STRUCTURE magazine - March 2020 by structuremag

STRUCTURE MARCH 2020

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WIND/SEISMIC

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Pile Structural Capacity Seismic Loading Protocols High Capacity Timber Connection

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WASHINGTON STATE FERRIES TIMBER TRESTLE SEISMIC HAZARD ANALYSIS AND VASHON TRESTLE SEISMIC REHABILITATION

Erin Conaway, P.E. AISC, Littleton, CO Linda M. Kaplan, P.E. Pennoni, Pittsburgh, PA Charles “Chuck” F. King, P.E. Urban Engineers of New York, New York, NY Emily B. Lorenz, P.E. Chicago, IL

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Jason McCool, P.E. Robbins Engineering Consultants, Little Rock, AR Brian W. Miller Davis, CA Evans Mountzouris, P.E. The DiSalvo Engineering Group, Danbury, CT John “Buddy” Showalter, P.E. International Code Council, Washington, DC Eytan Solomon, P.E., LEED AP Silman, New York, NY Jeannette M. Torrents, P.E., S.E., LEED AP JVA, Inc., Boulder, CO STRUCTURE ® magazine (ISSN 1536 4283) is published monthly by The National Council of Structural Engineers Associations (a nonprofit Association), 20 N. Wacker Drive, Suite 750, Chicago, IL 60606 312.649.4600. Application to Mail at Periodicals Postage Prices is Pending at Chicago, IL and additional mailing offices. STRUCTURE magazine, Volume 27, Number 3, © 2020 by The National Council of Structural Engineers Associations, all rights reserved. Subscription services, back issues and subscription information tel: 312-649-4600, or write to STRUCTURE magazine Circulation, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606.The publication is distributed to members of The National Council of Structural Engineers Associations through a resolution to its bylaws, and to members of CASE and SEI paid by each organization as nominal price subscription for its members as a benefit of their membership. Yearly Subscription in USA $75; $40 For Students; Canada $90; $60 for Canadian Students; Foreign $135, $90 for foreign students. Editorial Office: Send editorial mail to: STRUCTURE magazine, Attn: Editorial, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606. POSTMASTER: Send Address changes to STRUCTURE magazine, 20 N. Wacker Drive, Suite 750, Chicago, IL 60606. STRUCTURE is a registered trademark of the National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

Contents M ARCH 2020

Cover Feature

26 NAPA COUNTY HISTORIC COURTHOUSE – PART 3 By Brett Shields, P.E., Luke Wilson, S.E., and Kevin Zucco, S.E.

Part 3 discusses repairs and an overview of the applicability of Fabric-Reinforced Cementitious Matrix (FRCM). FRCM is more homogeneous with existing brick stiffness and mechanical properties compared to epoxy-based overlays and allows historic brick to breathe.

30 ANATOMY OF A HIGH CAPACITY TIMBER CONNECTION By D. Scott Nyseth, P.E., S.E.

An 18-x18-inch timber was designed as the bottom chord of a pedestrian bridge truss. For the truss tension splice connection, choosing the optimal number of interior knife plates and a dowel diameter that effectively utilized the entire width of the timber allowed for higher capacity connections.

Columns and Departments 7

Editorial Going Global

By Anne Ellis, P.E.

Structural Rehabilitation

By Kirk Grundahl, P.E.

The IEBC’s Roof Diaphragm Evaluation Requirements 43

Structural Testing Lab Test Confidential: Seismic Loading Protocols By Matthew S. Speicher, Ph.D., and Bruce F. Maison, P.E., S.E.

Structural Systems Structure Design Considerations for Building Enclosures By Matthew L. Wagner, S.E., and Patrick Olechno, P.E.

Spotlight 181 Fremont By Ibbi Almufti, S.E., and Nate Warner, P.E.

Structural Components Pile Structural Capacity By H. Y. Ng, P.E., C.Eng

Historic Structures Albion Bridge Collapse By Frank Griggs, Jr., D.Eng., P.E.

By Dale Statler, P.E., and Jerry Maly, P.E.

Building Blocks Pre-Manufactured Wood Trusses

Structural Forum FEMA P-807 for Soft-Story Retrofits By Bruce F. Maison, P.E., S.E.

In Every Issue 4 40 44 46 48

Advertiser Index Resource Guide – Software Updates NCSEA News SEI Update CASE in Point

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, the Publisher, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions. MARCH 2020

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EDITORIAL Going Global

Introducing the SEI/ASCE Global Practice Guide By Anne Ellis, P.E., FACI, F.ASCE, NAC

id you know in some countries the workweek is Saturday – and addresses culture, design and construction, and legal and financial Thursday? That some countries use decimal commas where issues. The intent is to raise awareness of and seed inquiry into specific others use decimal periods when writing numbers of 4 places or topics from those that can provide appropriate assistance. The Guide more? That some countries have neither national, regional nor local will prove valuable whether you are contemplating global practice, building codes nor standards? When working outside the U.S., local considering go-no-go decisions on specific opportunities, and/or forpractices, different from our own, can wreak havoc on structural mulating project plans. An extensive list of topics is covered (see box). engineering projects if not adequately anticipated and addressed. In The Guide also provides country-specific information and examples. consideration of such, some may ask, why work outside of the U.S.? The Guide does not address considerations related to the establishStructural Engineering increasingly is becoming a profession that ment of an in-country office. crosses national borders. Advances in technology, increasing interdeStructural Engineers thriving in global practice recognize there is pendent economies, and emerging market investment in infrastructure no single way of doing things when working globally. They identify and real estate enable structural engineering practice opportunities out- and plan for the local differences. They recognize that every client, side of the U.S. These opportunities country, local society, and workforce are significant and allow structural is unique. These differences are anticiTopics Covered engineers to participate in projects pated, accommodated, and adequately • Team structure • Cultural intelligence of grand scale and challenge. These resourced. Legal, financial, and human • Standard of care and • Cultural dynamics ventures allow structural engineers resource professionals with global experother legal regimes • Communication to share and transfer their high-value tise are essential to informing and • Rule of law • Multicultural teams and knowledge. These global projects guiding the structural engineer’s success. • Registration workforce allow structural engineers to work Working outside your home country • Anticorruption • Business etiquette with different codes and standards can be a very attractive proposition. • Local tender rules • Codes and standards and be exposed to different materials, Many engineers are motivated to make • Contracts • Sustainability rating systems means, and methods. When consida difference where the need is greater. • Insurance • Resilience ering projects outside the U.S., the Some structural engineers may be • Professional liability • Construction practice, opportunities must be balanced by attracted by the idea of traveling to and including means and methods • Intellectual property and the multiplicity of risks, including even living somewhere different. The conﬁdential information • Labor: Skills and languages government instability, even regime idea of new technical challenges may • Export controls • Safety: Personal, liability, collapse, that may result in people attract others. And, for many, it may • Cybersecurity and responsibility and assets stranded for extensive peribe something never considered but an • Employment • Metric system ods and with little recourse. opportunity offered by their employer, • Cost and pricing • Project delivery types Successful global practice requires an opportunity that may enhance the • Currency considerations • Expectation of deliverables structural engineers to embrace chance of promotion, even fuel a career. • Taxation • Infrastructure: Technology, skills, traits, know-how, and awareWhatever the attraction or motivation, • Local accounting rules phone, power, and ness not only technically but also in opportunities are great for those who • Banking transportation areas outside their field of techniare willing to go global. This Guide cal training. The extension of skills, is intended to inform those structural traits, and know-how varies greatly depending on the client whose engineers of the basic considerations imperative for success. activities you are supporting. U.S.-based clients operating overseas and Are you interested in advancing structural engineering global practice? utilizing the codes, standards, specifications, and contract constructs If so, consider joining an SEI GAD committee. GAD is responsible common in the United States reduce many, but not all, of the risks of for increasing SEI members’ awareness of global issues that impact working outside the country. Compare that to foreign-government our profession and facilitating the development of skills that will allow clients who may lack legal and technical governance frameworks – SEI members to thrive in the world market. The Global Credential including governing codes, standards, specifications, and contracts Committee is charged with enabling global structural engineering – that those practicing in the U.S. take for granted. credentials; the Inter-organizational Collaboration Committee is To aid structural engineers in successful global practice, the Structural charged with establishing partnerships with other organizations to Engineering Institute (SEI) of ASCE developed the Global Practice mutually advance efforts of benefit to the profession. Learn more at Guide (the Guide), available to SEI/ASCE members for free download. www.asce.org/SEIGlobal. The Guide is a manifestation of the SEI Vision for the Future, a work Access and share your feedback on the Global Practice product of the SEI Global Activities Division (GAD), written by SEI Guide at https://bit.ly/2RFwcuM (member login required).■ members and produced with funding from the SEI Futures Fund. Anne Ellis is Executive Director of the Charles Pankow Foundation, advancing This SEI-led Guide also will prove valuable to those practicing civil better ways to design and build. She served as the American Concrete as well as other engineering disciplines. Institute President (2013-2014) and held enterprise positions at AECOM The Guide highlights areas requisite for global practice that are (2008-2016), advancing the global impact of both organizations. beyond the U.S.-based structural engineer’s domestic field of training STRUCTURE magazine

M A R C H 2 02 0

structural REHABILITATION The IEBC’s Roof Diaphragm Evaluation Requirements When Reroofing Requires a Lateral Analysis By Dale Statler, P.E., and Jerry Maly, P.E.

ince its inception in 2003, the International Existing Building Code (IEBC) has contained a provision that triggers the evaluation and possible retrofit of roof diaphragms when certain buildings are reroofed. This provision has gradually evolved within the Work Area Method, unnoticed by some practitioners and readily avoided by others by reverting to the International Building Code (IBC) Chapter 34, Existing Buildings and Structures, or by using the Prescriptive or Performance Methods within the IEBC. However, Chapter 34 was eliminated from the 2015 IBC and the subject provision metastasized into the Prescriptive Method in the name of consistency. So today, engineers, architects, and owners are forced to contend with it (except in the narrow instances where the Performance Method is applicable) and, for those affected, the consequences can be unduly burdensome. This article recounts the origin and evolution of the provision since its introduction in 2003, discusses fundamental flaws in its requirements, and argues for the limitation of its applicability to either: 1) repairs that can be made to correct visible deterioration and/or deficiencies that are readily observed and remedied in the normal course of a roof replacement, or 2) specific geographic regions or building types known to have extraordinary roof diaphragm vulnerabilities.

Origin and Evolution In the first edition of the IEBC (2003), a structural provision in Chapter 5, Alterations – Level 1, of the Work Area Method stated the following: 507.3 Roof Diaphragm. Where roofing materials are removed from more than 50 percent of the roof diaphragm of a building or section of a building where the roof diaphragm is a part of the main wind force-resisting system, the integrity of the roof diaphragm shall be evaluated and if found deficient because of insufficient or deteriorated connections, such connections shall be provided or replaced. To the authors’ knowledge, nothing similar to this provision existed in any of the three model codes or other documents that served as the primary basis for the first edition of the IEBC. Furthermore, nowhere in the June 2001 Working Draft of the 2003 IEBC, prepared by the 2003 IEBC Drafting Committee, was there any mention of structural evaluations and/or upgrades to roof diaphragms associated with reroofing. As such, it was surprising that this provision appeared in the August 2001 Final Draft of the 2003 IEBC, also prepared by the 2003 IEBC Drafting Committee. It was subsequently learned from International Code Council (ICC) Technical Services that, based on recollections of certain ICC staff, the drafting committee reportedly had “concerns about the working draft and the lack of protection for high wind, and the focus was on the connections because they were often the cause of failures in high winds.” Unfortunately, it appears that there are no meeting minutes or other written records 8 STRUCTURE magazine

Figure 1. Wood diaphragm connections and shear walls not visible from the top surface of the diaphragm.

that provide elaboration or documentation regarding these alleged failures, including 1) locations, 2) wind speeds, 3) types of storms, e.g., thunderstorm, tornado, hurricane, chinook, 4) diaphragm materials, e.g., wood, steel, concrete, gypsum, etc., 5) connections of concern, or 6) the extent to which diaphragms were actually affected. During their service lives, most buildings will be reroofed on multiple occasions, with the life of conventional roofing systems ranging from about 20 to 40 years. As elaborated in the IEBC Commentary since 2003, the provision intends to take advantage of this opportunity to observe and address potential problems that are otherwise obstructed from view. The provision applied only to diaphragm deficiencies from “insufficient and deteriorated connections,” which apparently were the original drafting committee’s focus. Any more extensive analytical evaluation would require an abundance of detailed information, including the locations and lengths of shear walls or frames and numerous connection details that are not necessarily observable from the top surface (Figure 1). As such, this provision appears to have been originally intended to identify and address obviously deficient or deteriorated connections based on a visual evaluation of a diaphragm’s top surface only; deficiencies or deterioration beyond this could not be observed or easily remedied in the relatively short period available between removal and replacement of a roofing system. However, as outlined below, this intent has been lost in subsequent revisions to the IEBC. Several modifications were made to the provision in the 2009 IEBC (606.3.2). One of these changes limited its applicability to “high-wind regions,” defined as areas where the basic wind speed was greater than 90 mph (the baseline design speed for non-coastal areas of the U.S.) or areas that were within special wind regions as defined in Section 1609 of the IBC. A second change mandated that the diaphragm evaluation be performed using design wind loads required by the IBC for new buildings, and stated explicitly that wind uplift was to be included in the analysis. Where diaphragms and/or their connections in their current condition were unable to resist these loads, strengthening or replacement was required. The 2012 IEBC (706.3.2) added clarification on the diaphragm connections that were to be addressed in the required evaluation,

explicitly including connections of the roof diaphragm to roof framing and roof-to-wall connections. This edition also reduced the design wind load criteria to 75 percent of that required for new buildings. Changes in the 2015 IEBC (707.3.2) consisted of updating the design wind speed consistent with the transition to ultimate loads in ASCE 7-10, Minimum Design Loads for Buildings and Other Structures, i.e., 90 mph became 115 mph and adding a virtually identical provision to the Prescriptive Method (403.8). Also in 2015, Chapter 34, Existing Buildings and Structures, was removed from the IBC, leaving regulation of existing buildings solely up to the IEBC. These requirements were unchanged in both the Work Area Method (706.3.2) and the Prescriptive Method (503.12) of the 2018 IEBC. The current provision in 706.3.2 reads as follows: 706.3.2 Roof diaphragms resisting wind loads in high-wind regions. Where roofing materials are removed from more than 50 percent of the roof diaphragm or section of a building located where the ultimate design wind speed, Vult , determined in accordance with Figure 1609.3(1) of the International Building Code, is greater than 115 mph (51 m/s) or in a special wind region, as defined in Section 1609 of the International Building Code, roof diaphragms, connections of the roof diaphragm to roof framing members, and roof-to-wall connections shall be evaluated for the wind loads specified in the International Building Code, including wind uplift. If the diaphragms and connections in their current condition are not capable of resisting 75 percent of those wind loads, they shall be replaced or strengthened in accordance with the loads specified in the International Building Code.

See Figure 2 (page 16 ) for the reproduction of 2018 IBC Figure 1609.3(1) with areas conforming to the IEBC definition of “highwind regions” highlighted.

Ramifications A diaphragm evaluation strictly conforming to the current provision and its stated intent would ostensibly involve the following: 1) removal of all existing roofing down to the structural diaphragm for observation and, except where drawings are available and sufficiently detailed, collection of data to support the structural analysis; 2) engineering calculations, which cannot be performed extemporaneously in the field, evaluating the diaphragm and connection strengths to resist the prescribed design wind forces; 3) installation of temporary protection for the roof in anticipation of the possibility of resulting structural retrofit work; 4) both demobilization and subsequent remobilization of the roofing crew; 5) design and permitting of any necessary structural retrofits, 6) potentially hiring a subcontractor capable of installing the necessary structural retrofits, and 7) resuming installation of the replacement roofing system. The authors suspect that such a sequence of events has rarely, if ever, occurred. More likely, the provision has been avoided by referencing an alternate chapter in the adopted code(s), the design professional, contractor, and/or building official never knew that the provision applied, or it was ignored. However, for conscientious design professionals working under the authority of attentive building officials, the only rational option has become to consult the construction drawings for the critical details well in advance of the work. In the absence of comprehensive construction documents, which is frequently the case,

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the evaluators are compelled to document the structure themselves, making pre-construction destructive openings in the roofing, inspecting below-deck conditions from the interior, and then analyzing and designing any necessary structural retrofits to be bid and permitted in conjunction with the roofing contract. This work can result in significant increases in costs for the routine exercise of reroofing.

