STRUCTURE magazine | March 2014 by structuremag

March 2014 Seismic

A Joint Publication of NCSEA | CASE | SEI

STRUCTURE

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CONTENTS

FEATURES 38

March 2014

The Transformation of the Historic First National State Bank Building Part 2

DEPARTMENTS

By D. Matthew Stuart, P.E., S.E., SECB and Ed D. Cahan, P.E., S.E.

The adaptive reuse of the First National State Bank Building is an exemplary rehabilitation project which would not have been possible without creative, yet simple and constructible, structural engineering solutions for the complex renovation changes. This article discusses life safety improvements and enhancements to the 12th floor and roof spaces.

43 Special Section

60 Legal Perspectives The Hyatt Regency Disaster Revisited

By Matthew R. Rechtien, P.E., Esq.

66 Engineer’s Notebook Advanced Seismic Systems and Code Evolution

By Jerod G. Johnson, Ph.D., S.E.

An Overall Strong Year for Steel Expected, Companies Say

67 Spotlight UC Berkeley California Memorial Stadium

By Larry Kahaner

By David Friedman, S.E., et al.

The steel fabrication business is predicting a strong 2014, which bodes well for the building industry. Firms continue to innovate and improve offerings across all sectors.

74 Structural Forum Are Sustainable Structures Compatible with Common Sense? By Bill Addis, Ph.D.

COLUMNS 7 Editorial

Lateral Load Path and Capacity of Exterior Decks

By Stan R. Caldwell, P.E., SECB

32 InSights

Land of Enchantment

3D In-Model Shop Drawing Review

By Jon A. Schmidt, P.E., SECB

By Mark Hershberg, P.E., S.E., et al.

10 Structural Design

34 Historic Structures

Design for Blast and Seismic

Trenton Bridge

By Frank Griggs, Jr., P.E.

By Monique Head, Ph.D., et al.

By Benton Johnson, P.E., S.E., et al.

By Benjamin Schafer, Ph.D., P.E.

Testing Tension-Only Steel Anchor Rods Embedded in Reinforced Concrete Slabs

By W. Andrew Fennell, P.E., CPEng, SECB, et al.

22 Structural Sustainability

The CFS-NEES Effort

52 Structural Testing

Nonlinear Analysis in Modern Earthquake Engineering Practice

58 Just the FAQs Reinforcing Shear Walls in Seismic Zones

By Max L. Porter, Ph.D., P.E. and Gregory P. Baenziger, P.E.

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, C 3 Ink, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

STRUCTURE magazine

March 2014

ON THE COVER

The Pennoni Philadelphia Structural Division developed innovative solutions for the transformation of the historic First National State Bank Building in Newark, New Jersey into a hotel. The conclusion of the article on this project on page 38 discusses life safety improvements and enhancements to utilization of the 12 th floor and roof spaces. A Joint Publication of NCSEA | CASE | SEI

16 Structural Performance

STRUCTURE

By Eric L. Sammarco, P.E., et al.

24 Building Blocks

8 Advertiser Index 62 Resource Guide (Software Updates) 68 NCSEA News 70 SEI Structural Columns 72 CASE in Point

By Brian J. Parsons, et al.

9 InFocus

Timber Tower Research Project

IN EVERY ISSUE

28 Structural Forensics

Investing in the Future of Our Profession

March 2014 Seismic

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STEEL PRODUCTS

GRADES

SIZES

SERVICES

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Electronic Data Interchange (EDI)

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Editorial

Investing new trends, new techniques in theand Future current industry of Our issues Profession By Stan R. Caldwell, P.E., SECB, F. ASCE, F.SEI, F.AEI

