quoted in the American Railroad that the invention is based upon scientific Haupt’s plan for Rider’s 109 th Street Bridge.

Journal of August 24, 1850, as follows, principles, and that the combination of This iron bridge, the giving way of which the trusses is such as to throw the weight of the passing body almost caused this accident, was the only one of a kind on the road, and exclusively in the longitudinal direction of the wrought iron tension when put up was supposed to be perfectly safe, and to possess many bars, thereby securing a power of endurance, such as iron alone can give, advantages over wooden structures in its greater strength, durability and which in practice is found to be more than equal to any emergency and exemption from fire. which the heaviest trains can subject them to. Where these bridges are It was not any experiment; but had been used before on several other now in daily use, we have observed that they appear to possess almost roads, and had proved satisfactory; It was supposed to be perfectly safe the solidity of terra firma; no sensible depression can be detected by the up to the moment of giving way – but the fact of its giving way without most vigilant observer during the passage of the trains. any previous warning decides the question in the mind of the directors The company furnishing these bridges are practiced and careful men, as to the entire disuse of iron bridges for the future…There are two determined that every step taken shall be justified by previous experience; more small iron bridges [by Squire Whipple] remaining on the road, this it is which has hitherto restrained them, in most cases from extending and those of different make and pattern, and supposed to be perfectly the span much beyond a hundred feet – though a bridge is just now safe, but they will be removed as soon as possible, and in their places completed at Buffalo of 160 feet, which will be tested during the present supplied with wooden bridges; until their removal they will be secure week, and which they doubt not will prove in all respects satisfactory, beyond the reach of accident by wooden supporters. Without expressing and authorize the building of longer spans, till finally they reach 300 any opinion as to the comparative safety of iron or wooden bridges, the feet – a limit which they deem entirely feasible. directors have decided to use no more iron bridges...

Herman Haupt, in his 1850 General Theory of Bridge Construction, Whipple wrote a long letter to the New York Tribune and the American published a plan of Rider’s 70-foot span for the 109th Street Bridge Railroad Journal under the heading The Breaking of the Iron Bridge on on the New York and Harlem Railroad that was built in 1847. After the New York & Erie Railroad, in which he said, in part, giving all the dimensions and weights of members, he noted, “The What was the cause of the accident? And how should they affect the result of this calculation shows that, with the dimensions assumed, confidence in iron bridges generally on railroads are questions in which the ties are stronger than the chords, and that heavier proportions the public are generally interested, and in which I, being a bridge are required to sustain a load of one ton per foot in addition to the engineer and builder, am particularly interested. The latter consideration weight of the structure.” induces me to attempt an answer to these inquiries and the former, I Until the summer of 1850, Rider’s company (he died in 1848) had hope will induce others to read and consider, according to their merits, received the praise of many and convinced several railroads to adopt the following remarks and statements… and build bridges to his design despite the criticism of Whipple and, Now, more than one person can testify to my having frequently remarked, to a lesser degree, Haupt. On July 31, 1850, his iron bridge on the in relation to those bridges, that they were badly proportioned and that New York & Erie Railroad failed. in certain parts which I have pointed out, they did not contain half The N.Y. Herald on August 3, 1850, reported the bridge which enough iron to render them safe and reliable; that I was surprised that crossed Hulburt’s Brook went down at “fifteen minutes past one” under they endured as much as they did and should not be disappointed to the load of an eastbound freight train carrying sheep, cattle, and hogs. hear of their failure at any time…

The paper reported, “There were seventeen cars besides the engine and The cause of the failure, therefore, I conclude to be the bad proportions tender. At the time the train approached the bridge, it was going at of the structure and weakness, i.e., want of proper size in some of the 54 STRUCTURE magazine

parts, defects noticed and frequently spoken of by me (in relation to other bridges on the same plan) years ago; that the accident affords no just grounds of apprehension from iron bridges properly proportioned and constructed, nor of discouragement from the introduction of such bridges on railroads.

Now the most important practical lesson taught by the sad catastrophe which gave occasion to this article, as it appears to me, is that a better understanding of the mechanical principles involved in the construction of bridges should prevail among engineers, and those having charge of such works, and that no structures of the kind, either of wood or iron, should be adopted or admitted, unless scientifically and systematically planned and proportioned throughout.