Flawed Foundation Wind can and does cause structural damage to buildings due to shortcomings in the original codes, problems with the design, construction defects, accumulated deterioration, or some combination of these factors. However, while model codes, as well as design and construction practices, have generally improved over time, the safety and sufficiency of existing structures are only rarely revisited unless significant damage has occurred or if a proposed structural alteration or occupancy change triggers compliance with the provisions for new structures. One such instance is presented next to the subject diaphragm reroofing passages in the IEBC: the requirement that unreinforced masonry (URM) bearing wall parapets be braced when reroofing buildings in high seismic regions. This provision addresses an exceptional hazard demonstrated by repeated poor performance (in many cases, even in events much less severe than design), arguably justifying the imposition of costs on a building owner to abate a significant latent danger to the public. To justify the high costs of retroactive diaphragm evaluations and upgrades, the authors believe there should be a commensurate extraordinary risk from wind-related diaphragm vulnerabilities. Such vulnerabilities may be regional, such as the URM parapet provision that only applies in Seismic Design Categories D through F. Likewise, diaphragm wind upgrades should be limited to regions or building types where extraordinary vulnerabilities have been observed. Coastal 10 STRUCTURE magazine

hurricane regions may be in this category, but the authors are not aware of any rigorous study that substantiates the existence or extent of any such extraordinary hazard associated with roof diaphragm performance. However, anecdotal evidence does suggest that buildings do collapse with some frequency in hurricane winds after roof diaphragm integrity is lost. The authors reside and practice structural engineering along Coloradoâ&#x20AC;&#x2122;s Front Range in a special wind region where basic wind speeds range from 115 to 225 mph. Based upon their knowledge and experience investigating structural failures in this extraordinary wind climate, they are unaware of any remarkable incidence of diaphragm failures from high winds. Similarly, the results of an informal survey conducted among professional members of the Structural Engineers Association of Colorado in 2017 indicated no prevailing evidence of diaphragm vulnerabilities in Coloradoâ&#x20AC;&#x2122;s special wind region.

2021 IEBC and Beyond ICC has approved modifications to the 2018 diaphragm provisions for inclusion in the 2021 IEBC. The revised provision in 706.3.2 will read as follows: 706.3.2 Roof diaphragms resisting wind loads in high-wind regions. Where roofing materials are removed from more than 50 percent of the roof diaphragm or section of a building located where the ultimate design wind speed, Vult, determined in accordance with Figure 1609.3(1) of the International Building Code, is greater than 130 mph (58 m/s), roof diaphragms, connections of the roof diaphragm to roof framing members, and roof-to-wall connections shall be evaluated for the wind loads specified in the International Building Code, including wind uplift. If the diaphragms and connections in their current condition are not capable of resisting 75 percent of those wind loads, they shall be replaced or strengthened in accordance with the loads specified in the International Building Code.

Exception: Buildings that have been designed to comply with to only hurricane-prone regions would limit the provision’s the wind load provisions in ASCE 7-88 or later editions. scope to suspected areas of vulnerability that are threatened The changes include: 1) an increase in the threshold wind speed from by extraordinary winds.■ 115 to 130 mph (the wind speed above which glazed openings must be protected from impact in hurricane-prone regions), 2) elimination Dale Statler is a Senior Associate in the Denver, Colorado office of Wiss, of any reference to special wind regions, and 3) an exception to the Janney, Elstner Associates, Inc. Mr. Statler is an active member of the Existing provision when the building under consideration has been designed Structures Committee of the Structural Engineers Association of Colorado. to comply with what are judged to be comprehensive modern wind INFO Jerry Maly is a Principal in the Denver, Colorado SPECSoffice of Wiss, Janney, load requirements. FileElstner Name: Associates, 20-1246 Structure Mag_March_System Flat Size: Inc. Mr. Maly is aSolutions past president of the Structural The authors worked to develop and promote the acceptance of these Finished Size: 5” ×Existing 7.5” XXXX MKT: 20-1246 of Colorado (SEAC), a member PR:Engineers Association of the changes; however, in our opinion, they do not go far enough. If all the Designer: Georgina Morra Email: gmorra@mapei.com Bleed: Yes Amount: .125” 114 4 E. Newpor t Center Dr. Buildings Subcommittee of NCSEA’s Code Advisory Committee, and a proposed changes had been adopted, diaphragm Colors: 4/0 Date: February 6, 2020 12:20 PM D e e r f i e l d evaluations B e a c h , F L 3 3in 4 4 the 2 member of the Existing Structures Committee of SEAC. 2021 IEBC would be triggered only for buildings N O T E : C O L located O R S V I E W Ein D OhurricaneN - S C R E E N A R E I N T E N D E D F O R V I S U A L R E F E R E N C E O N L Y A N D M A Y N O T M A T C H T H E F I N A L P R I N T E D P R O D U C T. prone regions where the ultimate design wind speed exceeds 130 mph. The authors believe this is important because, to their knowledge, there is no historical evidence substantiating the existence of any extraordinary diaphragm vulnerabilities to wind outside of hurricane-prone areas. Coastal wind regions differ significantly System solutions for bridge restoration from those farther inland in the relationship between frequency and severity of high winds. Design for coastal regions is driven predominantly by extreme random events in an otherwise unexceptional wind Overhead Repair Solutions climate. Compare this to downslope chinook winds along the Colorado Front Range driven by the weather phenomenon of air rising and falling over the Column Repair Solutions Rocky Mountains, in which foothills communities experience high winds on a regular basis. Such winds are neither unusual nor unexpected, and local design and construction practices have necessarily evolved to keep buildings upright with their roofs intact. The relatively high inland frequency of such winds leaves a substantially smaller margin for deficiencies to remain undetected, as may have happened historically on the coasts. Bridge Deck Solutions The requirement that a building undergoes a diaphragm evaluation, involving a significant investigative and analytical effort by an engineer with the possibility of costly structural upgrades, is an extraordinary burden that should only MAPEI offers a full spectrum of products for concrete be justifiable based on a commensurately restoration, below-grade waterproofing and structural extraordinary hazard. Otherwise, it is strengthening. Globally, MAPEI’s system solutions have been logical, appropriate, and consistent with utilized for bridges, highways, parking garages, stadiums, longstanding engineering practice to let buildings and other structures. grandfathered structures stand unaffected by the increasingly complex regulations Visit www.mapei.us for details on all MAPEI products. governing new structures. It is unreasonable to attempt to keep the entire building stock up to date with model codes as they continue to evolve. Retroactive upgrades are an appropriate tool when the costs of inaction definitively outweigh the costs of action. But that burden should be limited to where there is sufficient evidence of major structural concern. Tying the trigger

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structural COMPONENTS Pile Structural Capacity A Comparison of Three Design Codes By H.Y. Ng, MSc, P.E. , C.Eng, MIStructE

oundations for a building are required to support the structure so that it remains safe and functional, and so that they meet serviceability limits. Buildings with excessive cracking and movement (e.g. total settlement, differential settlement, tilt, etc.) are alarming and not acceptable. When column loads are heavy, as in high-rise buildings for example, large piles or groups of piles are usually required to carry the loads down to a competent stratum. Bored piles, also referred to as augercast piles, are large-diameter, cast-in-place concrete piles, usually lightly reinforced by steel reinforcing bars (0.5%) under gravity loads. Circular boreholes are drilled in the ground before placing concrete into the borehole. The reinforcing steel can be set in the holes before the concrete is placed or “wet-set” before the concrete is allowed to set. Common bored-pile sizes range from 30 inches to 60 inches (750 mm to 1,500mm) in increments of 4 inches (100mm). This article compares the structural capacity calculation of bored piles in compression using three design codes: ACI-318, Building Code Requirements for Structural Concrete and Commentary; the Eurocode (EC2); and CP4:2003 (Singapore CP4).

Design Codes There are generally two types of design methods: allowable stress design (ASD) and load and resistance factor design (LRFD). In many parts of the world, ASD and LRFD are known as working stress (permissible stress) design and limit-state design, respectively. ASD compares capacities derived from the allowable stress (factored down from ultimate) against the service loads without any load factors, while limit-state design has factors for loadings and partial factors for materials (Table 1). In the design of piles, both the geotechnical (soil strength) capacity and structural (material strength) capacity must be checked. For an economical design, the structural capacity and geotechnical capacity should be as close as possible.

Geotechnical Design The geotechnical capacity of piles can be calculated based on site investigation data. For piles founded in soil, unit shaft friction and end-bearing are summarized in Table 2. In some parts of the world, where there is prior pile design experience in similar ground conditions, it is possible to correlate the shaft friction and end-bearing to standard penetration test (SPT) N values. For example, bored piles in a specific locality may have the values shown in Table 3. (Example: A soil with SPT value of 100 may have unit shaft friction of 3x100 = 300 kPa = 44 psi) In design, it is often prudent to impose an upper bound on shaft friction and end-bearing as a safety precaution to avoid using values that are not achievable or come with excessive settlement. Designers need to recognize that end-bearing requires much more pile settlement to mobilize as compared to shaft friction. The condition of the pile toe is difficult to ascertain. For these reasons, 14 STRUCTURE magazine

Table 1. Codes and design method.

Code

Type of Design

ACI-318

LRFD (ASD is also allowed)

EC2

Limit state design

CP4

Allowable stress design

designers may want to exercise extra caution when selecting end bearing values in design. Codes typically impose a higher factor of safety for end-bearing as compared to shaft friction. The different partial safety factors applied to the calculated shaft friction and end bearing may also depend on conditions such as whether load tests were carried out and the type of loading (compression or tension). To illustrate, Table 4 shows the different factors used in the different codes. Designers need to be mindful of the different factors that are required on shaft friction and end bearing, and the appropriate corresponding factors to be applied to the loadings, depending on the code and type of design adopted. Due to the uncertainties of pile design and construction, load tests are desirable to verify that piles can meet settlement criteria under the design loads. For example, the governing code may specify that, when a test pile is loaded to 1.5 times the unfactored column load, the pile settlement should not exceed 0.6 inches (15 mm) (from CP4).

Structural Design Bored piles are usually designed to carry compression loads, similar to columns in a building. The main difference is that columns are cast above the ground while piles are cast underground. Other differences are summarized in Table 5, page 16. Because of these differences, the structural capacity of a pile determined from code equations usually is lower than column capacity of a similar size and reinforcement. Table 2. Unit shaft friction and end bearing.

Total Stress (Clay) Effective Stress (Sand) Unit shaft friction

αcu

βσv'

Unit end bearing

9cu

Nqσv'

cu = undrained shear strength σv' = average vertical effective stress α = coefficient to reduce shear strength (usually 0.5 to 1.0, depending on the strength of clay) β = coefficient related to interface friction between pile and soil and Ks which is the ratio of horizontal stress to vertical stress Nq = bearing capacity factor based on soil friction angle Table 3. Example of shaft friction and end bearing based on SPT.

Coefficient

Maximum (kPa)

Unit shaft friction

1.5 – 3

150 – 300

Unit end bearing

40 – 120

10,000

Under gravity loads, bored piles in compression can be nominally reinforced (using minimum reinforcement). Usually, it is more economical to use a larger pile with nominal reinforcements compared to a smaller pile that is heavily reinforced. However, smaller diameter piles with heavier reinforcements may be adopted in certain situations such as space constraints or low headroom. For comparison in this article, only nominally reinforced bored piles founded in soil are discussed.

Structural Capacity Based on ACI-318 In the U.S., ACI-318 is a commonly used standard for the design of reinforced concrete structures. ACI-318 is primarily an LRFD code, but ASD is also allowed. Bored piles are known as drilled shafts, drilled piers, or auger-cast piles. The structural design of drilled piers is similar to a beam-column. However, in most practical cases, the design can be simplified to a short column by assuming the bending moment is negligible (gravity loads only) and the pier is laterally restrained, unless in the case of very soft soil (for example, less than 1.5 psi (10 kPa) shear strength). U.S. practice also includes seismic considerations in certain areas, which may require special detailing requirements, such as spiral hoops (for added shear strength) and more stringent reinforcement spacing and limits. For column design, ACI-318 provides the following well-established equation for column capacity: φPn = φ(0.85f´c Ac + fyAst) ACI’s strength reduction factor, φ, is an overall factor to reduce nominal strength, similar to the partial safety factor for materials in the Eurocodes (e.g., 1.5 for concrete).

Table 4. Partial factors on shaft friction and end bearing.

Shaft Friction Safety Factor

End Bearing Safety Factor

1.5 to 5

EC2*

1.68

2.04

CP4

1.5 to 2

ACI-336^

^ LRFD is denoted as strength design method, ASD is denoted as alternate design method * The 1.68 and 2.04 factors for EC2 were calculated based on Design Approach 1 – Combination 2 (usually governing for pile design) where there is a 1.3 factor on live loads and 1.2 model factor applied to 1.4 and 1.7 for shaft and base, respectively. These factors assume that appropriate load tests were carried out.

A strength reduction factor, φ, of 0.75 is used for spirally reinforced columns and 0.65 for tied columns. There is a further reduction factor of 0.85 and 0.8 for spiral and tied columns, respectively, to account for eccentricities. Ignoring the contribution of steel (for a nominally reinforced pile), the ultimate capacity for a tied column is: φPn = (0.65)(0.8)(0.85)fckAc = 0.442 fckAc = 0.35 fcuAc This gives a working load of 0.25fcuAc after dividing by a combined load factor of 1.4. Note that ACI-336.3R for drilled piers specifies load factors of 1.4 and 1.7 for dead load and live load, respectively, whereas ACI-318 specifies load factors of 1.2 and 1.6. The difference in load factors for drilled piers and columns suggest that underground concrete is more uncertain and requires a higher factor of safety compared to concrete columns in a superstructure. continued on next page

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Table 5. Difference between a column and a bored pile.

Columns

Bored Piles

Superstructure – above ground (better quality control)

Sub-structure – below ground (harder to control quality)

Unrestrained between adjacent story levels

Supported by soil along the shaft (unless soil is very soft)

Can be easily inspected after construction

Cannot be easily inspected after construction

Usually rectangular with heavy reinforcements

Circular with light reinforcements

Higher concrete strength, e.g., fcu = 5,800 psi (40 MPa)

Lower concrete strength, e.g., fcu = 4,350 psi (30 MPa)

Less concrete cover

Greater concrete cover

(fck = 0.8fcu, where fck = fc´ = cylinder strength and fcu = cube strength)

The U.S. model building code is the International Building Code (IBC). In IBC (allowable stress design), the allowable stress in concrete is 0.33fc´ (0.26fcu). This is in line with a general rule of thumb that design stress is a third of material strength for piling.