ave you ever given any thought to the future of our profession? Not the profession next month or next year, but rather the profession that we will leave to our successors. Driven by automation, globalization, contractor-led procurement, and many other forces, our profession is changing like never before. Will we sit quietly on the sidelines and accept whatever role society crafts for us, including possible obsolescence, or will we actively work to steer the profession toward a thriving future? The Structural Engineering Institute (SEI) of ASCE, with more than 25,000 diverse members and more than 150 active committees, has deliberately chosen the latter, proactive approach. Each year, SEI undertakes a broad range of activities that advance the art, science, and practice of structural engineering – helping to create a brighter future for the profession. Because some of these activities fall outside the constraints of SEI’s annual operating budget, philanthropic support is needed from individuals and organizations. Launched in November 2013 as the successor to the SEI Endowment Fund, the SEI Futures Fund leverages philanthropic contributions to support vital SEI activities and initiatives. Managed by its own Board of Directors, the SEI Futures Fund has established four strategic areas for funding. Gifts to the SEI Futures Fund will help: • Promote student interest in structural engineering. • Support younger member involvement in SEI. • Provide opportunities for professional development. • Invest in the future of the profession. These priorities will ensure a strong future for the structural engineering profession by attracting the best and brightest students, creating pathways to encourage creativity and innovation by young professionals, and building a vibrant culture of lifelong learning for all of our members. Most importantly, the SEI Futures Fund will invest in new opportunities to secure and expand the role of structural engineers as stewards of the built environment. For example, SEI recently released a ground-breaking report, A Vision for the Future of Structural Engineering and Structural Engineers: A Case for Change, with ten specific initiatives designed to help structural engineers evolve as leaders and innovators in the global economy of the future. The report is currently available as a free download on the SEI website at www.asce.org/SEI. In my view, this is a “must read.” As SEI moves forward with some of the initiatives in the report, the SEI Futures Fund will likely provide critical financial support. As I write this article, the SEI Futures Fund is less than two months old and its fund-raising campaign has not yet formally commenced. Nevertheless, its message and mission have been spread by word of mouth and are already resonating among those structural engineers with a long view of our profession. An engineer in Chicago has pledged $20,000 over the next three years. Another, in Washington, has pledged $10,000 over that period. An engineer in Dallas has mailed in a check for $7,500. Gifts of $1,000 or more have been made online or by mail from engineers in Hawaii, California, Florida, Delaware … and the list is growing. Organizations have stepped up as well. An engineering firm in New York has pledged $5,000 annually. Another, in Rhode Island, has pledged $2,000 annually. You get the idea; the SEI Futures Fund is clearly gaining momentum. STRUCTURE magazine

ASCE Foundation President Michael Goodkind and SEI Futures Fund Chair Stan Caldwell sign a Memorandum of Understanding.

SEI has the potential to bring tremendous benefits, to you and to our profession, with financial support from the SEI Futures Fund. With your help, the activities and initiatives supported by the SEI Futures Fund will become increasingly significant over time. Please join me in building the SEI Futures Fund. Your gift, along with those of your colleagues, will allow robust financial support in all of the strategic areas outlined above. The future of our profession is not assured, so please consider giving generously. All gifts will be sincerely appreciated and gratefully acknowledged. Your gift can be made by sending a check to: SEI Futures Fund, c/o ASCE Foundation, 1801 Alexander Bell Drive, Reston, Virginia, 20191. Alternatively, you can make your gift securely online. Please be assured that 100% of your gift will benefit the SEI Futures Fund, free of any administrative burdens. SEI is a part of ASCE, a 501(c)(3) tax exempt organization. The SEI Futures Fund is partnering with the ASCE Foundation, also a 501(c)(3) tax exempt organization, to leverage their expertise in managing philanthropic gifts. Your contributions are therefore tax deductible to the fullest extent of the law, and an acknowledgement with a receipt will be provided. Please check with your tax advisor as appropriate. For questions about making a gift, contact Natalie Zundel by email at nzundel@asce.org or by telephone at 703-295-6347. For more information about the SEI Futures Fund, or to make a gift online, visit www.asce.org/SEIFuturesFund.▪ Stan R. Caldwell, P.E., SECB, F. ASCE, F.SEI, F.AEI (www.stancaldwellpe.com), is a consulting structural engineer in Plano, Texas. He currently serves as the chair of the SEI Futures Fund Board of Directors. He also serves as a member of the SEI Board of Governors, the SECB Board of Directors, and the Structural Engineering Licensure Coalition Steering Committee.