He then quoted from his 1847 article in the American Railroad Journal in which he pointed out the poor design of the bridge. He was chiefly concerned as the railroad removed several iron bridges he had built on the line. He also wrote a long illustrated letter to Appleton’s Mechanics Magazine and Engineer’s Journal that was published in their January 1851 issue.

Whipple’s colleague, John A. Roebling, who was building suspension aqueducts for the Delaware & Hudson Canal Company in the area, wrote, “Mr. Whipple’s remarks, as far as they go, are perfectly correct, in relation to the ‘Rider’ bridge, although I believe that the weakest points in the combination are at the ends next to the abutments. The investigations of that gentleman will always command the respect of those best able to judge, he having proved himself competent to the task in all his publications on bridge building, a qualification which few bridge builders have a right to claim – most structures of wood or of iron being put up at random, or by the ‘rule of thumb,’ the

inventors and constructors very frequently not even understanding the first elements of statics, much less the application of their principles…

In conclusion, I would express a hope, that the civil engineers of the United States, in view of their professional standing, will in a body disapprove of the wholesale veto, which the president and directors of the New York and Erie Railroad have seen fit in their wisdom to pass indiscriminately upon all iron bridges.”

Other Rider Bridges also failed shortly after but some survived for extended periods. The Baltimore & Ohio Railroad, under Benjamin Henry Latrobe, did not follow the lead of the New York & Erie and built many iron bridges to the patents of Wendel Bollman and Albert Fink in the 1850s. Squire Whipple built his first double intersection bridge with a span of 146 feet on the Albany Northern Railroad outside of Watervliet (West Troy) in 1853. His design became almost the standard railroad bridge until about 1890.

The failure, the first of an iron railroad bridge, did not stop the construction and failure of many poorly designed and built iron bridges. It did, however, start the slow process in which the railroads retained men, like Whipple, Roebling, Fink, etc., with engineering training to design and build their bridges rather than relying on bridge companies building bridges to their own plans or patents.■ Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He was formerly Director of Historic Bridge Programs for Clough, Harbour & Associates LLP in Albany, NY, and is now an Independent Consulting Engineer. (fgriggsjr@twc.com)

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SPOTLIGHT Structural Design for Art

By Ahraaz Qureishi, P.E.

C

olumbus, IN, located about 45 minutes south of Indianapolis, is home to a surprising amount of pioneering architecture given its modest population of 50,000. The city has added so much to the industry that the American Institute of Architects has named it sixth in the country for contributing to architectural innovation and design. The Exhibit Columbus festival celebrates the city’s design heritage each year by presenting current art and architecture. Wiikiaami, presented at the 2017 exhibit, is a steel art structure with an asymmetric, conical shape inspired by the homes of the Myaamia people indigenous to Indiana.

Wiikiaami translates to “wigwam” in the native language and is a modern take on the dwellings that were once present in Indiana. The structure tapers from a 37- by 22-foot ellipse at the base to a 6- by 4-foot ellipse at the top. One face of the cone is nearly vertical while the other face gradually slopes up over the 30-foot structure height – from 25° to nearly vertical.

For Wiikiaami’s skin, the architect/artist chose perforated steel scales reminiscent of traditional textiles that wrapped wigwams. A discreet structural frame was desired to promote visibility through the light-pervious scales and achieve the architect’s vision. After considering several options to accommodate the unique geometry and limited budget, A706 weldable reinforcement was chosen to comprise the intricate frame as it was much less expensive than curved HSS sections, and most bar sizes could be bent on-site to the required shape.

The shape’s curvature and steep slopes made geometry-dependent loads, such as wind, snow drifts, and ice, particularly challenging.

Making assumptions to simplify or generalize loads over the full height resulted in a rebar frame too dense for the architect’s vision and bar diameters too large to bend in-field.

Therefore, the structure was discretized over its height and loads were calculated for each section individually. Ice loads were significant due to the unusually high surface area-tovolume ratio of the skin and rebar frame, and governing gravity loads came from the accumulation of snow and ice.

Lateral translation under gravity loads was significant due to lack of plumbness and further compounded high P-Δ effects and wind displacements. Frame-action from vertical members (termed “spines”) and horizontal members (termed “ribs”) alone did not provide enough lateral stiffness. Considering STRUCTURE magazine

the perforated scales as sheathing to transfer in-plane shear was ineffective as they were curved (thus too flexible) and required too much costly field welding. Diagonal members (termed “struts”) spiraling up the sloping face were added to brace the structure.