Structural Capacity Based on CP4 For rock-socketed piles with full-length reinforcement, using a short column formula, CP4 states that the ultimate structural capacity is given by the sum of stress multiplied by area for both concrete and steel components: Pu = 0.4fcuAc + 0.75fyAs where fcu and fy are concrete and steel strength, respectively, and A is area. To derive the working load, using a minimum safety factor of two, the equation becomes: Pu = 0.2fcuAc + 0.375fyAs In CP4, structural capacity (working stress) of nominally reinforced bored piles is calculated using 0.25fcu (ultimate cube strength) but limited to a maximum of 1,088 psi (7.5 MPa). Some engineers are tempted to view 0.25 as a “safety factor” for structural capacity using a permissible stress perspective, giving a false sense of safety. However, this is not strictly correct because the 0.25 is a value obtained after accounting for several aspects of cube strength which are different compared to actual concrete cast-in piles. The derivation of 0.25 was strongly influenced by BS8110 (or local Singapore CP65), which was the corresponding reinforced concrete design code used in conjunction with CP4. When a column is loaded to failure in compression, the ultimate capacity is the sum of concrete and steel components, and it is given by an empirical formula: N = 0.67fcuAc + fyAs (Note that 0.67fcu = 0.85fck, using fck = 0.8fcu) where fck = fć = cylinder strength and fcu = cube strength. This is the maximum load, independent of creep and shrinkage effects. The 0.67 factor applied to cube compression strength of concrete is to account for differences such as size (actual structural element is much larger than cube), boundary conditions (actual building load on column versus loading using compression testing machine), rate of loading (much faster rate in cube test), and quality of compaction (cube test is properly compacted). An additional “partial safety factor” of 1.5 needs to be applied for design against ultimate collapse (note that this factor is not meant

to bring it down to working load capacity). Because bored piles are lightly reinforced, the contribution from steel can be ignored. This means the ultimate pile structural capacity reduces to: N = 0.45fcuAc The coefficient 0.45 is further reduced by 10% to 0.4 to account for eccentricity and tolerances in construction. To bring the ultimate capacity to working load capacity, 0.4 is divided by 1.5 (equivalent to a combined load factor) to obtain a coefficient of 0.267 (note: 1.4 and 1.6 were load factors for dead and live load, respectively, based on British Standards). Therefore, CP4 recommended the pile structural capacity (working stress) to be 0.25fcu. This capacity is to be compared to column loads (serviceability limit state) acting on the pile, without any load factors. When CP4 was in use, the common concrete strength for bored piles was C30 (fcu = 4,350 psi or 30MPa), which means that the allowable stress was 1,088 psi (7.5 MPa). Also, CP4 limits concrete strength to 1,088 psi (7.5 MPa) to account for quality control issues when pouring concrete into a hole underground. Even with higher concrete strengths, there is a need to be mindful that such a high strength of concrete may not be adequately compacted and subjected to issues associated with bored pile construction, such as mixing with water and soil, necking, etc. For this reason, 1,088 psi (7.5 MPa) was the maximum working stress allowed in concrete, even if much higher strength of concrete was used.

Structural Capacity Based on EC2

Using Eurocodes, the structural design of reinforced concrete is in accordance with EC2. EC2 provides the following equation for predicting the ultimate capacity of reinforced concrete piles: NRd,p = Acfcd,p where fcd,p = αcc,p fck/γc,f According to EC2, αcc “is the coefficient taking account of longterm effects on the compressive strength and of unfavorable effects resulting from the way the load is applied” and 0.85. γc,f is the partial safety factor for concrete (1.5 x 1.1; 1.1 being required for casting piles without a permanent casing). With all these factors, the ultimate stress in concrete becomes: fcd,p = αcc,pfck/γc,f = 0.85 x (0.8fcu)/(1.5x1.1) = 0.412fcu Under Eurocodes, the load factors for permanent (dead load) and variable action (live load) are 1.35 and 1.5, respectively. Because permanent loads are much higher than variable Table 6. Construction tolerances for bored piles. loads for most structures, a combined load factor Code Tolerance can be assumed to be approximately 1.4. ACI-336 4% of the diameter or 3 inches (75mm), whichever is less The working stress of concrete then becomes 0.29fcu (higher than 0.25fcu using ACI-318 or EN1536 (execution 4 inches (100 mm) (≤40 inches diameter) 0.25fcu using CP4). standard for bored piles) 0.1D (40<D≤60 inches diameter) By comparing the working stress allowed for 6 inches (150mm) (>60inches diameter) concrete, it appears that EC2 allowed a 16% CP4 3 inches (75mm) higher value as compared to ACI-318 and CP4.

16 STRUCTURE magazine

and specify higher strength concrete for bored piles. However, designers should be cautioned on the need to ensure stringent quality control measures during pile construction and verify that concrete strength can be achieved on-site.â&#x2013;

Table 7. Allowable concrete stress in bored piles in compression.

Code

Allowable Concrete Stress (psi or MPa)

ACI-318

0.25fcu

EC2

0.26fcu

CP4

0.25fcu and not greater than 1088 psi (7.5 MPa) INFO

The online version of this article contains references. Please visit www.STRUCTUREmag.org. SPECS

However, in EC2, it is necessary to reduce the design diameter of a FileName:19-1670_Ad_1/2IslandStructure_July_BridgeRepairSolutions Page Size: 5w" x 7.5h" bleed bored pile by 2 inches (50mm) for diameters greater than or equal N/A Pages: 1in reviewing to 40 inches (1,000mm) when there is no permanent casing. Job#: This 19-1670 H. Y. Ng isPR#: a Principal Engineer with a localNumber authorityofinvolved Artist: Georgina Morra Email: gmorra@mapei.com Bleed: Yes Amount: .125" 1 1 4top 4 E of . Nthe e w p1.1 o r t factor C e n t e applied r Dr. design diameter reduction is on to the and granting approval for structural and geotechnical design. Deerfield Beach, FL 33442 Date: June 7, 2019 10:29 AM Colors: CMYK Process, 4/0 concrete partial factor of 1.5. Such a reduction in capacity using (xyhng@hotmail.com) N O T E : C O L O R S V I E W E D O N - S C R E E N A R E I N T E N D E D F O R V I S U A L R E F E R E N C E O N L Y A N D M A Y N O T M A T C H T H E F I N A L P R I N T E D P R O D U C T. EC2 equations is to allow for greater uncertainties in casting concrete underground without a permanent casing. With these two explicit provisions addressing uncertainties in the construction process, EC2 uses a slightly higher allowable concrete stress of 0.29fcu. If the 2-inch (50mm) reduction in design diameter is accounted for, the concrete stress in EC2 reduces by another 10% (using 382/402 = 0.9) to 0.26fcu.

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Conclusion Although there is no upper limit on concrete stress in ACI-318 and EC2, designers may not want to use the maximum stress allowed. For example, a concrete with fcu = 5,800 psi (40 MPa) has allowable stress of 1,711 psi (11.8 MPa) designed with EC2. In practice, the designer may choose to limit the stress to 1,450 psi (10 MPa) as an additional safety precaution. Allowable compressive stress in concrete for bored piles appears to be consistent across the three different codes (Table 7). Most codes recognize that concrete cast underground is more uncertain compared to a column cast in a superstructure and, therefore, a safety factor that accounts for this is required. CP4 uses a 1,088 psi (7.5 MPa) stress limit to guard against quality issues and therefore places a disincentive for using higher strength concrete, while the other codes do not prescribe concrete strength limits. Designers can take full advantage of the higher concrete stress

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Column loads are usually transferred to bored piles through a pile cap. Pile caps help to distribute loads to piles in a group and minimize the effects of eccentricity, that is, variation in the individual pile positions. Pile caps may need to be redesigned if there are piles constructed exceeding the allowable tolerances. Construction tolerances allowed by the codes are summarized in Table 6.

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M A R C H 2 02 0

structural TESTING Lab Test Confidential: Seismic Loading Protocols By Matthew S. Speicher, Ph.D., and Bruce F. Maison, P.E., S.E.

he rise in performance-based engineering, in which a structure is proportioned to meet certain predictable performance requirements,

necessitates reasonable estimates of component behavior during earthquakes. It is customary to determine component properties via physical lab tests. For components such as concrete anchors, verification of the ultimate strength is required and quasi-static pull tests are sufficient. The situation is more involved for other components, such as beam-to-column assemblies, since an earthquake produces dynamic back-and-forth cyclic actions and the component is often expected to deform inelastically. Although earthquake loadings are dynamic, quasi-static component tests are routinely performed due to the complexity and expense of dynamic testing (Figure 1). The most widely used loading patterns (protocols) consist of fully-reversed cyclic loading having progressively increasing displacement amplitudes. These are often referred to as simply “cyclic” tests and are termed here as “standard” protocols. Figure 2 contrasts a standard protocol to that of a simulated earthquake response from a building undergoing inelastic actions. Notice the standard protocol has many cycles as opposed to the earthquake response having relatively few cycles with a one-direction bias. The Applied Technology Council project ATC-62 was one of the first extensive studies that found standard protocols too demanding compared with actual earthquake loadings, and the use of standard protocols can lead to the overly-conservative representation of component seismic performance (FEMA, 2009). Reports on lab tests often do not discuss how a component response is influenced by the loading protocol, so engineers are left to accept the results as intrinsic “seismic” properties. This behind-the-scenes aspect is the subject of this article. The purposes are to:

Figure 1. Illustration for a lab test of beam-to-column assembly and various quasi-static loading protocols.

• Illustrate differences in test results from using standard protocols and actual earthquake loading patterns. • Point out that lab test data based solely on standard protocols can lead to unduly conservative component acceptance criteria as well as computer models that overestimate building response in performance-based engineering. • Encourage future experimental projects to use protocols reflecting actual earthquake response patterns to better estimate component seismic behaviors.

Realistic Protocols

With the advancement and gaining popularity of nonlinear structural analysis, a better understanding of actual seismic response has led researchers to propose different protocols that may more appropriately reflect earthquake inelastic building response (Figure 3, page 20). Such protocols are termed here as “realistic.” Realistic protocols are different from standard protocols by having fewer cycles and a one-direction bias. Tests using realistic protocols have been relatively infrequent, but they do show that components generally have more ductility compared to results using standard protocols. For example, Figure 4 ( page 20) shows the drifts describing various damage states of steel columns from tests conducted by Elkady et al (2018). Use of a realistic collapse-consistent protocol indicates the columns have about twice the inelastic deformation capacity than those from using standard protocols. Accordingly, column acceptance criteria would be very conservative should they be based entirely Figure 2. Typical standard protocol compared to simulated building inter-story drift earthquake response (Maison et al., 2016) on the standard protocol test results. continued on next page M A R C H 2 02 0

Backbone Curves In performance-based engineering, the backbone curve is the customary way of describing component behaviors over a range of deformations. It is formulated as an envelope of hysteresis loops from component lab tests and is the chief factor for displacement-controlled component Figure 3. Realistic protocols reflecting earthquake inelastic building response. modeling and acceptance criteria. Figure 5 illustrates the strong influence the protocol can have on envelopes (backbones) of test results Additional Studies from identical reinforced concrete bridge piers (FEMA, 2009). The backbones were essentially the same out to about 2% drift, but they The National Institute of Standards and Technology (NIST) perdiffered significantly for larger drifts depending on the protocol. formed comprehensive analytical investigations into the correlation Notice how the standard protocols produced backbones with the between the seismic performance of steel buildings designed to current smallest drift capacities. standards and their performance as quantified using performanceThe trend is also evident in other materials. Figure 6 shows results based engineering (Harris and Speicher, 2015). It was found that from identical plywood shear walls subjected to different protocols performance-based engineering often rejects “new” buildings as being from a test conducted by Gatto and Uang (2003). There was little unsafe even though they meet current building codes. One likely difference between the cyclic envelope and the monotonic test out reason is conservatism in performance-based engineering acceptance to about a 3% drift. However, at about 4% drift, the cyclic envelope and modeling criteria since the vast majority of existing lab test data strength was less than one-half that of the monotonic test strength. used standard protocols having the shortcomings discussed above Figure 7 shows more test results from identical plywood shear (Speicher et. al, 2018). walls but, in this case, a more realistic protocol is compared to the Likewise, the ongoing Applied Technology Council project ATC-116 monotonic. The envelope from the realistic test is similar to the investigates the paradox of the better-than-expected performance of monotonic test suggesting that earthquake inelastic behavior is short period buildings. One principal finding is the important effect better represented by the monotonic as opposed to the standard of the component post-peak residual strength (Kircher et al., 2018). protocol. In fact, Professor Helmut Krawinkler, who was responsible To arrive at performance observed in actual earthquakes and to be for the development of several popular loading protocols, advocated consistent with the judgment of earthquake engineers, the ATCcomplementing standard tests with other tests, including monotonic, 116-improved analytical models have significant residual strength whose loading histories better represent response close to collapse beyond 10% drift (30% to 60% of component ultimate strength). (Krawinkler, 2009). Lab tests using realistic protocols generally have much greater residual Therefore, key points about backbone curves derived from standard strength than those using standard protocols (Figures 6 and 7 ). This lab tests are as follows: gives support to the notion that lab tests using standard protocols lead • Standard protocols do not mimic actual earthquake demands. to conservative seismic performance criteria (Speicher et al, 2018). Lab tests using such protocols generally result in backbones having progressively decreasing deformation capacities accordASCE 41 Performance-Based Engineering ing to increasing numbers of fully-reversed loading cycles. • For moderate drifts, say up to about 2%, backbone curves are ASCE 41, Seismic Evaluation and Retrofit of Existing Buildings, is generally independent of the loading protocol. an industry standard incorporating performance-based engineering • For large drifts, say greater than about 2%, backbone curves methods (ASCE, 2017). Figure 8 depicts the ASCE 41 construction can be strongly influenced by the loading protocol. of a backbone curve as an envelope of data from a component lab test using a standard protocol. The peak displacement reversal points in the hysteresis loops govern where the abrupt decline in strength occurs in the backbone, giving the impression that the component has zero strength beyond deformation E. However, this is misleading since the use of a protocol having different displacement reversal points would shift the deformation to E´ as indicated. Hence, the backbone curve is dependent on the loading protocol, and standard protocols do not reflect building seismic inelastic response. The latest edition of ASCE 41, ASCE 41-17, recognizes the importance of loading protocols and provides additional freedom in protocol selection to better represent actual seismic loading patterns. It does not prescribe a specific “one-size-fits-all” loading protocol due to the variety of factors involved with a particular component, e.g., perFigure 4. Drifts defining column damage states from lab tests using different loading protocols formance objective, type of structure, and seismic setting. (Elkady et al., 2018). 20 STRUCTURE magazine

Figure 5. Envelopes of cyclic test results (backbones) from six identical reinforced concrete bridge piers subjected to various loading protocols (adapted from Figure 2-20 of the ATC-62 project; FEMA P-440A).

Figure 6. Lab test results from tests of two identical plywood shear walls using different loading protocols. (Adapted from Figure 6g in Gatto et al., 2003)

Since most previous component tests were performed using standard protocols, ASCE 41-17 allows such test data to be supplemented to better define behavior at near-collapse displacements (Section 7.6). To ensure reasonable protocols are selected for a particular component and project, concurrence is required by independent peer reviewers. The rationale for the current ASCE 41 provisions is in a paper by Maison and Speicher (2016).

2) Derived backbone curves used in analysis models for seismic evaluation can lead to over-estimation of peak inelastic displacements. The above shortcomings have a compounding effect, causing rejection of buildings that would otherwise be considered acceptable should component behaviors be based on tests using realistic earthquake loading patterns. When using lab test data, it is important to be aware of the loading protocol used. The test results must be scrutinized within the context of the loading protocol. If a standard protocol is used, then it is likely the data represents a very conservative description of component inelastic behavior under earthquake loadings. In this case, it may be appropriate to modify the data for use in performance-based engineering. The current ASCE 41-17 outlines one way this can be done. It is encouraged that future lab tests include realistic earthquake loading protocols so that the results are best suited for performance-based engineering.â&#x2013;

Conclusion

With the emergence of performance-based engineering, it is essential to have reasonable estimates of component behaviors during actual earthquakes. Standard lab test loading protocols typically consist of fully-reversed cyclic loading with progressively increasing displacement amplitudes. However, realistic earthquake loading patterns are not like standard protocols and can lead to different conclusions about component performance. There are two underlying shortcomings when using component lab test data derived from standard protocols. 1) Ductility can be underestimated, which, in turn, can lead Figure 7. Lab test results from two identical plywood shear walls to overly conservative accepusing different loading protocols. (Adapted from Figure 6h; tance criteria. Gatto et al., 2003)

The online version of this article contains references. Please visit www.STRUCTUREmag.org.

Matthew S. Speicher is a Research Structural Engineer at the National Institute of Standards and Technology (NIST) in Gaithersburg, MD. Matthew is an active member of the ASCE 41 Analysis Subcommittee and conducts research to advance the use of performance-based seismic design. This article is an extension of his presentation at the 12 th Canadian Conference on Earthquake Engineering titled â&#x20AC;&#x153;The blind side: using 'canned' loading protocols in seismic testing" (Speicher and Maison, 2019). (matthew.speicher@nist.gov)

Figure 8. Construction of ASCE 41 component backbone curves as envelopes of cyclic test data (adapted from Figure 7-5 in previous ASCE 41-13). The displacement reversal point is at E in one protocol, whereas it is at EÂ´ in another protocol.