March 2014

ADVERTISER INDEX

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Bentley Systems, Inc. ............................. 75 Canadian Wood Council ....................... 65 Construction Specialties ........................ 11 CTP Inc. ............................................... 40 CTS Cement Manufacturing Corp........ 35 Design Data .......................................... 49 Enercalc, Inc. .......................................... 3 Engineering International Inc................ 54 ESAB Welding and Cutting Products .... 51 Fyfe ....................................................... 57 Hayward Baker, Inc. .............................. 55 Integrated Engineering Software, Inc..... 41 Independence Tube Corporation ............. 6

ITW BCG Construction Hardware....... 46 ITW BCG TrusSteel.............................. 48 ITW Red Head ..................................... 37 JEC Americas ........................................ 53 KPFF Consulting Engineers .................. 50 LNA Solutions ........................................ 8 MMFX Steel Corporation of America ... 44 New Millennium Building Systems ....... 47 Nucor Vulcraft Group ........................... 42 Powers Fasteners, Inc. .............................. 2 QuakeWrap ........................................... 19 Ram Jack Systems Distribution ............. 26 RISA Technologies ................................ 76

Editorial Board

S-Frame Software, Inc. ............................ 4 SidePlate Systems, Inc. .......................... 45 Simpson Strong-Tie......................... 15, 21 Soc. of Naval Arch. & Marine Eng. ....... 17 StructurePoint ....................................... 59 Struware, Inc. ........................................ 18 Taylor Devices, Inc. ............................... 13 USP Structural Connectors ................... 20

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STRUCTURE ® (Volume 21, Number 3). ISSN 1536-4283. Publications Agreement No. 40675118. Owned by the National Council of Structural Engineers Associations and published in cooperation with CASE and SEI monthly by C3 Ink. The publication is distributed free of charge to members of NCSEA, CASE and SEI; the non-member subscription rate is $75/yr domestic; $40/ yr student; $90/yr Canada; $60/yr Canadian student; $135/yr foreign; $90/yr foreign student. For change of address or duplicate copies, contact subscriptions@STRUCTUREmag.org. Any opinions expressed in STRUCTURE magazine are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, C3 Ink, or the STRUCTURE Editorial Board. STRUCTURE® is a registered trademark of National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

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A K E E S A F E T Y C O M PA N Y

March 2014

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inFocus

Land new trends,of newEnchantment techniques and current industry issues By Jon A. Schmidt, P.E., SECB

elcome to a rare instance when this column has nothing to do with engineering or philosophy! Ever since I spent several days in New Mexico with my family during a vacation last summer, I feel like the abundant sunshine and thin, dry air have put me under some kind of spell. I find myself eager to return as soon as possible and seriously considering the prospect of retiring there someday. In the meantime, I have read nearly a dozen books – travel guides, geography and biology, customs and culture, history, and even (unusual for me) fiction – all in an effort not only to acquire information, but also to gain an intangible sense of the place. For example, New Mexico’s flag is one of the simplest and most distinctive of the fifty states, with a red symbol centered on a yellow field. Reflecting the region’s unique history and ethnic makeup, the colors are those of old Spain, while the symbol comes from the Native American people of Zia Pueblo. Dr. Harry P. Mera, a physician and anthropologist who lived in Santa Fe, created the contest-winning design, which the legislature officially adopted in 1925. He was familiar with the symbol from its frequent appearance on clay pottery fashioned by the Zia Pueblo residents. The symbol itself is rich in meaning, primarily representing the sun as the “Giver of all Good Gifts.” The official salute to the state flag further associates it with “perfect friendship among united cultures.” However, its significance is much deeper for the Zia people. Four is their sacred number, and the symbol includes four groups of four rays, which correspond to the four principal directions – north, south, east, and west; the four seasons of the year – spring, summer, autumn, and winter; the four phases of the day – morning, noon, evening, and night; and the four stages of life – childhood, youth, adulthood, and old age. The symbol thus encompasses all of space and time with their ongoing cycles, including human existence, joined together by a circle of love that has no beginning or end. It also invokes four specific obligations that the Zia people recognize: a strong body, a clear mind, a pure spirit, and a devotion to the welfare of one’s people. Alternatively, Christians may perceive it as a form of cross, and the obligations in reverse order are reminiscent of the commandment to love the Lord your God with all your heart, with all your soul, with all your mind, and with all your strength. Of course, spreading Christianity to Native Americans was one of the professed motivations for the arrival of the Spanish in New Mexico, which predates the efforts by the English along the Atlantic coast that are far more prominent in American textbooks. In fact, the name of the territory is considerably older than that of the country to our south, which was known as New Spain until its independence in 1821. Previously, “Mexico” referred primarily to the fabulous capital city of the Aztecs, conquered three hundred years earlier. Explorers were hoping to find similar riches to the north – i.e., a new Mexico. Although nothing so spectacular ever materialized, the name stuck, and the state’s official seal is delightfully clear in its symbolism – it depicts an American bald eagle taking a Mexican golden eagle under its wing. STRUCTURE magazine