Added stiffness from the struts, though required, introduced additional complications.

Struts presented a more direct load path as they were inherently much stiffer than spines, but strut size was unreasonable in initial layouts since they could not be bent in-field. The final layout was optimized by positioning struts to have less direct load paths to foundation elements so that the load was shared with spines.

A 3-D model of the full structure accounted for curved members, P-δ/Δ effects, and increased flexural stiffness from continuous members and welded connections. Maximum vector deflection from vertical loads was 1 inch and occurred 20 feet up the sloped face. However, lateral displacement contributed to over 60% of this deflection, which is an artifact of Wiikiaami’s out-of-plumb shape.

Rib sizes ranged from #8 bars at the base to #4 bars at the top and were located every 2 feet along the structure height. A total of 20 spines, mostly consisting of #7 and #8 bars with some of the straighter spines being #9s, ran full height and were stiffened by 14 struts. Struts on the sloped face were mostly #8s, while the vertical face required much more curvature and could only be accommodated by #6 and #7 bars. The limitation of being able to field-bend bars was especially challenging for members near the base subjected to nearly 4,000 pounds of compression, a considerable force for slender, unbraced rebar up to 4 feet long. The force required to bend the larger bars with small radii of curvature was studied and quantified, ensuring specified bar sizes were constructible.

Nodal connections between spines, struts, and ribs needed to be quickly assembled and costeffective, but also rigid enough for adequate stability. Effective buckling lengths for bars in compression were decreased by accounting for fixity at member ends so rotational stiffness at nodes was necessary. This was especially critical as the mid-point of many curved members were out-of-plane and experienced P-δ moments from the onset of loading. Though bolted options were considered, welded connections were ultimately specified due to their stiffness and ability to transfer moments between intersecting members. Welded

Pierce Engineers, Inc., was an Award Winner for its Wiikiaami project in the 2018 Annual Excellence in Structural Engineering Awards Program in the Category – Other Structures. Courtesy of Hadley Fruits.

connections also engaged bars instantly without any bolt slip or “lag,” essential for stability considering the number of connections present. Thus, rebar at connections was welded with 5⁄16 -inch U-groove welds on each side of the bar. Similarly, rebar was lapped 12 inches at splices and welded on each side.

To accurately fabricate the complex geometry, plywood platforms were CNC cut to the correct elliptic dimensions and erected with scaffolding.

Spines and ribs were then welded along this precise skeleton. Rebar landed on a 4XS pipe section, factory curved to the elliptic shape and supported by helical anchors. Wiikiaami’s conical shape generated substantial thrust forces at the base, despite its small, lightweight frame.

Eight 23⁄8-inch-diameter helical anchors provided adequate lateral and uplift capacities. Specific anchors were purposely located away from struts and spines to force load distribution in the base pipe to adjacent, lighter loaded anchors.

Wiikiaami was very well received at Exhibit Columbus and was a 2017 Miller Prize recipient. Though it was anticipated to be a temporary structure, it is one of 4 installations from the exhibit remaining in place while a new, permanent location is sought.■ Ahraaz Qureishi is a Structural Engineer with Pierce Engineers in their Madison, WI office.

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NCSEA NCSEA News

National Council of Structural Engineers Associations

2019 Summit: Fun, Sun, Disney and the Best in SE Education!

November 12–15, 2019 · Disneyland ® Hotel · Anaheim, CA The Summit draws the best of the structural engineering field together for high-quality education by expert speakers, a dynamic trade show with over 60 exhibitors, and compelling peer-to-peer networking at a variety of events and receptions. Join us for this growing and dynamic event designed to advance the industry!

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• New Format! Beginning on Tuesday and ending Friday afternoon, the program will offer more education (over 16 hours) and less overlap.

• The SE3 National Symposium, the first Structural Engineering Engagement and Equity (SE3) symposium to be held in conjunction with a national engineering conference • NCSEA's Award Celebration, honoring the ingenuity, creativity, and innovation of structural engineering and structural engineers across the association and around the world!