Bruce F. Maison is a Consulting Engineer practicing in El Cerrito, CA. Bruce is an active member of the Existing Buildings Committee of the Structural Engineers Association of Northern California (SEAONC). He is also a member of the ASCE 41 Analysis Subcommittee. (maison@netscape.com)

M A R C H 2 02 0

structural SYSTEMS Structure Design Considerations for Building Enclosures By Matthew L. Wagner, S.E., and Patrick Olechno, P.E.

he International Building Code (IBC) and design standards used by structural engineers are developed with the intent to provide minimum requirements for the design and construction of a structure to help assure the public health, safety, and welfare. While a particular structural design is required to satisfy the IBC, as well as material design standards and public safety requirements, the IBC does not directly address the effects of the vertical movement of the building structure on the building enclosure. For example, what are the deflection limits for spandrel beams or slab edges that support the building enclosure? The IBC provides deflection limitations based only on the span of the structural member; however, more restrictive limitations may govern based on properties of the building enclosure system or architectural vertical movement joint details. Enclosure systems and framing systems exhibit unique behaviors, and interaction and structural behavior between them may not be adequately accommodated using customary span-to-deflection ratios. The structural engineer will need to understand how the enclosure is expected to behave during construction and throughout the service life of the building to design a structural frame that properly accommodates the movement limitations of the building enclosure. This article is focused only on issues associated with vertical deflections. Lateral movement limitations, such as the case of whole building drift limitations, are more clearly defined in the IBC and ASCE/SEI 7, Minimum Design Loads for Buildings and Other Structures, and are not subject to creep, varying superimposed loads, or long-term deflection effects.

IBC and Material Standards There are two types of deflection limits: the traditional limit prescribed in codes and standards based on the span of the structural member and absolute dimensional limits based on the movement capabilities of joints designed to accommodate the deflections. Deflections based on ratios of span length consider the maximum rate of change between two points along a deflected shape and are useful for preventing damage of the nonstructural material or system that is being supported by the structural member or even the member itself. With the advancement of construction materials and the increase of member spans, the traditional span ratio limit may be inadequate in some cases to prevent damage of nonstructural systems that are supported or attached to the structure. Structural engineers should consider several interacting effects when designing structural members that support building enclosures to provide structural systems with deflection characteristics that are compatible with the enclosure system. However, deflection requirements provided in the IBC do not address this situation. The summary below describes how the IBC and design standards approach structural deflections, and how building enclosure considerations are addressed. 22 STRUCTURE magazine

Figure 1. Common building enclosure types; a) Infill wall; b) Curtain wall; c) Column-supported wall; and d) Multi-span wall.

IBC The serviceability requirements in the 2018 IBC limit the deflection of structural members to the more restrictive case of either the referenced material-specific standard or those deflection limit values provided in Table 1604.3. The deflection limits in this table are based on the member span length and are meant to address serviceability and limit damage to architectural finishes. Advancements in framing systems and materials have led to longer spans and greater deflections. While some modern architectural finishes have been able to accommodate these larger deflections, many building enclosure systems may not be able to accommodate the larger deflections.

ASCE/SEI 7 The Commentary Section CC.2.1 for Appendix C in ASCE/SEI 7-16 states that, in certain conditions with long spans, it may be appropriate to limit maximum deflections to prevent damage of nonstructural elements, along with the suggestion to limit deflection to 3â &#x201E;8 inch for the case of non-load-bearing partitions.

ACI-318 Deflection requirements for concrete are governed by ACI-318, Building Code Requirements for Structural Concrete and Commentary, Chapter 24, which specifies a serviceability deflection limit for concrete members of L/480 in Table 24.2.2. A footnote to the table allows greater deflection if the deflection does not damage supported or attached elements.

AISC-360 For steel structures, AISC-360, Specification for Structural Steel Buildings, does not explicitly state serviceability deflection limits for steel members supporting building enclosures but instead references AISC Steel

Design Guide 3, Serviceability Design Considerations, and Design Guide 22, Façade Attachments to Steel-Framed Buildings. These guides contain useful information and guidance for developing spandrel member design criteria.

TMS 402

at each floor level or designed to span multiple levels (Figure 1d ). It is essential to understand the interaction of the wall system support locations and vertical movement joint details and capabilities, considering the various combinations of loads that may be present in-service. This is particularly true when curtain walls accommodate the accumulated movement of multiple structural levels.

Section 5.2.1.4.1 of the 2016 Edition of TMS-402, Building Code Requirements and Specification for Masonry Structures, prescribes a design deflection limit of Column-Supported Walls L/600 for serviceability. Earlier editions of TMS-402, such as the 2005 Edition, In some cases, enclosure panels are included an additional overall deflecdesigned to be supported from columns. tion limit of 0.3 inches. The 0.3-inch A common type is a precast spandrel deflection limit has since been removed, panel installed in front of the floor slab, starting with the 2008 Edition; many typically supporting an infill ribbon structural designs still apply this crite- Figure 2. Relative spandrel member deflections (DB1 and DB2) window system (Figure 1c). Though the rion on most applications. floor does not support the enclosure, are similar, but the global deflection along B2 is additive to the Consideration should be given to limit- cantilevered end deflections. floor deflection relative to the columning differential spandrel deflection to 3⁄8 supported panel should be considered. inch based on ASCE commentary and historic masonry requirements for infill wall systems, and 1⁄2 inch for unitized curtain wall systems Cantilevered Perimeter Framing designed to accommodate vertical movements between floors within a manufactured stack joint assembly. Increased vertical movements Traditional deflection limits based on individual member spans do are possible and require additional effort and coordination among not adequately address cantilevered perimeter framing. The traditional project team members to accommodate proper support and move- deflection limits apply only to relative member deflections, which ments of the building enclosure. are measured as the difference between member ends and deflection

Building Enclosure Systems When designing the structure for the building enclosure, understanding and accommodating for the structural behaviors of the enclosure system and materials to be used is critical. The enclosure system must support itself and allow for predictable movements of the structural building frame.

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The many varieties of infill walls, such as translucent window walls and solid-framed walls, are commonly supported by each floor level, extending and attaching to the floor above (Figure 1a). Masonry veneers, for example, are often laterally supported by an infill wall system and gravity supported at each level with a shelf angle. Variations of floor elevation are adjusted using shims at the base of the infill wall, allowing conformance with defined vertical installation tolerances. A vertical movement joint at the top of the wall system below the floor. or below the shelf angle in the case of masonry veneer, accommodates the differential vertical displacement between adjacent floors.

Curtain Walls Curtain wall systems are typically thin-framed wall constructions often including infills of glass, metal panels, or thin stone. The most recognizable framing for curtain walls is aluminum. The curtain wall framing is installed outboard of the structure, extending past the outer face of the structural framing. As shown in Figure 1b, the curtain wall framing is directly attached to the main building structure and does not carry the floor or roof loads of the building. Lateral and gravity loading imposed on the curtain wall are transferred to the structural building frame. Curtain walls are typically laterally supported at each level. Gravity supports can be located

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along its span (Figure 2). While individual members of a cantilevered system can meet IBC deflection limitations, when relative deflections are combined, the resultant deflection can be significant and must be accommodated by the building enclosure system (Figure 2).

Live Load Deflections

Reduction of live loads when determining spandrel deflections is appropriate. Reductions can be based on code-calculated values or a more realistic percentage depending on the application. ASCE/SEI 7-16 Appendix C Commentary suggests using 50% live load when combined with long-term Sizing Vertical effects. In addition to a uniform reduction of live loads for deflection calculations, one Figure 3. Deflected member shape based on uniform Movement Joints and pattern loading of a cantilevered member. should also consider pattern loads where the The detailing of horizontal sealant and curfloor plate is cantilevered. In this specific case, tain wall joints to accommodate vertical movement is not typically a loaded back span could result in upward movement at the cantilever in the structural engineer's scope of services. However, the structural edge, whereas loading just the cantilever portion may result in more engineer must be knowledgeable of the structural behavior of the significant deflections when compared to uniform loadings (Figure 3). project's building enclosure and assist the design team by providing sizing criteria for items such as joints and defining movement requireLong-Term Deflections ments, including limitations for the building enclosure. Variables that affect vertical movement within the horizontal joints In addition to initial elastic deflection, concrete also exhibits addiare broken into two categories â&#x20AC;&#x201C; enclosure movement and structural tive long-term deflections due to creep and shrinkage. Long-term frame deflections. flexural deflection of horizontal members and axial shortening of columns due to both initial and long-term effects all need to Dimensional Changes of Enclosure Materials be carefully considered when dealing with taller buildings, as Volumetric changes of materials need to be considered when deter- the building enclosure installation will likely begin before the mining appropriate widths for enclosure vertical movement joints. structure is complete and before initial and long-term deflections Temperature affects all materials differently. The rate of thermal expan- have taken place. sion for aluminum is approximately twice the rate for concrete or steel and four times that of clay masonry. Since the curtain wall and infill Coordination window wall systems are typically built with aluminum extrusions and can experience extreme changes in temperature, the calculation of the There must be communication and design coordination between thermal volumetric change is fundamental to determining appropriate all members of the project design team, including the building joint widths. While manufacturers of fabricated enclosure systems, structural engineer, architect, and various design professionals. such as curtain walls, will determine the expected volumetric changes There should be a clear understanding of how the structure will within their system, the architect will be responsible for determining interact with the building enclosure, including the proposed the volumetric change of most infill wall systems, such as concrete support locations, loads, and deflection limitations. The design and clay masonry veneers which will experience some permanent parameters should be conveyed in the contract documents and contraction or expansion, respectively, over time. then checked throughout the submittal process. Whatever deflection limitation or assumption is used during the design of the Structural Frame Deflections structural frame, it should be included in the building enclosure When sizing enclosure vertical movement joints at locations where specification or defined preferably on the architectural drawings. floor stiffness and loadings are similar between adjacent levels, the Well-coordinated building enclosure and structural support systems differential deflection between the levels can be estimated using a will significantly reduce possible constructability and performance portion of the calculated deflections. However, where stiffness and issues related to the interaction of the building structure and wall loading vary (e.g., a 1st-floor wall with a stable and stiff foundation), enclosure systems. the joint between the elevated framing and the foundation will need to account for all of the elevated frame deflection. Final Thought Even when the structural floor framing is repetitive, changes in curtain wall weight, floor finishes, and construction methods could The IBC and industry standards do not provide prescriptive requireresult in differential deflection between levels. The enclosure system ments with respect to resultant deflection limits of perimeter floor weight can vary due to changes in materials or height of wall sections, framing members, which are attached to or support building enclosuch as where lower or upper-level sections of a hung curtain wall may sures. As such, the design team must consider the behavioral differ from the typical floor-to-floor section. Floor finishes or dead differences and provide an appropriate analysis that addresses load may also vary between floors. In the case of an office tower, the the impact of wall enclosure systems on each project.â&#x2013; flooring may not be selected or installed until a tenant has signed a lease. In this case, some floors may have installed flooring while the Matthew L. Wagner is a Senior Project Engineer at Raths, Raths & Johnson, adjacent floor may remain unfinished for years. Though the structural Inc. (mlwagner@rrj.com) design may account for heavy thickset tile flooring, a tenant may opt Patrick Olechno is a Professional Engineer at Stantec. for a lightweight floor, such as wood or carpet. Lastly, even though the (patrick.olechno@stantec.com) design stiffness may be identical between floors, as-built irregularities may cause the two floors to deflect differently. 24 STRUCTURE magazine

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Napa County HISTORIC

Courthouse PART 3

By Brett Shields, P.E., Luke Wilson, S.E., and Kevin Zucco, S.E. Figure 1. Entry showing damage taken the morning of the earthquake.

n August 24, 2014, the South Napa Earthquake left the Napa product requires surface preparation to receive a base coat for finished County Historic Courthouse heavily damaged with partially plaster. The epoxy-based resin creates a sealed surface over the historic collapsed walls, ceilings, and extensive wall cracking (Figure 1). The brick, restricting the brick’s natural ability to breathe. Maintaining City of Napa red-tagged the courthouse as unoccupiable, which began this breathability was critical for the preservation of the historic brick. the extensive damage documentation and repair effort. The overarchDuring concept design, an overlay product used extensively in ing goal throughout this process was to Europe, Fabric-Reinforced Cementitious provide a solution to repair and preserve Matrix (FRCM), was being introduced to as much of the historic building as practithe California market by manufacturers cal while providing improved detailing. including Simpson Strong-Tie. FRCM The historic courthouse building is a consists of either uni-directional or bi140-year-old, two-story, unreinforced brick directional carbon fiber fabric (Figure 2) bearing wall structure with wood-framed embedded between lifts of cementitious floors and roof, located in downtown Napa. matrix installed in ¼- to ½-inch lifts. The The building had a significant remodel and lifts can either be installed by hand similar retrofit in 1977 which included concrete to plaster, or as a spray installation similar to frames for new openings, concrete or conshotcrete. The fabric, which comes in rolls crete masonry unit (CMU) infill of existing up to 77 inches wide, is pressed into the openings, and many other small renovations. Figure 2. FRCM uni-directional (left), bi-directional (right). base lift before having a cover lift of matrix The observed earthquake damage in the brick walls varied from installed. The total thickness for one layer of FRCM is approximately miscellaneous small cracks to significant cracking with permanent 1 inch plus ½ inch for each additional layer of FRCM. As a porous in-plane and out-of-plane displacements in both principal directions, cementitious material, FRCM is more homogeneous with existing to the partial collapse of wall sections. The repair approach needed to brick stiffness and mechanical properties compared to epoxy-based provide a similarly diverse set of options to match observed condi- overlays and allows the historic brick to breathe. The FRCM surface tions. Traditional brick repair methodologies, repointing, and grout preparation only requires a surface clear of loose debris, cleaned, and injection were used in areas of minor damage where appropriate. saturated surface dry for cementitious matrix adhesion and curing. The Complete wall reconstruction using specially detailed CMU was used FRCM can double as the base coat for plaster installation, removing in areas with permanent deflections and partial collapse. However, an extra preparation step required for FRP. a third repair approach was needed to address the majority of wall The FRCM, combined with grout injection of large cracks in the brick areas exhibiting extensive cracking and minimal displacement. The substrate, was used to restore in-plane capacity to extensively cracked damage documentation and general repair efforts were outlined in brick walls in place of typical traditional brick repair approaches. The the December 2019 and January 2020 editions of STRUCTURE. matrix’s tolerance in lift thickness, and its ability to locally fill voids of up to 2 inches, provided flexibility for the structure’s variable (brick, concrete, and CMU) surface conditions and interfaces. Additionally, Fabric Reinforced Cementitious Matrix the FRCM was used to provide nominal continuity through structural Early in the repair design, the design team considered using tradi- elements such as floors and wall intersections using bundled splay tional Fiber Reinforced Polymer (FRP) overlay on brick walls that anchor ties, laps between different materials, and nominal tension demonstrated extensive cracking. However, FRP presents challenges in ties at reentrant corners exhibiting spandrel joint damage. a historic brick application. It requires significant surface preparation During design, Simpson Strong-Tie was in the process of gaining seisof the brick to provide a flat surface for fiber application. The finished mic approvals for FRCM in California. Therefore, the design process was 26 STRUCTURE magazine

a collaborative performance-based specification design-build approach. Simpson Strong-Tie acted as the engineer of record for the design of the FRCM product using American Concrete Institute (ACI) 549.4R-13, Guide to Design and Construction of Externally Bonded Fabric Reinforced Cementitious Matrix Systems for Repair and Strengthening Concrete and Masonry Structures, with direct input and oversight from ZFA. Per ACI 549.4R, there is no strengthening limit for earthquake and wind forces as they are not considered likely to damage the unprotected FRCM compared to a fire or blast loads. However, FRCM strengthening should not exceed 50 percent of the capacity of the original structure to limit force transfer to the brick. And, strengthening should be limited to 12-inch-thick walls maximum. Additionally, the combined brick and FRCM shear capacity should be compared to the limit state shear capacity of substrate toe crushing. Based on the reduction in capacity determined in the damage documentation process, ZFA provided target minimum in-plane strengthening loads, ranging from 10 to 25 percent of the undamaged brick wall capacity, for the design of the FRCM to repair the wall to its undamaged capacity. ZFA also provided construction details for the installation of FRCM to walls and their intersections with surrounding structural elements. The design team coordinated the number of layers and directionality of FRCM on each wall face to minimize impact to significant architectural areas. The result was that each wall to receive FRCM had custom reinforcing layups, including at least one layer of bi-directional fiber reinforcing for continuity and added ductility.