The New Mexico State Capitol’s rotunda floor features the state seal within the Zia symbol.

Juan de Oñate led the first group of European settlers up the Rio Grande – then known as Rio del Norte, or River of the North – in 1598, nine years before Jamestown. The positive and negative aspects of this event are meticulously described in an epic poem written by one of Oñate’s subordinate officers, Gaspar de Villagrá, and published in Spain in 1610 as La Historia de la Nueva Mexico. That same year, Governor Pedro de Peralta founded the city of Santa Fe – a decade before the Pilgrims arrived at Plymouth Rock, making it the oldest capital in the United States. It is also the highest, at an elevation of 7,260 feet above sea level. Visiting state capitol buildings was actually the main objective that brought us to New Mexico in the first place. We have been on a quest to see all fifty of them since my son was six and my daughter was barely a year old – they are fourteen and nine now – and this was state capitol number thirty-eight. Santa Fe is home to the nation’s oldest continuously occupied public building, the Palace of the Governors (1610), and one of its newest state capitols (1966), which residents affectionately call “the Roundhouse” since it is the only such facility in the country with a primarily circular shape. I was impressed with the apparent efficiency of its unique layout, as well as its distinctive Territorial architecture, which is the only style permitted by law in the historic portion of the city other than the even more traditional adobe look of Pueblo Revival. Needless to say, New Mexico has made an indelible impact on me. Time will tell if its magic induces multiple future visits and ultimately draws me to make the Land of Enchantment my home.▪ Jon A. Schmidt, P.E., SECB (chair@STRUCTUREmag.org), is an associate structural engineer at Burns & McDonnell in Kansas City, Missouri and chairs the STRUCTURE magazine Editorial Board.

March 2014

Structural DeSign design issues for structural engineers

ver the past few decades, significant advances have been made in the areas of earthquake engineering and seismic design. A growing database of strong motion records, refined ground motion attenuation relationships, and probabilistic seismic hazard methodologies have led to an improved design basis for seismic events. In addition, multi-scale component-level and system-level research have given rise to innovative energy dissipation and kinematic isolation concepts, enhanced structural detailing provisions, and performance-based design methodologies. Perhaps most importantly, many of these technological advances are currently being implemented in practice and taught in colleges and universities. The protection of buildings against airblast due to explosions has been a national interest for many years. For at least half a century, the U.S. Government has invested in physical testing, research, and development efforts focused on wartime defense scenarios involving both nuclear weapons and high-explosive detonations. Moreover, the heavy industrial sector has long been concerned with damage and injury mitigation from accidental explosions occurring in petrochemical facilities. With the rise in international and domestic terrorism, the vulnerability and state-of-security of the nation’s buildings and infrastructure have become national concerns. As a result, interest in blast effects and protective design has increased among the general structural engineering community. Recent public domain research has led to a number of significant technological advances related to blast threat mitigation and anti-terrorist/force protection (ATFP) design. Blast-resistant design guidance is available in specialized building design standards, handbooks, and guidance documents such as ASCE 59-11 Blast Protection of Buildings, UFC 4-010-01 DoD Minimum Anti-Terrorist Standards for Buildings, the compilation text entitled Handbook for Blast-Resistant Design of Buildings, and FEMA 427 Primer for Design of Commercial Buildings to Mitigate Terrorist Attacks. However, unlike earthquake engineering, the integration of fundamental blast-resistant analysis/design principles with the general structural engineering community and major college curricula has been slow at best. Consequently, understanding that blast and seismic are both dynamic phenomena, many structural engineers are left drawing from their seismic knowledge when faced with a blast-resistant analysis and/or design scenario. Extrapolating in this manner is ill-advised because an adequate seismic design does not necessarily imply adequacy from a blast-resistant design perspective. Recognizing the risk presented by