• A Celebration of Structural Engineering hosted by Computers and Structures, Inc. • Five Keynote presentations: • A Perspective on the Future of Consulting Engineers, Stacy J. Bartoletti • Talk Nerdy to Me: Science Not Communicated is Science Not Done, Melissa Marshall • The Power of Connection, Avery Bang, Ph.D • Moving Beyond Life Safety for Community Recovery, Lucy Jones, Ph.D, • Structural Engineering: Indispensable to Civilization So Why Don't We Have More Influence? Why Don't We Make More Money?, Ashraf Habibullah, S.E. • A Welcome Event sponsored by the Structural Engineers Association of California (SEAOC) • And much, much more!

Visit www.ncsea.com to learn more about this year's Summit and to reserve your room at the Disneyland® Hotel; rooms are going fast!

2019 Excellence in Structural Engineering Awards NCSEA's Excellence in Structural Engineering Awards annually highlights some of the best examples of structural engineering ingenuity throughout the world. Projects are judged on innovative design, engineering achievement, and creativity. Multiple winners are presented in seven categories with an outstanding winner being chosen and announced at NCSEA's Structural Engineering Summit in Anaheim, California this November.

The awards are presented in the following categories: • New Buildings Under $20 Million • New Buildings $20 Million to $100 Million • New Buildings over $100 Million • New Bridges/Transportation Structures • Forensic/Renovation/Retrofit/Rehabilitation Structures up to $20 Million • Forensic/Renovation/Retrofit/Rehabilitation Structures over $20 Million • Other Structures

Entries are due by 11:59 pm on July 16, 2019. Structural engineers and structural engineering firms are encouraged to enter.

More information about the awards along with submission instructions can be found on www.ncsea.com.

Now Accepting Nominations for NCSEA Special Awards

NCSEA's Special Awards are presented each year at the Structural Engineering Summit. These awards are presented to NCSEA members who have provided outstanding service and commitment to the association and to the structural engineering field. Special Awards are granted to worthy recipients in four different categories: the NCSEA Service Award, the Robert Cornforth Award, the Susan M. Frey NCSEA Educator Award, and the James Delahay Award.

The NCSEA Service Award is presented to an individual who has worked for the betterment of NCSEA to a degree that is beyond the norm of volunteerism. It is given to someone who has made a clear and indisputable contribution to the organization and therefore to the profession. 2018 Recipient: Barry Arnold, P.E., S.E. The Robert Cornforth Award is presented to an individual for exceptional dedication and exemplary service to a Member organization and to the profession. 2018 Recipient: Ryan A. Kersting, S.E. The Susan M. Frey NCSEA Educator Award is presented to an individual who has a genuine interest in, and extraordinary talent for, effective instruction for practicing structural engineers. 2018 Recipient: Ronald O. Hamburger, S.E. The James Delahay Award is presented at the recommendation of the NCSEA Code Advisory Committee, to recognize outstanding individual contributions towards the development of building codes and standards. 2018 Recipient: Jonathan (Jon) C. Siu, P.E., S.E. Visit www.ncsea.com to submit your nomination by July 19, 2019.

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News from the National Council of Structural Engineers Associations

LEANING OUT

July 18, 2019

A BASIA + LEONARD MYSZYNSKI FILM

A Can’t Miss Event for Structural Engineers

10:00 am Pacific, 11:00 am Mountain, 12:00 pm Central, 1:00 pm Eastern Members: $45 Nonmembers: $75 Course will award 1 hour of continuing education. Diamond Review Approved in all 50 states.

Leaning Out is the story of the World Trade Center's lead structural engineer, Leslie E Robertson. Leaning Out is not only a film about innovation, wind engineering, and visionary collaborations, it is a film about Robertson, a man who oversaw the construction of the tallest building on the planet at the time, and is haunted by its collapse and the events of 9/11.

This online-event will include a screening of the film and a Q&A session with Leslie Robertson. Attendees will earn 1 hour of Diamond Review-approved continuing education.

Register by visiting www.ncsea.com.

NCSEA Webinars

Register by visiting www.ncsea.com.

June 12, 2019

CalOES Safety Assessment Program Douglas L. Fell, P.E.

This SAP training course provides engineers, architects, and code-enforcement professionals with the basic skills required to perform safety assessments of structures following disasters.

June 18, 2019

Post-Tensioning Concepts, Repair, Modifications & Evaluation of Existing PT Structures Nate Poen

This presentation will discuss the evolution of typical prestressing steel, anchors, sheathing, and protection systems, the problems they created, repair strategies, and the long-term solutions provided by the new encapsulated post-tensioned systems that are used today.