Mockup

Figure 3. Mockup prior to installation of FRCM.

Construction As one of the first installations of the FRCM product in California, there were challenges to overcome. FRCM requires the substrate to be saturated surface dry (SSD). SSD is the condition at which the wall substrate is saturated, refuses to absorb additional water, and the surface is dry to the touch. The building has been enclosed for 140 years with central heat since the 1970s, which led the walls to be dry. The walls were wetted every 30 to 60 minutes during the day and covered overnight for approximately 48 hours before the walls reached SSD. This is critical to the installation of FRCM to prevent the wall substrate from absorbing the moisture in the thin lift layers of cementitious matrix, causing the material to flash or surface tear while the contractor finishes the installation. This challenge was not identified in the mockup because the mockup was new construction and was exposed to the outside elements before the installation of FRCM. FRCM was installed with a pump, hose, and nozzle, similar to shotcrete. Typical shotcrete lifts are three to four inches minimum, meaning there was a learning curve to install the ½-inch lifts required for FRCM. The size of the walls complicated this. The typical wall height was 16 feet on the first floor and 18 to 20 feet on the second floor, requiring multiple levels of scaffolding to access (Figure 5). Matrix had to be installed around scaffolding, leading to shadowing and/or thin spots at the scaffolding planks and legs and thick spots at

Due to the complexity of detailing and installing CMU, historic brick, wood framing, wall anchorage, and use of a new product in FRCM, the design team recommended the construction of a mockup. The mockup was intended to test installation techniques on similar conditions to the existing building and to identify potential issues before full mobilization during construction. The mockup served as a minimum quality of work example and later served as a surface for mockup of architectural finishes installed over the FRCM. The mockup consisted of approximately 19 linear feet of a six-foottall wall in an ‘L’ configuration (plan view). It included the two typical window head configurations, an exterior corner pilaster, multiple CMU to brick interfaces, control joints, and a wood ledger to install FRCM splay anchors through. Once the structure was completed, FRCM was installed over each of the four sides on two separate days. The mockup was completed early in the construction schedule and observed by the Design Team, Owner’s Representative, General Contractor, multiple representatives from Simpson Strong-Tie, and Project Inspector of Record (IOR). The group debriefed after each installation and circulated lessons learned. Through this process, multiple installation techniques were tested to ensure ½-inch lift heights. Ultimately, the most effective process was determined to be having preset feeler gauges, spot-checking the material as it was installed and using wires to set the final depth (Figures 3 and 4 ) similar to traditional shotcrete Figure 4. FRCM fabric being installed into matrix. installation techniques.

Figure 5. FRCM installation. MARCH 2020

the corners, floors, and ceilings. These variations had to be trimmed to meet flatness requirements for future wall finishes. If the matrix was too thin, the fabric could be unintentionally moved or damaged while worked into the first lift of matrix. Additionally, scaffolding presented an obstacle to the installation of the large rolls of fabric. The fabric could not be sharply bent without damaging the fiber bundles or disconnecting the unidirectional grid from tie strands. Therefore, the FRCM had to be installed from a rolled bundle challenging to work around scaffolding. While the mockup was an excellent proof of concept and uncovered many installation hurdles, it had size limitations that did not uncover challenges of large-scale construction means and methods.

Figure 6. Finished FRCM installation on first floor (left), second floor (right).

Conclusions FRCMs were used to provide in-plane and out-of-plane repair/ strengthening throughout the building, maximizing the amount of historic brick wall that was preserved. Its relatively thin application, substrate tolerance, material compatibility with the existing brick, and its allowance for the walls to naturally breathe were critical to providing a structural solution while minimizing the effect on the historic character of the building (Figure 6 ). FRP would have required significantly more surface preparation in comparison to FRCM, and discrete shotcrete walls would have affected room sizes and required strengthening of collectors and foundations.

While FRCM had multiple benefits, using a new product to California presented several challenges during design and construction. These ranged from working with Simpson Strong-Tie as they developed design guidelines and material information, preparation of 140-year-old brick walls, and translating installation techniques from the mockup to working within the building. A dynamic and collaborative team approach was required by all involved to overcome these challenges.â&#x2013; All authors are with ZFA Structural Engineers in Santa Rosa, California. Brett Shields is an Engineer. (bretts@zfa.com) Luke Wilson is an Associate Principal. (lukew@zfa.com) Kevin Zucco is an Executive Principal. (kevinz@zfa.com)

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Anatomy

OF A HIGH CAPACITY

Timber Connection By D. Scott Nyseth, P.E., S.E.

new timber truss pedestrian bridge in Oregon is using high-capacity timber connections. The span is 114 feet with a truss height and width of 16 feet between centerlines of chords. Historically, timber truss bridges have light roof structures when compared to the timber truss and deck structure. The roof for this bridge weighs more than the truss and deck structure. This provided unique challenges in a high seismic region. Bottom chord seismic tension loads are approximately 300 kips and typical knife plate connections could not develop the strength required for the bottom chords. The recent introduction of proprietary connections utilizing multiple internal steel plates and screws provided a challenge for a much larger connection. The initial bottom chord size from the architect was proposed as 14x18 visually graded sawn timber. When analyzing typical external connection plates with bolts, the connection became so long that the group action factor, based on the National Design Specification® (NDS®) for Wood Construction, became too large, even when looking at various bolt diameters, spacing, and larger chords. The author proposed a continuous glulam member to avoid the tension connections, but that was not the aesthetic the architect wanted for this bridge. Once there was an agreement to use multiple internal knife plates, a design that almost met the tensile capacity of the chord member itself was sought. The upper design strength of the connection is the allowable tensile capacity of the individual wood segments between knife plates, so removal of wood for the knife plates would never allow for the full allowable tensile capacity of the chord away from the connection. After analysis, truss bottom chords were required to be 18x18 Douglas Fir-Larch (DF-L) Select Structural, with a non-incised pressure preservative treatment.

18-inch timber off the shelf is not customary. The author’s firm, Stonewood Structural Engineers, worked directly with the contractor and lumber broker to source the timber and a mill which has on-site graders to get the right timber to meet the specifications. The green milled size of this timber would be 17½ inches square based on the manufacturing and grading rules. Minimizing movement after the CNC processing of the timber is essential for proper fit-up, so Kiln Dried (KD) timber was specified. The Moisture Content (MC) can vary significantly throughout the depth of the timber. At the time a DF-L tree is felled, the MC is about 37% for the heart-wood (center-of-tree) and about 115% for the Sapwood (outer surface) although it dries out much faster than the heart-wood. The KD process is an art form to get a dry timber while minimizing checking and avoiding “killing” the timber and making it brittle, so choosing an experienced kiln operator is important. When specifying MC19 (19% moisture content) KD timber, the moisture content is typically measured with a standard prong moisture meter which may have prongs less than 3 ⁄8-inch-long (Figure 1). With this approach, there is no guarantee that the MC deeper below the surface is even close to MC19. The measurement was specified to be taken 3 inches below the surface (Figure 2) to have a better chance of having an average MC19 throughout the wood When a timber is milled at MC19, versus green, the final dressed size requirement is different. The MC19 minimum dressed size is 17x17 inches versus 17½ x 17½ inches when milled green. The mill started to plane the timber at the smaller dry dimensions once the timber was removed from the kiln. The moisture content was found to be closer to 25% moisture content at 3 inches below the surface. Milling was stopped at that point. The kiln operators felt that additional time in the kiln to lower the moisture content further would create excessive checking, yet the members did not meet the criteria for milling at the 17- x 17-inch criteria. A rule of thumb is a 1%-dimensional change for every 4% change in moisture content to determine the acceptable dressed size. Due to the higher moisture content, it was determined that the final milled size needed to be 173⁄16 x 173⁄16 inches. For member design, the full 17½-inch dimension is used (grade rules account for this) and, for connection design, the 17-inch dimension is used. The equilibrium moisture content (EMC) for the location of the bridge ranges from about 12% to 17% throughout the year. Right before fabrication of the timber was scheduled to begin, the project was delayed. This allowed Stonewood to cut two sections of

Physical Properties of Timbers The following discussion explains some of the issues related to the physical properties of timbers and how they impact the design process. For a project such as this, buying an 18- x 30 STRUCTURE magazine

Figure 1. Initial moisture content with standard prong.

Figure 2. Initial moisture content 3 inches below surface.

timber off the mock-up to determine the actual moisture content of the timber, instead of a single measured point 3 inches below the surface. Over two months, the timber block, which initially measured 17 x 17 x 12 inches and weighed 77.6 pounds (38.8 pcf ), went into the author’s oven at home at 170 degrees Fahrenheit. It underwent about 22 cycles of about 12-18 hours before reaching an oven-dry weight of 62.0 pounds (31 pcf ). The initial moisture content is determined by the formula MC = (initial weight – oven dry weight)/oven dry weight. For our sample, this equates to 25.2% moisture content. As a reference, the initial surface moisture content with a Figure 4. Boxed heart manufacturing yields smaller standard prong was 11.4% (Figure 1) and at 3 Figure 3. Blue outline of smaller oven-dry timber over initial timber. checks and allows for better utilization of the tree. inches below the surface was 22.8% (Figure 2). Our initial direction to measure moisture content at 3 inches below the surface slightly underestimated the actual 6) Apply pressure preservative treatment so that it penetrates all moisture content of the timber. end grain, cuts, and holes. Additionally, from the initial moisture content of 25.2% to the ovendry condition, the timber shrank on average from 17 inches to 16¼ Multiple Internal Plate Connection Design Issues inches. The diagonal dimension shrank from 237⁄8 inches to 233⁄8 inches. The timber shrank relatively less across the diagonal when compared The NDS has a lower bound connection design approach in section to the face dimensions. This created the slightly cupped faces that can 12.3.9. The method of checking the bolt-wood yield mode capacity be seen in Figure 3. The actual face dimensions shrank down to 16¼ at each shear plane, as outlined in this article, rather than using NDS inches, where the rule of thumb of 1%-dimensional change for each 12.3.9, is a rational method which is allowed in the code but can be 4% MC change predicts a 15.9-inch oven-dry face dimension. This is challenging to shepherd through the plan review process depending likely due in part to a comment from the kiln drying representative, on the jurisdiction and its familiarity with wood design. Also note made before the kiln drying started, that large square timbers similar that values for drift pins have a 25% reduction in the NDS due to to what Stonewood was using will often shrink less than timbers that the lack of heads, nuts, and washers. The choice of connection plate are more rectilinear, due to three-dimensional restraint that occurs. numbers and locations is critical to the design. Figure 5a (page 32) Boxed heart timber was used because it allows for better utiliza- shows the standard internal knife plate. This is the typical approach tion of the tree and requires a smaller (less expensive) tree. It can for a hidden connection. Based on a preliminary NDS yield limit also yield smaller checks which are more evenly spread around the analysis, the 4-inch dimensions, noted as x(IIIs), at the exterior wood perimeter, where Free of Heart Center (FOHC) will have one face of segments, represent the minimum width required to force yield the timber that is cut close to the heart, which is where many larger mode IV and not allow a smaller mode IIIs capacity. The double checks originate. For 18x18 Boxed Heart, a 30-inch-diameter log may be required, and for 18x18 FOHC, a 50-inch-diameter log may be required (Figure 4).

1) Order 50% more timber than required to be able to choose the correct grade, appearance, and check locations at critical connection zones. 2) Mill oversized in its green state. 3) Kiln dry until MC19 measured at 3 inches below the surface. 4) Select timbers and mill rough sawn section with true 17-x17-inchsquare dimensions, or larger size based on actual moisture content higher than MC19. 5) CNC all cuts and bolt holes per final connection design (discussed in the Multiple Internal Plate Connection Design Issues Section).

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Timber Specification and Detailing Process

MARCH 2020

(4) rows of (7) 7⁄8-inch-diameter bolts (28 total). The author chose to show bolt capacities in this table instead of the reduced drift pin capacities because the higher bolt capacities create more “net tension” and “tearout” controlled designs, where the reduced drift pin capacities controlled the design except where the inner members' thickness was 6 inches or less. The final connection shown Figure 6. Net section tension failure mode. in Figure 7 utilized the 8-inch inner section and (4) rows of (8) 7⁄8-inch diameter drift pins. Table 2 shows the final design capacity of this connection. Note that this drift pin connection has 10% less capacity even though it has (4) more dowels, because of the 25% reduction for drift pins.

Figure 5. a) Standard internal steel knife plate; b) Multiple steel knife plates.

crosshatch represents the region of excess wood not being utilized for the dowel capacity. Figure 5b shows how the use of multiple knife plates doubles the capacity of each dowel in the connection. The 2½-inch dimensions, noted as x(Is), represent the minimum width required to force yield mode IV and not allow a smaller mode Is capacity. The single cross hatch represents the area not required to maximize the dowel capacity, but which is required to preclude a net tension failure of the inner section, as explained below. As noted above, each shear plane was analyzed to determine bolt capacities. For a multiple interior knife plate connection, the assumption is that Modes II, IIIs, and IIIm cannot occur at the inner sections, and Modes II and IIIm cannot occur in the outer sections. The spacing between knife plates is important when determining the capacity of the dowel, but of more importance is the impact the spacing of the plates has on the net wood tension failure mode and wood tear-out failure modes that must be checked at all connections (NDS Appendix E provides an approach for this). The net tension failure mode is shown in Figure 6. The inner section between plates resists about twice the load as the outer sections, as can be seen at the bottom of Figure 5b. Therefore, as a first design step, set the inner section as twice the width of the outer sections. Table 1 shows the comparison of adjusted bolt capacities and net section capacities for a double 3⁄8-inch knife plate connection with

Connection Design Steps 1) Determine dowel diameter and number of knife plates that maximize the wood section use for yield mode IV dowel capacity. 2) Determine yield mode capacity at each bolt shear plane. 3) Determine load in each wood section based on the number of shear planes and capacities in step two above. 4) Check net tension, group tear-out, and row tear-out. 5) Modify variables to eliminate failure modes and/or optimize the connection. 6) Check steel plate capacities per appropriate steel standards.

Technical Topics Note that the lower bound approach in NDS 12.3.9 can be unconservative for this specific type of connection design. When proportioning the spacing of members and knife plates, group tear-out and net tension failure of the members in the connection area is a critical factor. If member size or knife plate spacing is based on a lower bound value per 12.3.9 when the more rigorous calculation anticipates a higher

Table 1. Comparison of adjusted bolt capacities and net section capacities for the double 3/8 - inch knife plate connection with (4) rows of (7) 7/8 - inch-diameter bolts (28 total).

Inner member thickness (in.)

Total Adjusted Connection capacity nZ’ (kips)

Total Adjusted Bolt Capacity nZ' (kips)

279.3

299.8

7 8 9

Adjusted Bolt Capacity nZ' Per Section (kips)

Row Tear-out ZRT' (kips)

Net Section Tension ZNT' (kips)

Group Tear-out ZGT' (kips)

Inner sections

Each Outer section

Inner section

Each Outer section

Inner section

Each Outer section

Inner section

Each Outer section

357.1

178.6

89.3

100.7

113.3

190.4

214.2

105.3

118.5

357.9

179

89.5

120.8

103.2

228.5

195.2

126.4

108.0

318.2

354.4

177.2

88.6

141.0

93.1

266.6

176.1

147.5

97.4

327.3

350.2

175.1

87.6

161.1

83.1

304.6

157.1

168.5

86.9

318.5

334.8

172.5

81.2

181.3

73.0

342.7

138.0

189.6

76.4

Table 2. Comparison of adjusted drift pin capacities and net section capacities for the double 3/8-inch knife plate connection with (4) rows of (8) 7/8 -inch-diameter drift pins (32 total) used on bridge bottom chord splice.