Design for Blast and Seismic Acknowledging Differences and Leveraging Synergies By Eric L. Sammarco, P.E., M. ASCE, Cliff A. Jones, P.E., M. ASCE, Eric B. Williamson, Ph.D., P.E., M. ASCE and Harold O. Sprague, P.E., F. ASCE

10 March 2014

unqualified engineers performing blast-resistant design, the United Kingdom has taken a proactive approach by initiating the Register of Security Engineers and Specialists (RSES) through the Institution of Civil Engineers. The Register aims to ensure that registrants have achieved a recognized competence standard, accepted a code of ethics, and are committed to continuing professional development. While no such specialized register or certification currently exists in the U.S., it is important for structural engineers to be mindful of the fact that design adequacy for one load case does not guarantee design adequacy for the other. As the title of this article suggests, there are important differences between seismic-resistant design and blast-resistant design, despite the dynamic nature of both. By acknowledging the differences and leveraging the synergies between the two design methodologies, structural engineers can improve the overall efficiency, effectiveness, and robustness of their building designs. This is not a new topic; however, past treatments have typically been cursory and fragmented. This article aims to provide a relatively comprehensive overview of the blast versus seismic topic by addressing demand, system response, component response, and design synergies in a practical way that will hopefully benefit the structural engineering community.

Differences from a Demand Perspective Aside from the dynamic nature of both types of loads, earthquake ground motion characteristics are markedly different from those of a blast-induced overpressure history. Figure 1 ( page 12) shows a comparison between a normalized ground acceleration record from the 1989 Loma Prieta earthquake and a normalized freefield overpressure history from a high-explosive (HE) detonation. The duration of an unconfined blast pulse from a high-explosive detonation is generally on the order of microseconds to milliseconds, whereas the strong motion duration of a typical earthquake record is generally on the order of several seconds and can last over a minute. In Figure 1, note the cyclic nature of the ground motion acceleration record, which includes multiple peaks. Conversely, for the blast overpressure history, note the nearly instantaneous rise to peak overpressure followed by a rapid decay to a subatmospheric “negative pressure” condition. Blast overpressure is often reported as a gauge pressure relative to ambient atmospheric conditions. Thus, a negative gauge pressure condition represents sub-atmospheric pressure, resulting in a temporary suction effect. Earthquake demand input is kinematic in nature, where near-field seismic waves excite continued on page 12

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to deterministic, risk-informed approaches for predicting design-basis demand input.

Differences at the System Level

Figure 1: Comparison of typical demand input.

the foundation of an affected structure and engage its entire lateral force resisting system (LFRS). Seismic forces are derived from accelerating the mass of an affected structure and inducing relative displacements between structural components. In contrast, blast demand input is force-based in nature. A high-explosive detonation gives rise to a shock wave that impinges upon exposed surfaces of nearby structural components, imparting a highly transient reflected pressure pulse. The shock wave does not engage the affected structure’s entire LFRS at once; rather, its effect is phased in time and highly variable in magnitude. Finally, an important distinction can be made with regard to the origin of both demand inputs. Earthquakes are a naturally occurring phenomenon. Historical strong motion data and fault studies are As a quick aside, ground motion spatial incoherence and soil-structure interaction effects are typically neglected in practice – high-importance structures such as nuclear power plants and mission-critical military facilities being exceptions. It is encouraged that such effects, particularly soil-structure interaction, be considered more frequently in practice as they are phenomena that do exist in reality and can strongly influence structural response in certain situations. The interested reader should consult NIST Report GCR 12-917-21 Soil-Structure Interaction for Building Structures for additional information.