July 11, 2019

Strength Design of Masonry Walls  The Code & Beyond John Hochwalt, P.E.

The design of masonry walls for in-plane and out-of-plane loads using the strength design provisions of TMS 402-16 will be discussed. Rules of thumb for initial proportioning of masonry walls will be shared, the strength design provisions will be reviewed, and some new and lesser used provisions that can allow for more efficient wall designs will be highlighted.

Courses award 1.5 hours of continuing education after the completion of a quiz. Diamond Review approved in all 50 states.

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61

SEI Update Learning / Networking

Global Practice Guide

Developed by and for the structural engineering community, this Guide highlights areas required for global practice beyond the structural engineer’s domestic field of training –including culture, design and construction, and legal and financial issues. The intent is to raise awareness of topics that can provide assistance whether you are contemplating global practice, considering go-no-go decisions on specific opportunities, and/or formulating project planning.

Technical, business, and professional considerations are highlighted that affect the ability to succeed. The Guide is the manifestation of the SEI Vision for the Future, written by SEI members and produced with funding from the SEI Futures Fund. Access with SEI/ASCE member login at https://collaborate.asce.org/integratedstructures/home.

NEW Free Resources for Members

Connect and access resources via ASCE Collaborate: Integrated Buildings and Structures. Use your SEI/ASCE member login to engage in member discussions, access NEW Global Practice Guide, Grenfell Tower Session from Structures Congress, ASCE 7-16 Overview, Wind, and Seismic Load presentations. https://collaborate.asce.org/integratedstructures/home.

ASCE/SEI 7-16 Wind Loads Presentation June 5 in Chicago

Also presenting the ASCE/SEI standard development process and future of performance-based codes, and promoting opportunities to get involved. Made possible by the SEI Futures Fund in collaboration with the ASCE Foundation. https://bit.ly/2PTGFjI

New Books

Resilience-Based Performance: Next Generation Guidelines for Buildings and Lifeline Standards

Wood Pole Structures for Electrical Transmission Lines: Recommended Practice for Design and Use (MOP 141)

Focuses on the enhancements needed in design and construction of buildings and lifeline systems to support a community’s social stability, economic vitality, and environmental sustainability. Current codebased standards focus on the performance of individual facilities and are out of sync with the resilience needs of the broader community.

This book provides the basis of a new approach of interest to engineers, economists, planners, and government officials.

Provides comprehensive knowledge of the principles and methods for the design and use of wood poles for overhead utility line structures. The use of wood pole structures, properly designed utilizing consistent structural engineering principles, may provide a simple, cost-effective, and more resilient option than some of the other pole materials commonly used. MOP 141 is valuable to engineers involved in utility, electrical, and structural engineering.

SAVE THE DATE STRUCTURAL ENGINEERING INSTITUTE

STRUCTURES CONGRESS 2020 St. Louis, Missouri I April 5-8

Interact with and learn from academic/practice experts on innovative topics: • Blast & Structural Response • Bridge & Transportation Structures • Building • Business & Professional • Career Development

• • • • • •

CALL FOR PROPOSALS Be part of the program - Submit an Abstract or Session Proposal Deadline: June 5, 2019

The Premier Event in Structural Engineering

Education Forensic Natural Disaster Special Structures Nonstructural Research

Students & Young Professionals: Apply for Scholarship to Participate. Learn more www.structurescongress.org 62 STRUCTURE magazine

News of the Structural Engineering Institute of ASCE For best rate, register

by June 7. Registrations not accepted after September 3.

• Dialog on advancing global interoperability and improving collaboration among international teams • Explore financial and economic considerations of large-scale international design/construction projects • Discuss lessons learned on standards of professional competence, building codes, and legal/regulatory • Learn from the experts on: o Large Volume Public Occupancy: Museum of the Future (Dubai), Singapore Sports Hub, Tottenham Hotspur New Stadium (UK) o Unusual Structures – typical codes do not apply: Dubai Frame, London Eye, Vegas High Roller o Performance-Based Design of Tall Buildings: Jeddah Tower, Salesforce Tower (San Francisco, USA), Shanghai Tower

Give to SEI Futures Fund and Make 4x the Impact

His enthusiasm and support for the profession do not stop with sponsoring SEI Structures Congress or encouraging our members to take their place as leaders and innovators. Donate to the SEI Futures Fund through August to support the Future of Structural Engineering, and it will be matched 4 for 1 with a generous gift of up to $40,000 from Ashraf Habibullah, CEO of Computer & Structures, Inc. Make your donation today! www.asce.org/SEIFuturesFund

Thank you

Thank you to the SEI Futures Fund and Donors for investing in the Future of Structural Engineering at Structures Congress in Orlando by supporting 45 Student and Young Professional Scholarship Recipients.