Inner Total Adjusted Total Adjusted Adjusted Drift Pin Capacity nZ' Per Section (kips) member Connection Drift Pin Inner Each Outer thickness capacity nZ’ Capacity nZ' sections section (in.) (kips) (kips) 8

292.8

32 STRUCTURE magazine

292.8

146.4

73.2

Net Section Tension ZNT' (kips)

Row Tear-out ZRT' (kips)

Group Tear-out ZGT' (kips)

Inner section

Each Outer section

Inner section

Each Outer section

Inner section

Each Outer section

161.1

83.1

348.2

179.5

179.4

92.5

capacity, a potential group tear-out or net tension failure mechanism could be forced in a connection. One more lesson learned on this project was that compressive member end bearing requires a 20 gauge or thicker bearing plate. Without this plate, the NDS requires a 25% reduction in bearing capacity. This affected the endto-end bearing capacity of truss top chord members.

Fabrication Adjustments

Figure 7. Final connection in progress at concrete abutment

Figure 8 shows the mockup of the typical not yet placed. bottom chord connection with only 28 pins on each side. Four more pins were added to each side on the structural detail to account for the drift pin reduction. The 7⁄8-inch dowels were installed flush and proud of the chord face. As a result of the mockup, dowels were lengthened to extend 3⁄8 inches proud of each face, knife plate slots were reduced to be 1⁄16 inch over steel plate width, and sections of wood marked 1-4 were removed for oven-dry moisture testing and measurement of shrinkage. Because the mockup moisture content was higher than specified, more vertical short slotted holes were used in the steel plates and the milled size was increased from 17 inches square to 17 3⁄16 inches square. Note that, for the two blocks of wood cut for drying, the 17-x17-x12-inch section fit into the home oven with ¼ inch to spare, and the drying process made the house smell great.

Figure 8. Mockup of the typical bottom chord connection, which led to fabrication adjustments.

Summary An 18-x18-inch timber was designed as the bottom chord of a bridge truss. For the truss tension splice connection, choosing the optimal number of interior knife plates, and determining a dowel diameter that most effectively utilized the entire width of the timber, allowed for higher capacity and more compact connections.■ D. Scott Nyseth is the President of Stonewood Structural Engineers, Inc., in Portland, Oregon. Mr. Nyseth is an active participant in the American Wood Council's Wood Design Standards Committee. He is currently a member of the PRG-320 CLT Manufacturing Standard Committee. (scott.nyseth@stonewoodstructural.com)

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building BLOCKS Pre-Manufactured Wood Trusses The Impact of Deferred Submissions and Cost Silos By Kirk Grundahl, P.E.

omponent Manufacturers (CMs) are often a misunderstood business in the construction industry, specifically by the structural

engineering community. CMs are a key supplier of load path resisting structural elements. Premanufactured roof trusses are one such element. Typically, the process of developing both the architectural and structural plans for a building designates“trusses by others.”By definition, trusses then become a deferred submittal. Given that trusses are the primary structural framework that provide resistance to the load path as load flows from the roof to the walls and floors to the walls and foundation, the concept of trusses being a deferred submittal presents both engineering and communication challenges. Given these challenges, why is this the preferred method in the market? Generally, cost; much of the root cause of the problems we all face in the construction industry can be traced to a singular focus on lowest cost. The engineer’s silo of work is bid out. This used to be illegal, due to concerns over cutting corners given that structural design should place life-safety above cost. Engineering service fees used to be one to two-and-a-half percent of the project value. From personal experience, fees can now be one-half percent or less. Each silo of work has essentially become a commodity. That cost squeeze has had a significant impact on the ability of the supply chain to collaborate and communicate well.

ANSI/TPI 1 Scopes of Work CMs utilize engineering in their value proposition because they supply the primary load path resisting elements. As mentioned above, the building owner or the Building Designer/Structural Engineer does not contract with the truss manufacturer to provide their engineering and manufacturing expertise at the design development stage of the project. Thus, the designation “trusses by others.” To help navigate the potential for misunderstandings and misperceptions, and also for contract relationships, CMs rely on the scopes of work (SOW) outlined in ANSI/TPI-1 (TPI) Chapter 2, Standard

Responsibilities in the Design and Application of Metal-Plate-Connected Wood Trusses. This standard, as Brent Maxfield outlined in his STRUCTURE magazine articles (March and April, 2019) regarding wood trusses, works incredibly well. There are, however, instances where SOWs are exceeded due mainly to a breakdown in execution. As the Structural Building Component Association’s (SBCA) Executive Director since 1992 and a P.E., the author is in a unique position to regularly see interactions between Building Designers and CMs. He has witnessed countless successful applications of trusses inside a properly functioning supply chain. Just think about the billions of dollars of structural framing that are installed each year that utilize trusses. If truss use presented a systemic problem to the construction industry, SBCA would hear about it daily. In instances where the SOW breaks down, it is due primarily to a failure in some form of communication. Communication, specifically between Building Designers/Engineers and CMs, can be improved and can provide solutions to many of the problems that arise in the implementation of deferred submissions and review processes. For example, one solution that works well is Building Designers and General Contractors (GCs) who commit to work with a specific CM early in the project life cycle. Communication and collaboration at the design development stage of any project solve many of the problems that typically present themselves during a deferred submission review and revision process. Those problems typically result in costly rework by the Building Designer and CM. When communication starts early in a project, it is more likely to continue throughout that project to everyone’s benefit!

Scope of Work Creep SBCA’s Construction Industry Work Flow Initiative looks to map the way information, products and services flow throughout the construcution industry in a series of articles and graphical models.

34 STRUCTURE magazine

In recent years, commoditization of the construction supply chain has led to the need to do more with a smaller budget. This

drives SOW creep down the supply chain. A good illustration is provided through the many plans CMs receive today that need to be fixed, whether it be dimensions that do not close or load path details that do not work. To manufacture the trusses that will create the proper load path, CMs help by providing solutions (usually under a very tight timeline) to keep the project moving. Unfortunately, the SOW creep starts with the Architect, who often designs structures without full consideration of the load resisting elements. From there, in a cost-cutting effort, Building Designers/Engineers are bid out and forced to abandon their traditional SOWs in hopes of maintaining profitability through streamlining the engineering process. Digging into the dimensional weeds and providing specific load path details is very time-consuming. Consequently, many have created processes like “standard details” to speed up project completion. It is not out of the ordinary to see a partially complete detail on a set of “for construction” plans. As a result, CMs are forced to exceed their defined SOW in hopes of pleasing their customer, typically the GC. Often, information is wrong or missing from the design documents as provided to the GC. The GC delegates problems to the CM to interpret or, in many cases, fix. These “fixes” need to be done before the CMs can perform necessary tasks to model the project and ultimately deliver their product. The goal is to fix what needs to be fixed, given that project time pressure generally does not allow a lengthy review and approval process. This is exacerbated by the fact that the Architect and Building Designer/Engineer generally are not in a position where their original fees will tolerate additional costs.

Why the Truss Industry Functions as it Does

Serving Best Interests What is the best path forward to address the engineering and communication challenges outlined above? The Building Designer/Engineer is in the best position to have a full understanding of the intent of the building design in the context of

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CMs are compensated based on the volume of product delivered and not the engineering value they provide to the structure. Many Building Designers/Engineers know that CMs provide fixes “for free.” Getting the party that needs to fix the plans to get truss designs to work, so that they can be manufactured and out the door, can leave more money in the building design/engineering budget. Hence, it is useful if the fix work can be done where it does not negatively affect the construction budget. Behind the CM is the Truss Design Engineer (Truss Designer), who designs and seals individual trusses based on input parameter files that are interpreted by the CM’s Truss Technician. Like the CM, the Truss Designer relies on a defined SOW to perform his or her duties. The Truss Designer is a delegated engineer, removed from the specifics of the project and not as intimately knowledgeable about the project as the Building Designer/Engineer. The key to a successful project for both the CM and the Truss Designer is that

they “shall be permitted to rely on the accuracy and completeness of information furnished in the Construction Documents or otherwise furnished in writing by the Building Designer and/ or Contractor.” Mr. Maxfield’s suggestion of injecting additional engineering services, outside of the traditional SOW of the Building Designer, is a serious transfer of load path responsibility. This also jeopardizes the role and value that Building Designers/Engineers have within the supply chain. Several unintended consequences may occur if the CM takes on additional SOW that should rightfully be performed by the Building Designer/Engineer. If the CM hires or employs a building design engineer to design the roof system as Mr. Maxfield suggests, could they design the rest of the structure, rendering the traditional Building Designer/Engineer obsolete? Savvy Architects and GCs/Project Owners will, in a cost-cutting effort, quickly take advantage of this new load path engineering service and look to replace the Building Designer and use the CM’s engineer instead. CMs will be forced to solicit more of this work to cover the costs of Mr. Maxfield’s suggestions. The author has no problem with this strategy as long as CMs are compensated, at competitive engineering service rates, for any expansion of their SOW. However, serious consideration should be given to the unintended consequences of this suggestion for Building Designers and the structural engineering community.

MARCH 2020

expected load paths and to specify building conditions such as snowdrift, HVAC units, load path to footings, etc. The SOW, as outlined in TPI Chapter 2, should be followed regardless of who is performing the various design duties. It is important to engage the CM as early in the design development process as possible. This will immediately improve communication. Design reviews should be conducted and are necessary if the process is to be successful. Mistakes are occasionally going to be made. These are often due to communication and execution breakdowns because the silos of work are isolated and bid out to obtain the lowest cost. Several opportunities exist to improve communication between the engineering community and CMs, specifically with regard to using common specifications, contract language, standard details, and so forth. SBCA has a longstanding working relationship with NCSEA, most recently An excerpt from the collaborative work of NCSEA and SBCA to bring clarity to the IBC on working together on IBC lateral restraint and diagonal best methods for installing permanent lateral restraint and diagonal bracing of individual bracing related code change proposals. NCSEA’s point wood truss members. of view was also instrumental in TPI 1 Chapter 2 and Building Component Safety Information (BCSI). communication and execution, while also being adequately comOpportunities to collaborate remain. Successful construction projects pensated in the process, will lead to both excellent communication certainly require that the SOW of the Building Designer/Engineer, and much better construction quality.■ GC, CM, Truss Designer, truss installer, and so forth to be approURL references for graphics and text are live in the digital priately compensated in order to be successful. version of this article. CMs are a key supplier of load path resisting structural elements. The Building Designer/Engineer is in the best position to have a full Kirk Grundahl is the Executive Director of the Structural Building Components understanding of the intent of the building design in the context of Association (SBCA). (kgrundahl@sbcindustry.com) expected load paths. How both groups work together to improve ADVERTISEMENT–For Advertiser Information, visit STRUCTUREmag.org

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historic STRUCTURES Albion Bridge Collapse By Frank Griggs, Jr., Dist. M.ASCE, D.Eng., P.E., P.L.S.

quire Whipple (STRUCTURE, September 2005, January 2015) built one of his early bowstring iron

trusses over the Enlarged Erie Canal in Albion, New York, in 1848. He wrote of his early bridges: “About the same period (1840), my own attention was directed to the subject of iron bridges, and I designed the plan of my Patent Iron Trussed Bridge, patented 1841, and built an experimental bridge of 72 feet span, subsequently erected and now in use on the Erie Canal at Newville near Rome. I had, however, in the winter of 1841-42 pre-

Whipple Bridge with sidewalks similar to that at Albion.

vious to the erection of the above bridge on the canal, built a bridge of about 80 feet on First Street in Utica, which is the oldest iron bridge now in use on the canal; Trumbull’s first bridge, built a few months sooner than mine, having failed and been rebuilt since, as before stated.” “Soon after, this work of enlargement was suspended, but little the various bolts as circumstances required, the sad calamity would progress was made in the introduction of iron bridges on the canal never have happened.” for several years. In the meantime, I frequently urged on the Canal The Orleans County Fair had opened about a half-mile from the Commissioners the policy of having a few iron bridges built, for bridge, and the village was filled with residents and visitors. A young the purpose of more thoroughly testing their economy, that on the man from Brockport stretched a rope between two buildings, the resumption of the enlargement, sufficient knowledge might be had Mansion House and the Pierpont Dryer’s, on opposite sides of the for judging accurately as to the expediency of a more extensive adop- canal and on the west side of the bridge. He advertised that he would tion of them for the use of the canal. But it was a subject that could walk the rope across the canal at 5:00 PM, so a large crowd gathered bear putting off. at the bridge to witness the feat. The following newspaper account, In the spring of 1848, commissioner Hinds, with a discriminating one of many, described the collapse: liberality highly creditable, contracted with me for building an iron “The iron bridge near the rope was crowded with people on foot bridge at the village of Albion and also one at Holley.” and in wagons, the place being one of the best that could be found The cost for both bridges was $3,196. The Albion bridge served for a view of the rope, When the rope walker was just walking out, well until 1859 when it and all eyes were fixed upon was decided to replace it. It him, the bridge went down had been raised and set on with a crash, carrying into wood blocks on the west the water in a promiscuous side. One of the men who mass hundreds of people, built the bridge stated, “that including all ages and both through the neglect of the sexes, and with them, of state authorities it had been course, fell the horse and in an unsafe condition for wagons, iron timbers and over a year, and that within other materials of the a few days he had noticed a bridge. decided vertical inclination, It appears that the largest and spoken of it. Had proper number of people were on care been taken in underpinthe west side of the bridge, ning the bridge when raised, that being nearest to the and in examining the strucrope, and that went down Whipple Bridge at Holley, NY, built at the same time as the Albion Bridge. ture, tightening or loosening first, thus giving the falling 38 STRUCTURE magazine