relatively abundant, which has led to the identification of statistical trends and development of probabilistic and deterministic approaches for predicting design basis demand input. In contrast, highexplosive detonations are often the result of an intentional act, and thus are regarded as being more of a random event. The man-made nature of these blast loads, coupled with the lack of meaningful statistical data relating geographic location or recurrence period to specific threat scenarios, has largely relegated blast engineers

The primary difference between blast and seismic loading from a system response perspective is the area over which the load is distributed. Because seismic loads are a secondary effect of base excitation, they effectively engage the entire structure and require system response to resist the forces. In contrast, primary blast effects from an external detonation are typically localized, distressing isolated areas along the exterior of an affected structure, and often creating less overall demand on the LFRS than earthquakes. While design for both blast and seismic is performed with the intent of protecting people and assets, the focus of each design effort is quite different. The main goal for seismic-resistant design has historically been to mitigate overall structural damage and prevent global collapse. This is achieved by limiting inter-story drifts, allowing for controlled and distributed plastic deformations, and anchoring non-structural components. Global mass and stiffness distribution are generally key considerations for seismic design. In contrast, blast design focuses on protecting building occupants and critical assets from localized hazards. This is achieved by mitigating primary and secondary debris, preventing failure of various components of the building envelope, and providing continuity between structural

Figure 2: Illustration of strain-rate effects for concrete (adapted from Tedesco, 1999).

STRUCTURE magazine

March 2014

elements to prevent disproportionate collapse due to extreme damage to a localized area of a structure. Exterior building envelope hardening is generally a key consideration for blast-resistant design.

more) important to the structural integrity to the time at which the entire component is set of the entire LFRS as the vertical elements. in motion, and it is chiefly driven by the effects While response to seismic excitation is very of the impinging shock wave as it propagates much a battle of attrition, where attention through the component material and interis paid to hysteretic energy dissipation and acts with cross-sectional bounding surfaces. “riding out” the relatively long duration ground These early-time wave propagation effects motion, response to blast loading is quite dif- can lead to material damage such as spall and Differences at the ferent. Blast-loaded structural components breach, which can cause locally reduced section undergo a complex response evolution involv- capacity and hazardous secondary blast-borne Component Level ing early-time local material response followed fragments before the entire component is even There are many unique aspects of compoby “global” component response measured in set in motion. Conversely, global component nent-level response that pertain specifically milliseconds. Local material response refers to response refers to dynamic modes of response, to seismic or blast applications. For instance, TAY24253 BraceYrslfStrctrMag.qxd 9/3/09 10:09 AM Page 1 early-time material behavior that occurs prior such as flexure and direct shear, which engage strain rates in blast-loaded components can be orders of magnitude higher than those generated during a seismic event. It has been shown through experimental Y O U B U I L D I T. testing that common construction materiW E ’ L L P R O T E C T I T. als, such as concrete and steel, experience strain-rate-dependent dynamic strength increases beyond certain threshold limits. In practice, these apparent strength increases are typically captured through the use of dynamic increase factors (DIF) Stand firm. Don’t settle for less than the seismic protection applied to nominal yield and/or ultimate of Taylor Fluid Viscous Dampers. As a world leader in material strengths. Figure 2 illustrates the science of shock isolation, we are the team you DIFs for concrete associated with differwant between your structure and the undeniable forces ent types of loads. In general, strain-rate of nature. Others agree. Taylor Fluid Viscous Dampers effects tend to increase yield and ultimate are currently providing earthquake, wind, and motion strengths while reducing material-level protection on more than 240 buildings and bridges. ductility. Stiffness remains largely unafFrom the historic Los Angeles City Hall to Mexico’s fected by strain-rate effects. Also, as can Torre Mayor and the new Shin-Yokohama High-speed be seen from Figure 2, strain-rate effects Train Station in Japan, owners, architects, engineers, for earthquake events are most often negand contractors trust the proven ligible and are typically not considered technology of Taylor Devices’ in practice. Fluid Viscous Dampers. Structural components respond to seismic excitation in a cyclic manner. A well designed and detailed component will undergo numerous cycles of response without a major reduction in load carrying capacity. For structural components designed specifically for controlled plastic deformation, this sustained fidelity is of utmost importance from both an energy dissipation and a system-level structural integrity point of view. Loss of confinement, local material degradation, excessive rebar strain-penetration at connections, second-order effects due to excessive deflections, local buckling, Taylor Devices’ Fluid Viscous Dampers give you the seismic protection and fracture are all potential causes for you need and the architectural freedom you want. reduced component-level load carrying capacity and should be considered w w w. t a y l o r d e v i c e s . c o m during design. Many of these undesirable limit states can be avoided by ensuring adequate detailing. Because the entire LFRS plays a major role in North Tonawanda, NY 14120 - 0748 resisting seismic forces, horizontal elePhone: 716.694 .0800 • Fax: 716.695 .6015 ments such as diaphragms, collectors, and their connections are just as (if not