“I am extremely thankful for the opportunity to engage with industry leaders, practicing engineers, fellow researchers, and other students from around the world – the experience has been inspiring and insightful. I am in the process of connecting with other attendees through email and professional online platforms. Also, I am in the process of starting an SEI Grad Student Chapter at the University of Oklahoma, and I hope to have it officially up and running by the fall semester.” Jake Choate, S.M.ASCE

Membership

Vote in SEI Online Election for SEI Board Member by June 30

The SEI Board of Governors consists of two representatives from each of the five SEI Divisions (Business & Professional, Codes & Standards, Global, Local, and Technical Activities), one appointee from ASCE, the SEI President, SEI Past President, and the SEI Director as a nonvoting member. The Division representatives each serve a four-year term.

In accordance with the SEI Bylaws, SEI is conducting an online election for a Technical Activities representative to the Board, term effective October 1. The SEI Technical Activities Division Executive Committee has nominated: Jerome F. Hajjar, Ph.D., P.E., F.SEI, F.ASCE Current SEI members above the grade of Student will receive an email June 1 from Association Voting on how to verify and submit your secure ballot online. Ballots are due online no later than June 30.

SEI Online

SEI News

Read the latest news items at www.asce.org/SEI.

Errata

SEI Standards

Visit www.asce.org/SEIStandards to: • View ASCE 7 development cycle • Submit proposals to revise ASCE 7

SEI on Twitter

Follow us: @ASCE_SEI

SEI Standards Supplements and Errata including ASCE 7. See www.asce.org/SEI-Errata.

If you would like to submit errata, contact Jon Esslinger at jesslinger@asce.org.

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63

CASE in Point Did you know?

CASE has tools to help firms deal with a wide variety of business scenarios. Whether your firm needs to establish new procedures or simply update established programs, CASE has the tools you need!

If your firm needs to update its current Risk Management Program or establish a program within the firm, the following CASE documents will guide employees: 962-H: National Practice Guideline on Project and Business Risk Management Tool 1-1: Create a Culture for Managing Risks and Preventing Claims Tool 2-1: A Risk Evaluation Checklist Tool 2-4: Project Risk Management Plan Tool 3-1: A Risk Management Program Planning Structure

You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

NEW – CASE Guideline and Commentary on ASCE Wind Design Provisions

The purpose of this Guideline is to provide guidance and commentary on the wind provisions of ASCE/SEI 7 and make available a brief overview of the changes from ASCE/SEI 7-05 to ASCE/SEI 7-10, and again from ASCE/SEI 7-10 to ASCE/SEI 7-16.

The most recent revisions to the Standard have restructured the format of its wind design procedures and added step-by-step checklists for each procedure to help clarify how to use its provisions. The Standard is continually updating and editing its procedures based on the latest research, data, and studies.

You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

Donate to the CASE Scholarship Fund!

The ACEC Council of American Structural Engineers (CASE) is currently seeking contributions to help make the structural engineering scholarship program a success. The CASE scholarship, administered by the ACEC College of Fellows, is awarded to a student seeking a Bachelor’s degree, at a minimum, in an ABET-accredited engineering program. Since 2009, the CASE Scholarship program has given $29,000 to help engineering students pave their way to a bright future in structural engineering.

We have all witnessed the stiff competition from other disciplines and professions eager to obtain the best and brightest young talent from a dwindling pool of engineering graduates. One way to enhance the ability of students in pursuing their dreams to become professional engineers is to offer incentives in educational support.

Your monetary support is vital in helping CASE and ACEC increase scholarships to those students who are the future of our industry. All donations toward the program may be eligible for tax deduction, and you don’t have to be an ACEC member to donate! Contact Heather Talbert at htalbert@acec.org to donate.

ACEC’s 2019 Annual Convention and Legislative Summit

May 5-8 saw over a 1,000 ACEC members attend the ACEC Annual Convention in Washington, D.C., meeting with Senators, Congressmen, and Capitol Hill staffers to advocate for major infrastructure legislation in 2019.