mass of human beings and materials a sidewise as well as a down- calling extraordinary numbers of people upon it, and by the neglect of ward motion. The eastern section of the bridge falling last must have the officers and persons having charges thereof, had so been suffered buried many beneath its ruins. As the bridge was of iron, partly cast to remain for a long space of time, to wit, for three months past.” and partly wrought, it broke into sections They went on to condemn rope-walking, and fragments and went to the bottom of stating, the canal, burying, of course, everything “The jury are therefore of opinion that that went before its weight. The excitement the calamity cannot be justly attributed to which followed cannot be described – it may the neglect of the authorities of the village; perhaps be imagined…There were three but that it is the legitimate fruit of tolerteams upon the bridge, two carriages and a ating these pernicious exhibitions in the lumber team. The number of persons on the country, and the jury takes this occasion bridge is estimated at two to four hundred, to solemnly condemn the rope walking most of whom went into the water…The exhibitions now and lately so prevalent bridge was constructed about ten years ago, in the country, as injurious to the public we think on the Whipple plan. It was about morals, dangerous to human life, and pro60 feet long but not the full width of the ductive of evil only.” enlarged canal. It was shortly to be replaced Whipple did not testify and was not by a larger and better one, the materials for found liable. Usually, juries such as this which were already at Albion. It is safe to called in experts to testify as to why the Plaque describing the collapse. They erroneously call it a wooden bridge. assume that there was a defect in this strucbridge failed. Why they did not, in this ture, or that it has been neglected. These case, is unknown. The bridge was rebuilt iron bridges require attention, and at times the rods and bolts require and eventually replaced with a swing bridge that lasted until 1912. to be adjusted so that the weight will be equally distributed. When The 1859 Annual Report of the State Engineer and Surveyor stated, they are in order, they will sustain any number of human beings that “Said bridge was erected some ten years ago and was of the same style can find a place to stand upon them.” and pattern as several, erected at other places at about the same time, Another account stated, which are yet standing. The sectional area of its sustaining rods was “At 5:15, the gymnast began his walk but had not advanced more sufficiently great, if sound, to have supported more than four times than ten feet when an ominous creaking and groaning of the bridge was the load supposed to have been upon that part of the bridge which heard, and immediately a man was seen to leap far out into the water. first gave way when it fell. This being the case, its failure can only be The ropewalker threw him his balancing pole and then dropped down accounted for upon the ground that some of the rods were defective into a sitting posture upon the rope, in which position he remained but had escaped previous observation.” until the worst was over when he regained his starting place in safety. It is clear, therefore, that the bridge was poorly maintained, according This passed in a moment. In the next, the foot walk gave way under to the locals, and poorly inspected, according to the State Engineer. the pressure and was immediately followed by the remainder of the Why it was in the process of being replaced is not known. structure, carrying with it into the water about 500 persons, of whom Whipple usually designed his bridges to carry 100 psf uniformly a considerable number were forced under the remains.” spread across the sidewalks and roadway. In this case, the load was And another, very unevenly distributed with the cantilevered sidewalk breaking first “The rope was about two rods west of Main street bridge, an iron and west truss following. The same thing, unbalanced loading, had arched structure like most of the new canal bridges, which of course happened on a wooden bridge he designed to cross the Mohawk River offered an eligible standpoint from which to view the performance. at Cohoes to carry both carriage traffic and a cantilevered towpath for It was accordingly crowded with people and teams. The rope walker the Champlain Canal. The bridge did not fail but was significantly had got partway across the canal, when the bridge broke in two at criticized by the citizens of Cohoes and Waterford. It was destroyed the centre, precipitating all who were upon it into the middle of the in a fire and replaced with an iron bridge. canal. Men, women, children, horses and wagons, were all piled in an When the Enlargement was considered complete in 1862, a survey indiscriminate mass. The west half of the bridge went down first, and of the bridges reported that, of the 158 iron bridges, 116 were of the of course many of those who stood near the break were pitched off in Whipple Bowstring style indicating that the failure of the Albion such a way that when the east half of the bridge came down, which it Bridge did not deter the Canal Commissioners from adopting the did immediately, it fell upon and covered them up. Mr. Grant informs Bowstring as their standard iron bridge. Many bridges survived us that a pair of horses and a carriage full of people were crushed in until the New York State Barge Canal replaced the Erie Canal in this way so completely that nothing had been found of them…All the second decade of the 20th century and the old canal was filled eyes were immediately turned from the dexterity of the ropewalker in. Many bridges were removed at the time and rebuilt elsewhere to attend the call of suffering humanity. Very many were in a few in New York State. The author has restored bowstrings at Union moments rescued from the canal, but the falling structure wounded College, Boonville, and Vischer’s Ferry, and is working on many probably who never rose to the surface again.” a two-span Whipple Bridge at Claverack, all of which are A total of 14 were killed and many injured. An eight-man coroner’s in New York State.■ inquest was held on the next day and determined, “And the jurors aforesaid, upon their oaths aforesaid, do further Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having say that the aforesaid bridge, by reason of the raising thereof, and restored many 19 th Century cast and wrought iron bridges. He is now an the placing of same upon blocks of wood, though sufficiently safe Independent Consulting Engineer. (fgriggsjr@twc.com) for ordinary use, had become unsafe and dangerous upon occasions,

MARCH 2020

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SPOTLIGHT 181 Fremont

By Ibbi Almufti, S.E., and Nate Warner, P.E.

here is a common misconception among the general public that buildings designed according to modern building codes will not be damaged in an earthquake. Many building owners are similarly unaware that the seismic performance objectives outlined in the code are intended only to provide life safety for occupants; they do not prevent damage or ensure post-earthquake functionality. Significant financial losses and downtime for repairs can occur after an earthquake, which likely does not meet the expectations owners have for their investments. Jay Paul Company, the owners of 181 Fremont, envisioned a high-performance building with innovative sustainability strategies. After realizing that code-minimum earthquake performance did not align with their goals, they also chose to pursue a design strategy presented by Arup to achieve “beyond code” seismic resilience. This holistic “resilience-based” approach required identifying and attempting to mitigate all threats that could hinder re-occupancy and functionality. The building was designed to exceed the California Building Code (CBC) mandated seismic performance objectives and achieve immediate re-occupancy with limited disruption to functionality after a 475-year earthquake (the approximate recurrence interval for the CBC design basis earthquake). This is accomplished through enhanced design of both structural and non-structural components along with pre-disaster contingency planning. 181 Fremont is the third-tallest building in San Francisco, with the spire of the 56-story tower reaching a height of 802 feet. The lower 37 levels of the building are commercial office space, and the upper levels are condominiums. Arup is the Structural and Geotechnical Engineer of Record for 181 Fremont. The structure utilizes a perimeter steel mega-frame system to resist wind and seismic forces because a traditional concrete or steel core lateral force-resisting system is too slender, given the tower’s small footprint. Over the lower two-thirds of the building, the structural system features mega-braces spanning between nodes near ground level and at levels 20 and 37. A secondary system of perimeter special moment frames transfers individual floor demands up or down to the mega-nodes. In the residential levels above, the perimeter braces are made of large wide flange sections and a core of buckling-restrained braces (BRBs) acts as the secondary system. An inverted chevron braced frame provides lateral continuity between STRUCTURE magazine

the office and residential levels. Gravity loads are resisted by steel columns in the core and at the exterior as well as corner mega-columns. Transfer trusses wrapping around the building at level 3 carry all perimeter gravity loads to the mega-columns. Below ground, a 5-story basement is supported on a concrete mat with piles socketed into bedrock more than 200 feet below grade. Arup determined that using an integrated damping system would reduce seismic forces to facilitate enhanced performance and control wind vibrations in the tower. The mega-brace system provided an excellent opportunity to incorporate damping. Each mega-brace in the office levels consists of three parallel elements – a built-up-box primary brace in the middle with solid steel secondary braces on either side. Viscous dampers are introduced at one end of each secondary brace. The megabraces are restrained laterally at every floor to prevent buckling but slide freely along their length against polytetrafluoroethylene (PTFE) bearing pads. As the tower sways in wind or seismic events, elastic strains develop in the primary braces, causing them to lengthen or shorten between mega-nodes. Since the secondary braces are connected to the same mega-nodes, this activates the dampers and dissipates energy. BRBs were also introduced as a fuse to prevent damage to the braces, dampers, and mega-columns in the maximum considered earthquake (MCE). The system acts like a giant shock absorber to limit building drift and reduce floor accelerations. The structural design also features innovative uplifting corner mega-columns. As the building sways, significant tension demands develop in the mega-columns. To prevent damage, the columns are designed to uplift slightly (approximately 1 inch) at their bases in the MCE, which limits demands in the columns and foundation. The mega-columns are anchored by pre-tensioned rods tuned so that uplift does not occur in wind or smaller earthquake events. A shear key transmits shear from the columns into the foundation in the event of momentary uplift. Arup’s structural design saved approximately 2,700 tons of steel compared to a baseline design by another engineering firm – roughly 25% of the building weight – while satisfying the enhanced resilience objectives. Since the damping reduced seismic forces, steel tonnage could be decreased, which reduced the building’s

Arup was an Outstanding Award Winner for the 181 Fremont project in the 2019 Annual Excellence in Structural Engineering Awards Program in the Category – New Buildings over $100M. Courtesy of Jay Paul Company.

stiffness and increased its flexibility, leading to a cycle of material reduction. The increased flexibility decreased the seismic demands further, and the process was iterated until the design was tuned to meet the seismic and wind criteria. Since the tower is slender and lightweight, wind accelerations posed a challenge due to stringent criteria in the residential levels, but integrating viscous damping within the mega-frame also eliminated the need for a tuned mass damper near the top of the building. This resulted in additional material savings and freed up valuable real estate at the penthouse level. 181 Fremont was designed to avoid damage and achieve rapid recovery in the aftermath of a large earthquake, far exceeding code requirements and earning a REDi Gold rating for seismic resilience. The design strategy, including enhanced performance criteria for structural and non-structural components, was implemented with little cost premium, and will reduce the building’s overall life-cycle cost and environmental footprint significantly. While a resilience-based design approach extends beyond the typical purview of the structural engineer, 181 Fremont demonstrates how informed engineers are uniquely qualified to assist owners and stakeholders who desire better performing, “beyond code” buildings.■ Ibbi Almufti is an Associate Principal in the Advanced Technology + Research Group in Arup’s San Francisco office and the Project Manager for 181 Fremont. (ibrahim.almufti@arup.com) Nate Warner is an Engineer in the Structural Group in Arup’s San Francisco office and a Project Engineer for 181 Fremont. (nate.warner@arup.com)

MARCH 2020

NCSEA NCSEA News

National Council of Structural Engineers Associations NCSEA Corporate Members

ASSOCIATE MEMBERS

Nationally recognized bodies that are associated with the practice of structural engineering, regardless of location and membership, who are approved for this status by the Board of Directors.

American Wood Council Cantsink Fabreeka International, Inc. Insurance Institute for Business & Home Safety

International Code Council LNA Solutions Metal Building Manufacturers Association Precast/Prestressed Concrete Institute

PROSOCO/CTP Simpson Strong-Tie Steel Tube Institute USG Corporation â&#x20AC;&#x2019; Structural Solutions

AFFILIATE MEMBERS

Companies who provide supplies or services to structural engineers, including vendors of structural engineering applications software, insurance, and structural products used for construction.

ADAPT Corporation AISC Atlas Tube AZZ Galvanizing Services Bekaert Blind Bolt Cast Connex Corporation Chicago Clamp Company Cold-Formed Steel Engineers Institute Construction Tie Products, Inc. Copper Creek Companies, Inc. CoreBrace DBM VirCon DeWALT Dlubal Software, Inc. DuraFuse Frames Fox Rothschild LLP Freyssinet, Inc. Geopier

GIZA Steel Graitec GRM Custom Products Haselton Baker Risk Group Hayward Baker Headed Reinforcement Corporation (HRC) Hilti, Inc. Hubbell Power Systems (CHANCE) IAPMO Evaluation Service International Masonry Institute ITW Commercial Construction N. America Jordahl USA Inc. Kinemetrics Lindapter USA MeadowBurke Mitek Builder Products Myticon Timber Connectors Nelson Stud Welding New Millennium Building Systems

Nucor Peikko Pieresearch Post-Tensioning Institute Qnect RISA SE Solutions, LLC SidePlate Systems, Inc. SkyCiv Steel Deck Institute Steel Joist Institute Strand7 Taylor Devices Trimble Vector Corrosion Technologies Vitruvius Project Voss Engineering

SUSTAINING MEMBERS

Structural engineering firms, firms that employ structural engineers, or individual professional engineers practicing structural engineering. 4x Engineering Allan Klein PA Consulting Engineer ARW Engineers ASC Engineers, Inc. Barter & Associates, Inc. Burns & McDonnell Collins Engineers, Inc. Cowen Associates Consulting Engineers Criser Troutman Tanner Consulting Engineers CSA Engineering CSA Knoxville Davis Patrikios Criswell, Inc. DCI Engineers Deems Structural Engineering Degenkolb Engineers DiBlasi Associates, P.C. Dominick R. Pilla Associates DrJ Engineering, LLC

44 STRUCTURE magazine

ECM Engineering Solutions, LLC Gilsanz Murray Steficek Haskell Heyer Engineering Holmes Culley IBI Group Engineering Services (USA) Inc. Joe DeReuil Associates Katerra Krech Ojard & Associates Lance Engineering LLC LBYD, Inc. Mainland Engineering Consultants Corporation Mainstay Engineering Group, Inc. Mercer Engineering, PC Morabito Consultants, Inc. Mortier Ang Engineers

O'Donnell & Naccarato, Inc. Omega Structural Engineers, PLLC Professional StruCIVIL Engineers, Inc. Rimkus Consulting Group Ruby & Associates, Inc. Sanchez Civil Engineering Simpson, Gumpertz & Heger Stability Engineering Structural Design Professionals, PLLC Structural Engineers Group, Inc. STV, Inc. TEG Engineering, LLC TGRWA, LLC The Harman Group, Inc. Thornton Tomasetti Wallace Engineering WDP & Associates

News from the National Council of Structural Engineers Associations

Prepare for the PE Structural Exam with NCSEA

The Best Instructors. The Best Material. Available to you immediately when you register. NCSEAâ&#x20AC;&#x2122;s on-demand class provides the most economical PE Structural Exam Preparation Course available. The course includes 30 hours of instruction, 9 Vertical Sessions and 11 Lateral, and will give you preparation tips and problem-solving skills to pass the exam. All lectures are up-to-date on the most current codes with handouts and quizzes available. PLUSâ&#x20AC;Śstudents have access to a virtual classroom exclusively for course attendees! Ask the instructors directly whenever questions arise. This NCEES PE Structural Exam Preparation Course allows you to study at your pace but with instant access to the material and instructors. Several registration options are available; visit www.ncsea.com to register yourself or to learn more about special group pricing!

Call for 2020 Structural Engineering Summit Abstracts The 2020 NCSEA Structural Engineering Summit Committee is seeking presentations for the 2020 Summit in Las Vegas, NV, November 3-6, 2020. Ideal presentations are between 45 and 90 minutes, and deliver pertinent and useful information that is specific to the practicing structural engineer, in both technical and non-technical tracks. Submissions on best-design practices, new codes and standards, recent projects, advanced analysis techniques, management, business practices, the future of the profession, and other topics that would be of interest to practicing structural engineers are desired. Help make this year's Summit a sure bet! Submit your abstract March 20, 2020. Visit bit.ly/2020SESabstracts for more details.

Secure Training to Become a Second Respsonder

Register for the next NCSEA CalOES Safety Assesment Program on April 29, 2020 The California Office of Emergency Services (CalOES) Safety Assessment Program (SAP), hosted by NCSEA, is highly regarded as a standard throughout the country for engineer emergency responders. It is one of only two post-disaster assessment programs that will be compliant with the requirements of the Federal Resource Typing Standards for engineer emergency responders and has been reviewed and approved by FEMA's Office of Domestic Preparedness. Based on ATC-20/45 methodologies and forms, the SAP training course provides engineers, architects, and code-enforcement professionals with the basic skills required to perform safety assessments of structures following disasters. Register by visiting www.ncsea.com. This course is not included in the Live & Recorded Webinar Subscription. Doug Fell, P.E., is a CalOES Assessor, Coordinator, and Instructor. He is a licensed professional engineer (structural) in his home state of Minnesota as well as several other states. Doug is the managing principal of Structural Resource Center LLC. His practice includes structural engineering design and analysis for new and existing structures, structural assessments, forensic engineering, emergency response, development and review of safety programs, and project management services. Doug has responded to all types of emergencies and performed assessments all over the U.S. He was the lead structural engineer for the Minneapolis Metrodome roof collapse stabilization and return to service.

NCSEA Webinars

March 10, 2020

April 16, 2020

Sam Rubenzer, P.E., S.E.

Nicholas Miley, S.E.

Masonry Movement Joints This presentation addresses the movement characteristics of masonry wall systems, and compares the differences between architectural veneer and structural reinforced masonry. March 24, 2020

Structural Design and Embodied Carbon Meeting the goals of the UN Paris Agreement requires reductions in carbon emissions of at least 50% by 2030. This webinar will explain why structural engineers have an important and immediate role to play in meeting climate change targets.

Lintels for Masonry Walls Sam Rubenzer, P.E., S.E.