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the entire component and depend on characteristics such as boundary conditions, stiffness, mass, and blast pulse variation with time. Direct shear behavior is regarded as being independent of flexure and involves large, localized shear forces near component supports that result from high-frequency, multi-modal effects that take place prior to the onset of traditional flexural response. If a blast-loaded component survives the early-time wave propagation effects and is properly designed to resist direct shear forces, it will respond in flexure. Unlike structural response to seismic loads, where cyclic behavior is expected, response to blast loads typically involves a single, high-demand inbound incursion – rebound can also be important for scenarios involving stiff components, interior detonations, or blast pulses with a significant negative phase – followed by numerous cycles of relatively benign free vibration response. The peak deflection during initial inbound response can be very large depending on the desired performance objective, particularly if flexural hinging and perhaps even membrane response are permitted.

Synergies While design for seismic or blast loads alone does not guarantee adequate performance for both load cases, there are many synergies that exist between these design methodologies. In general, the shared benefits are related to design features implemented to ensure adequate behavior of members and systems loaded beyond their elastic limit. The three primary areas where synergies exist are capacity design, ductile detailing, and design for continuity – all of which are somewhat related. The capacity design methodology focuses on designing connections to allow for structural components to reach their full capacity and deform in a ductile manner up to failure. This precludes connection failure, as well as undesirable component failure modes such as shear and local buckling. In short, capacity design ensures connections are stronger than their connected structural members. Because various elements in the LFRS may be pushed beyond their elastic limit, seismic-resistant design typically requires critical elements (e.g., collector elements) to be connected for the calculated loads factored by an overstrength factor, Ωo, to ensure, indirectly, that the elements are connected to develop their full capacity. In addition, provisions in ASCE 7 Minimum Design Loads for Buildings and Other Structures allow for design of element connections to be based on the capacity of the connected element directly rather than using the calculated seismic

forces. Similarly, in blast-resistant design, efficiently designed members will exceed their elastic capacity during response. Therefore, their connections are commonly designed for the full member capacity and/or the peak calculated reaction. In this way, connections of critical members in the force resisting system for blast or seismic will oftentimes be designed for the full capacity of the member, allowing for quick and efficient connection design for both load cases. Ductile detailing, which is intimately related to capacity design, is achieved by designing members to exhibit “ductile” modes of response involving plastic deformations that occur prior to failure and away from connections. This is accomplished through adequate confinement, bracing/stiffening, and overall system connectivity. These are all recommended detailing practices for both seismic and blast design. By designing and detailing for ductility, members and systems can dissipate energy in a predictable and controlled manner without a premature loss in load carrying capacity. Ultimately, this results in reduced connection reactions when compared to a system designed to respond elastically. In addition, ductile detailing increases the robustness of a structural system and can often lead to a more economical design. Adequate connectivity of critical members is also required for most blast-resistant and seismic-resistant design applications. Design for progressive collapse – often termed disproportionate collapse – is frequently required under the umbrella of blast-resistant design. Although, strictly speaking, progressive collapse is regarded as a threat-independent phenomenon. Interested readers can refer to UFC 4-023-03 Design of Buildings to Resist Progressive Collapse for additional information. A fundamental concept of progressive collapse design is ensuring continuity of loadpath-critical framing members and floor slabs to allow the structure to bridge over removed or failed elements, thus maintaining structural integrity and preventing collapse of the structure. Similarly, in order to ensure load path continuity in a seismic-force-resisting system, seismic design typically involves the amplification of design forces for critical members that transfer loads. In addition, special material-specific seismic detailing provisions are typically required in design. For instance, continuity in concrete systems is provided in part by increasing development length, splice length, and hook length requirements for steel reinforcing bars. In steel structures, gravity members are designed for nominal axial loads while connections are designed for the gravity members’ full plastic capacity. In