More than 800 attended Tuesday night’s black-tie EEA Gala in Washington, D.C., hosted by Emmy award-winning comedian Ross Shafer, where 196 preeminent engineering accomplishments from throughout the nation and the world were honored.

Alaskan Way Viaduct Replacement Program, produced by engineering firm WSP USA, was honored with the 2019 ACEC Grand Conceptor Award. This project replaced one of Seattle’s most earthquake-vulnerable highways with a 1.7-mile underground tunnel containing a state-of-the-art earthquake-resilient double-deck thoroughfare. Located 200 feet beneath downtown, and within a highly active seismic region, the new tunnel has a revolutionary flexible concrete core that combines with its underground location to make it capable of withstanding a 9.0 earthquake.

Pioneering ventilation and fire control systems also make the tunnel one of the safest structures of its type in the world. Additionally, the project eliminates a half-century-old barrier separating downtown from its waterfront to pave the way for nine new acres of public-friendly space.

The new underground tunnel joins the Bayonne Bridge, Staten Island, NY (2018); the SR520 Floating Bridge Replacement, Seattle (2017); and the San Francisco Airport Air Traffic Control Tower, (2016) as recent distinguished ACEC Grand Conceptor Award winners.

ACEC’s Annual Convention also marks the induction of a new ACEC Executive Committee. Jaros, Baum & Bolles Managing Partner, Mitch Simpler took the gavel as 2019-2020 ACEC Chair, succeeding Manish Kothari of Sheladia Associates.

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News of the Council of American Structural Engineers Applying Expertise as an Engineering Expert Witness – Atlanta, GA Wednesday, June 26, 2019, and Thursday, June 27, 2019

Engineers are often asked to serve as expert witnesses in legal proceedings – but only the prepared and prudent engineer should take on these potentially lucrative assignments. If asked, would you be ready to say yes? Developed exclusively for engineers, architects, and surveyors, this unique course will show you how to prepare for and successfully provide expert testimony for discovery, depositions, the witness stand, and related legal proceedings.

Applying Expertise as an Engineering Expert Witness is a focused and engaging 1½-day course that will run you through each step of the qualifications, ramifications, and expectations of serving as an expert witness.

Get Full Program and Registration Details: https://bit.ly/2WSRW6g.

Fresh EJCDC Contracts to Meet Modern Market Demands

EJCDC’s newly released 2018 Construction (C-Series) Documents are a significant modernization, revision, and expansion of the 2013 C-series and now the state-of-the-art in construction contract documents.

The updated edition comprises 25 integrated documents including: • Fundamental contract documents such as the Standard General Conditions, the Small Project agreement, and Supplementary Conditions • Forms for gathering information needed to draft bidding documents • Instructions for bidders and a standard bid form • Bonds including bid, performance, warranty (new for 2018), and payment bonds • Administrative forms, such as change orders and a certificate of substantial completion EJCDC C-700, Standard General Conditions of the Construction Contract has been extensively refreshed and updated, too. The new EJCDC 2018 C-Series also includes expanded and updated “Notes to Users” and “Guidelines for Use” to provide more specific instructions, and it eliminates the need for notary and corporate seals.

You can purchase these and the other EJCDC documents at www.acec.org/bookstore.

Manual for New Consulting Engineers An HR Favorite for New Hires

ACEC’s best-seller, “Can I Borrow Your Watch?” A Beginner’s Guide to Succeeding in a Professional Consulting Organization offers new engineers a head start in the business of professional consulting.

This essential guide is tailored to the unique needs of engineering firms, and the skills and experiences rookie consultants need to be successful in a large organization, including: • Proposal Preparation • Financial Management • Client Relationships

• Project Management • Staff Management

With over 140 pages of consulting expertise, this resource is the perfect addition to any new staffer’s welcome pack or in-house orientation. It can even be a useful resource for more seasoned engineers looking to refine their skills.

To order this book, go to www.acec.org/bookstore. Bulk ordering is available, for more information contact Maureen Brown (mbrown@acec.org).

Follow ACEC Coalitions on Twitter – @ACECCoalitions.