This webinar will explain firsthand how different design problems are being solved with masonry lintels. Courses award 1.5 hours of continuing education after the completion of a quiz. Diamond Review approved in all 50 states. M A R C H 2 02 0

SEI Update Students and Young Professionals

Congrats and Welcome to the Next Generation of Future Structural Engineering Leaders

Thanks to the SEI Futures Fund – www.asce.org/SEIFuturesFund – and generous donors in collaboration with the ASCE Foundation, we welcome the following SEI Student & Young Professional Scholarship recipients to participate and get involved in SEI at Structures Congress in St. Louis:

Students Mahesh Acharya, S.M.ASCE, Idaho State University Wael Aloqaily, S.M.ASCE, University of Delaware Adrianna Bailey, S.M.ASCE, University of South Alabama Sherlock Banks II, S.M.ASCE, Virginia Tech Julie Bouwens, S.M.ASCE, Michigan Technological University Richard Campos, S.M.ASCE, University of Oklahoma Briana Clark, S.M.ASCE, Rochester Institute of Technology Elizabeth DePaola, S.M.ASCE, University of Notre Dame Merhad Dizaji, S.M.ASCE, University of Virginia Emily Durcan, S.M.ASCE, Rensselaer Polytechnic Institute Ebenezer Fanijo, S.M.ASCE, Virginia Tech Mahmoud Faytarouni, S.M.ASCE, Iowa State University Yijie Gao, S.M.ASCE, Stanford University Moheldeen Hejazi, S.M.ASCE, Istanbul Technical University, Turkey Mary Juno, S.M.ASCE, University of Kansas Farid Khosravikia, S.M.ASCE, University of Texas at Austin Dayakar Naik Lavadiya, S.M.ASCE, North Dakota State University Min Li, S.M.ASCE, Colorado State University Jessica Lopez, S.M.ASCE, University of Illinois at Chicago Maria Camila Lopez Ruiz, S.M.ASCE, University of California, Berkeley Daniela Lugo Romero, S.M.ASCE, Princeton University Nicholas Maloney, S.M.ASCE, Drexel University Erin Mills, S.M.ASCE, Kansas State University Jordan Nutter, S.M.ASCE, University of Kansas Ryan Olsen, S.M.ASCE, Michigan Technological University Hongrak Pak, S.M.ASCE, Texas A&M University Sarah Puchner, S.M.ASCE, University of Alabama in Huntsville Selene Renes, S.M.ASCE, South Dakota State University Paul Ryan, S.M.ASCE, Florida Institute of Technology Babak Salarieh, S.M.ASCE, University of Alabama in Huntsville Stefanie Schulze, S.M.ASCE, Oregon State University Niyam Shah, S.M.ASCE, New Jersey Institute of Technology Jay Shah, S.M.ASCE, Texas A&M University Missagh Shamshiri, S.M.ASCE, University of Texas at Arlington Samuel Steiner, S.M.ASCE, Bradley University Muhammad Salman Tahir, S.M.ASCE, National University of Sciences and Technology, Islamabad, Pakistan Luke Timber, S.M.ASCE, University of Delaware Shree Tripathi, S.M.ASCE, Southern Illinois University Carbondale Xuguang Wang, S.M.ASCE, University of Toronto, Canada Haifeng Wang, S.M.ASCE, University at Buffalo

Young Professional Teaching Faculty Mohamad Alipour, Ph.D., A.M.ASCE, University of Virginia Mohammad Omar Amini, Ph.D., EIT, A.M.ASCE, Colorado State University William Collins, Ph.D., P.E., M.ASCE, University of Kansas Mohamed Elhassan, Aff.M.ASCE, University of Khartoum, Sudan Laura Micheli, Ph.D., EIT, A.M.ASCE, Catholic University of America Ravi Yellavajjala, Ph.D., P.E., M.ASCE, North Dakota State University Tanmay Ramani, EIT, A.M.ASCE, Arlington Heights, IL 46 STRUCTURE magazine

Young Professionals Stefanie Rae Arizabal, P.E., M.ASCE, San Francisco, CA Dan Bergsagel, C.Eng, M.ASCE, Brooklyn, NY Christopher Bird, EIT, A.M.ASCE, Washington, DC Jillian Cayer, EIT, A.M.ASCE, New York, NY Halle Doenitz, P.E., M.ASCE, Playa del Rey, CA Megan Hanrahan, EIT, A.M.ASCE, Los Angeles, CA Nicholas Heim, EI, A.M.ASCE, Cleveland, OH Hannah Hillegas, P.E., M.ASCE, Kansas City, MO Jaynee Jhaveri, P.E., M.ASCE, Dallas, TX Bishal Khadka, EIT, A.M.ASCE, Washington, DC Daniel Koothoor, EIT, A.M.ASCE, Sunnyvale, CA Maissoun Ksara, A.M.ASCE, Philadelphia, PA Katherine ONeill, EIT, A.M.ASCE, Sarasota, FL Matthew Powell, S.E., Aff.M.ASCE, Leeds, UK Rebecca Reifel, A.M.ASCE, St. Louis, MO Matthew Shelden, EIT, A.M.ASCE, Petaluma, CA Margaret Stambaugh, P.E., M.ASCE, St. Louis, MO Ashfaq Syed, P.E., A.M.ASCE, Dallas, TX

News of the Structural Engineering Institute of ASCE Learning / Networking STRUCTURAL ENGINEERING INSTITUTE

STRUCTURES CONGRESS 2020 St. Louis, Missouri I April 5-8

Search keywords in the technical program to plan which sessions to attend – www.structurescongress.org. Expert Special Sessions on: • Millennium Tower • FIU Bridge Collapse • Conceptual Design of Bridges and Buildings • Professional Liability Case Study Marathon • Confidential Reporting on Structural Safety in the U.S.

Advancing the Profession

Embodied Carbon Reductions Endorsed by SEI Board of Governors – the SE 2050 Movement

On December 16, 2019, the SEI Board of Governors (BOG) unanimously voted to endorse the SE 2050 Challenge issued by the Carbon Leadership Forum stating: “We, the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE), support the vision and ambition of the SE 2050 Challenge. We, as a leading structural engineering organization in the United States, recognize the need for coordinated action across our profession to achieve the globally stated goal of net-zero carbon by 2050.” Learn more at www.seisustainability.org or www.se2050.org. And attend Climate Change and our Structural Systems: SE 2050 Initiative at Structures Congress in St. Louis.

SEI Online

ASCE 7-16 Errata

Update with Batch #3 (effective January 16, 2020) now available; Access Batches 1-3 from ascelibrary.org.

SEI News Read the latest at www.asce.org/SEINews SEI Standards Visit www.asce.org/SEIStandards to View ASCE 7 development cycle SEI on Twitter

Errata

SEI on Facebook Follow us: @SEIofASCE

SEI Standards Supplements and Errata including ASCE 7. See www.asce.org/SEI-Errata. If you would like to submit errata, contact Jon Esslinger at jesslinger@asce.org. M A R C H 2 02 0

CASE in Point Did you know? CASE has tools and practice guidelines to help firms deal with a wide variety of business scenarios that structural engineering firms face daily. Whether your firm needs to establish a new Quality Assurance Program, update its risk management program, keep track of the skills young engineers are learning at each level of experience, or need a sample contract document – CASE has the tools you need! CASE has several tools available for firms to use to enhance their internal policies and procedures – from office policy guides to employee reviews: Tool 1-3 Tool 2-2 Tool 2-3 Tool 2-5 Tool 4-3 Tool 5-3 Tool 5-5

Sample Policy Guide Interview Guide and Template Employee Evaluation Templates Insurance Management Sample Correspondence Guidelines Managing the Use of Computers/Software Project Management Training

You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

Save on CASE Membership!

Can you ever really be too successful? Keep your business thriving – no matter what your competition or the economy is doing – and say YES to membership in ACEC’s Coalition of American Structural Engineers (CASE). An “Association within an Association” that complements your ACEC National benefits. CASE, the oldest of ACEC’s four discipline-specific Coalitions, is a professional community for, of, and by structural engineers who want relevant, useful information – on BIM, international building codes, risk management, and more – to run their businesses better. Join CASE today, and you’ll qualify for: • Education: CASE offers a track of 3 dedicated education sessions at both the ACEC Fall Conference and Annual Spring Convention to keep members current with best practices and trends in structural engineering. As a member, you will also receive a discounted rate to ACEC webinars focused on structural engineering issues. CASE also provides education sessions at the AISC Steel Conference and the ASCE-SEI Structures Congress. • Resources: Coalition members get free access to over 145 contracts, tools, and publications (a total value of over $5,000!). CASE developed over 70 documents geared toward structural engineering firms. • Advocacy: Your voice matters! Coalition members are often the first ones contacted to share their expertise with Congress and government agencies in response to current legislation and relevant regulatory agendas. Save $75 off your first year’s dues through June 30, 2021! Join CASE by March 31, 2020, and get 15 months for the price of 12! Interested? Contact CASE’s Executive Director, Heather Talbert, at 202-682-4377 or email her at htalbert@acec.org.

Manual for New Consulting Engineers An HR Favorite for New Hires

ACEC’s best-seller, “Can I Borrow Your Watch” A Beginner’s Guide to Succeeding in a Professional Consulting Organization, offers new engineers a head start in the business of professional consulting. This essential guide is tailored to the unique needs of engineering firms, and the skills and experiences rookie consultants need to be successful in a large organization, including: • Proposal Preparation • Financial Management • Client Relationships • Project Management • Staff Management With over 140 pages of consulting expertise, this resource is the perfect addition to any new staffer’s welcome pack or in-house orientation. It can even be a useful resource for more seasoned engineers looking to refine their skills. To order this book, go to www.acec.org/bookstore. Bulk ordering is available; for more information, contact Maureen Brown (mbrown@acec.org). 48 STRUCTURE magazine

News of the Coalition of American Structural Engineers CASE Practice Guidelines Currently Available

CASE 962-H – National Practice Guideline on Project and Business Risk Management This guideline is intended to assist structural engineering companies in the management of risk associated with projects and to provide commentary regarding the management of risk associated with business practices. The guideline is organized in two sections that correspond with these two areas of risk, namely Project Risk Management and Business Practices Risk Management. The goal of the guideline is to educate and inform structural engineers about risk issues so that the risks they face in their practices can be effectively mitigated, thus making structural engineering firms more successful. Structural Engineer’s Guide to Fire Protection This publication is intended for structural engineers with no prior experience or training in fire protection engineering. It is a comprehensive and concise treatment of prescriptive and performance-based methods for designing structural fire protection systems in an easy to understand format.

CASE 504 – Proposal Preparation Spreadsheet The CASE Proposal Preparation Spreadsheet was developed to assist project managers and administrators in developing cost proposals for a project. The spreadsheet may be easily customized for any organization or project type. It also may be used as a checklist to see that all phases of a project are adequately staffed. CASE 976-A – Commentary on Value-Based Compensation for Structural Engineers Value-Based Compensation is a means to step out of the ordinary and establish your value to the team. Value-Based Compensation is based on the concept that there are specific services, which may vary from project to project, that provide valuable information to the client and whose impact on the success of the project is far in excess of the prevailing hourly rates. Value-Based Compensation is based on the increased value or savings that these innovative structural services will contribute to the project. As a result, the primary beneficiary of an innovative design or a concept is the owner, but the innovative engineer is adequately compensated for his knowledge and expertise in lieu of his time.

You can purchase these and the other CASE Risk Management Tools at www.acec.org/bookstore.

Donate to the CASE Scholarship Fund! The ACEC Coalition of American Structural Engineers (CASE) is currently seeking contributions to help make the structural engineering scholarship program a success. The CASE scholarship, administered by the ACEC College of Fellows, is awarded to a student seeking a bachelor’s degree, at a minimum, in an ABET-accredited engineering program. Since 2009, the CASE Scholarship program has given $31,000 to help engineering students pave their way to a bright future in structural engineering. We have all witnessed the stiff competition from other disciplines and professions eager to obtain the best and brightest young talent from a dwindling pool of engineering graduates. One way to enhance the ability of students to pursue their dreams to become professional engineers is to offer incentives in educational support. Your monetary support is vital in helping CASE and ACEC increase scholarships to those students who are the future of our industry. All donations toward the program may be eligible for a tax deduction, and you do not have to be an ACEC member to donate! Contact Heather Talbert at htalbert@acec.org to donate.

Follow ACEC Coalitions on Twitter – @ACECCoalitions. MARCH 2020

structural FORUM FEMA P-807 for Soft-Story Retrofits Technical Considerations for Engineers By Bruce F. Maison, P.E., S.E.

rescriptive Performance-Based Design: An Innovative Approach to Retrofitting Soft/Weak-Story Buildings (STRUCTURE, September 2019) describes the approach contained in the Federal Emergency Management Agency (FEMA) P-807 guideline. P-807 is a method to retrofit a weak first story of wood buildings to mitigate side-sway pancaketype collapse, as depicted in the Figure. The hazard posed by such buildings was underscored by their damage in the 1989 Loma Prieta earthquake affecting the San Francisco Bay area, as well as in the 1994 Northridge earthquake in the Los Angeles region. Some cities in California have enacted ordinances mandating retrofit of soft-story buildings. Performance-based engineering (PBE) is an evolving paradigm in earthquake engineering in which the goal is to proportion a building to meet specific, predictable performance requirements. The benefits of PBE are generally recognized, but predicting structural performance is challenging. This contrasts with traditional building codes that are mostly prescriptive and do not require explicit performance prediction. Prescriptive design is simpler than PBE. The “prescriptive performance-based design” advocated in the article could be appealing since it suggests the benefits of PBE are within the relative ease of prescriptive design. However, P-807 is a novel approach that quantifies performance in probabilistic terms. Implicit is the notion that the probability associated with actual building response can be estimated with reasonable accuracy. P-807 has not been thoroughly peer-reviewed and, as such, caution must be exercised regarding its use – especially on the efficacy of performance prediction. The shake table experiments mentioned in the article, in fact, do not validate P-807. (Please see reference 2 in the online version of this article for an explanation.) It is essential to recognize that P-807 has not been vetted through a rigorous ANSI type consensus process typically used in the development of codes and standards. It also lacks a formal mechanism for revision as necessitated by emerging new information. Hence, P-807 is not an industry consensus method and due diligence must be performed

50 STRUCTURE magazine

before proposing it as an alternative method in building codes. The author, in conjunction with several other San Francisco Bay Area practicing structural engineers, performed an independent review of P-807. Below is a summary of the technical aspects that engineers should be aware of before deciding on its use. • Component (e.g., wood structural panel) lateral load-drift relationships (i.e., backbone curves) are not indicative of those expected for wood-frame buildings – most notably by having relatively limited ductility. • The drift acceptance criteria do not reflect a severe condition such as near-collapse and, hence, do not signify damage states of practical value. Depiction of a collapse mechanism for a San Francisco type • Using the default perfor1920s soft-story apartment building. mance objective can result in first story retrofits having lateral strength greater than those from questionable accuracy for predicting actual IBC building code for new construcbuilding performance within a PBE context. tion. The default is a 20% probability That is, it cannot reliably compute the probabilof exceeding the drift associated with ity that a particular seismic intensity will result near-collapse under a maximum conin a meaningful state of damage for a specific sidered earthquake (20% POE under building. Engineers are encouraged to read refer100% MCE). ences 6 and 7 in the online version of • Using a relaxed performance objective this article that serves as the basis of such as that in the San Francisco softthe brief summary presented here.■ story building retrofit ordinance (30% POE under 50% MCE) can result The online version of this article in first-story retrofits having lateral contains references. Please visit strength smaller than that from tradiwww.STRUCTUREmag.org. tional retrofit practice (e.g., 75% code per IEBC Appendix Chapter A4). Bruce F. Maison is a Consulting Engineer • It used a single suite of earthquake practicing in El Cerrito, California. He is ground motions to account for all an active member of the Existing Buildings site classes (rock, soil, etc.), and, as a Committee of the Structural Engineers consequence, P-807 is likely to overAssociation of Northern California (SEAONC). estimate the ruggedness of buildings He is also a member of the ASCE/SEI located on soil sites (near-collapse Committee responsible for the performanceprobability too low). based engineering standard ASCE 41, Seismic It was concluded that P-807 might be an Evaluation and Retrofit of Existing Buildings. efficient methodology for relative ranking (maison@netscape.com) and selection of retrofit designs, but it has

MARCH 2020

Learn about the changes in ACI 318-19 Attend one of 25 public seminars titled “ACI 318-19: Changes to the Concrete Design Standard” being held throughout the United States this spring. FREE COPY OF ACI 318-19 FOR ALL SEMINAR ATTENDEES

With ACI 318-19 now 100 pages longer than the previous edition, two industry experts will walk seminar attendees through the new provisions and present major technical changes.

SEMINAR LOCATIONS Albany, NY Baltimore, MD Chicago, IL (Rosemont) Cleveland, OH Dallas, TX Denver, CO Des Moines, IA Detroit, MI (Farmington Hills) Emeryville, CA Houston, TX Indianapolis, IN Little Rock, AR Miami, FL Minneapolis, MN Nashville, TN New Brunswick, NJ

New Orleans, LA Pittsburgh, PA Portland, OR Raleigh, NC Richmond, VA San Diego, CA Savannah, GA St. Louis, MO Tampa, FL

Visit concrete.org/ACI318 for a complete seminar description and all registration information

THE NEW RISA-3D: DELIVERING MORE POWER AND A GREAT NEW INTERFACE

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