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business PRACTICES From Peer to Manager

By Jennifer Anderson

H

ave you been recently promoted to a leadership position? Congratulation on a great accomplishment! With a new leadership position comes greater responsibility and the opportunity to grow in ways that you cannot even imagine. This article covers five key points to navigate going from “peer” to “manager” to help you prepare for the transition.

Personal Brand

Social Media If you are friends on social media with your peers, you will likely want to “unfriend” them when you become their manager.

The main reason is to keep a professional distance. As a friend on Facebook, you will have access to their photos and posts, which exposes you to much personal information that could put you in a compromised position. For example, what if your work friend

‘‘

Team Member Interruptions As a peer, your “door” was likely always open for anyone to speak with you. When you become a manager, you will need to learn the power of protecting your schedule. Otherwise, you will be continuously interrupted and hard-pressed to get all of your work done. Certainly, you should make yourself available for an urgent situation but learning how to manage the interruptions will reduce stress (yes, your stress level is going to increase with the promotion).

One-on-One Time

‘‘

The first thing you should always keep in mind is your personal brand. Your brand does not change now that you are in a leadership position. What changed is your title. You are still you. You are still composed of the same set of skills, personality traits, and habits.

Yes, you will grow and learn new managerial skills, but your characteristics will remain with you for the rest of your life. Make sure to be clear about your personal brand (also known as your legacy or your mark on the world) so that you will know what kind of manager you want to be.

One of the most powerful ways you can connect with your direct reports is through regular, individual, one-on-one meetings.

These are not project status update meet-

With a new leadership position comes greater responsibility and the opportunity to grow in ways that you cannot even imagine.

posted photos of drinking at a bar while on a business trip, and then you know that they drove the company car home from the bar? Depending on your company’s social media policy, you may now have to fire your work friend. As his manager, you know that he did something illegal and against company policy.

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ings; one-on-one meetings are for your direct reports to get support and guidance from you about their career, personal and professional goals, and development at the company.

One-on-one meetings are time well spent.

A powerful connection happens during a closed-door, private conversation. One key thing about one-on-one meetings is to be

consistent … these meetings should never be rescheduled. Your direct reports will greatly appreciate your efforts to prioritize them.

Don’t Forget It is likely that there was a time when your manager did something that drove you nuts. At that moment, you might have said something to yourself like, “When I’m the manager, I will NEVER do that.” Well, now you are the manager! Do not do those things that will drive people nuts. Take some time to think about what kind of manager you do not want to be and write out what kind of manager you do want to be. Share that list with your direct reports and ask them to hold you accountable. Being vulnerable like this will help you to stay connected to your work peers in a way that is lasting to both of your professional careers.

Ultimately, going from peer to manager has many nuances that will be specific to your situation, but do not forget that it is YOUR career…make sure that you are growing and developing into your version of a leader. Be the manager that you’ve always wanted to work for!■ Born into a family of engineers but focusing on the people side of engineering, Jen Anderson has over 21 years of helping leaders build stronger careers for themselves and their teams. (www.CareerCoachJen.com)

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THE ULTIMATE RESOURCE IN TUBULAR PRODUCTS

IS HERE Independence Tube Corporation, Southland Tube, and Republic Conduit are now Nucor Tubular Products. As we come together as part of Nucor, North America’s leading steel company, we remain dedicated to working with you, our customer.

As a result, our HSS line now boasts a wider product range. But one thing hasn’t changed, our quality and service continues to be among the best in the industry. We pioneered on-line ordering with our 24/7 customer secure portal and our on time rolling schedule is considered to be second to none among our customers.

As part of our tubular family, Republic Conduit continues to offer its electrical conduit products designed to reduce installation costs and jobsite delays. This winning combination of products and innovation continues to support the reason why we have been so successful: working together and dedicated to providing our customers with the best products and services in the industry.

Our locations include: Birmingham, AL; Cedar Springs, GA; Chicago, IL; Decatur, AL; Louisville, KY; Marseilles, IL; and Trinity, AL.

NTP Grades include: • ASTM A500 • ASTM A252 • ASTM A1085 • ASTM A513 • A53 grade B Type E ERW • ASTM A135 and ASTM A795 Sprinkler Pipe

HSS Sizes include: Squares: ½" x 16" gauge through 12" x .625" wall Rectangles: 1 ½" x 1" x 16 gauge through 16" x .625" wall Rounds: .840" OD x .109" wall through 16" OD x .688" wall

Learn more at www.nucortubular.com