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March 2014

light frame construction, various tie, bond, and anchorage requirements are also specified to achieve continuity. By focusing on the synergies of blast-resistant and seismic-resistant design, more efficient and less costly structures can be designed, detailed, and constructed than would otherwise be achievable by addressing each load case and associated design criteria independently. The Federal Emergency Management Agency (FEMA) has recommended various strategies for leveraging these synergies in design – refer to FEMA reports 439A and P-439B for additional information. In addition, organizations like the MultiDisciplinary Center for Earthquake Engineering Research at the State University of New York at Buffalo have begun to explore multi-hazard design concepts to be implemented for both seismic and blast applications.

Conclusion As the structural engineering community forges ahead toward enhanced resiliency and security of the nation’s buildings and infrastructure, loads due to earthquakes and blast will continue to be challenging and interdependent facets of building design. It is the hope of the authors that this article was successful in highlighting some of the key similarities and differences between seismicresistant and blast-resistant design, as well as emphasizing the potential benefits of leveraging their synergies in building design.▪ Eric L. Sammarco, P.E., M. ASCE, formerly a Civil Engineer with Black & Veatch’s Nuclear Energy Division, is currently a Ph.D. candidate in Civil Engineering at the University of Texas at Austin. Mr. Sammarco can be reached at esammarco@utexas.edu. Cliff A. Jones, P.E., M. ASCE, formerly a Civil Engineer with Weidlinger Associates, is currently a Ph.D. candidate in Civil Engineering at the University of Texas at Austin. Mr. Jones can be reached at cliffjones@utexas.edu. Eric B. Williamson, Ph.D., P.E., M. ASCE, is the J. Hugh and Betty Liedtke Centennial Professor in Civil Engineering at the University of Texas at Austin. Dr. Williamson can be reached at ewilliamson@mail.utexas.edu. Harold O. Sprague, P.E., F. ASCE, is a Principal Technical Consultant with Parsons. Mr. Sprague can be reached at harold.sprague@parsons.com.

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Structural Performance performance issues relative to extreme events

dvances in computational tools for earthquake engineering analysis continue to broaden the structural engineer’s ability to conduct performance-based simulations, evaluate ductility, and make decisions on performance criteria that involve deformations of a structure beyond the elastic limit. However, with the advent of these tools, the industry has also seen a proliferation of software packages with input options and parameters, which must be understood to perform acceptable and meaningful nonlinear analyses. In 2012, a 17-question survey was conducted to ascertain existing perceptions relating to the application of nonlinear analysis methods in earthquake engineering, how nonlinear analysis is being applied, which tools/guidelines are being used to support nonlinear analysis, and what are the overall barriers to entry for conducting nonlinear analysis. The survey was designed to correlate the responses to the various groups of end-users so that trends in user behavior could be identified. A secondary motivation for the survey, and a primary motivation for this article, is to stimulate discussion and solicit further feedback from the earthquake engineering community regarding the use of nonlinear analysis in practice. Interested readers are invited to share their experiences and viewpoints by responding to the survey link listed on page 19.

Nonlinear Analysis in Modern Earthquake Engineering Practice By Monique Head, Ph.D., Sheri Dennis, Susendar Muthukumar, Ph.D., P.E., Bryant Nielson, Ph.D., S.E. and Kevin Mackie, Ph.D., P.E.

Survey Demographics The 76 survey respondents classified themselves in one of four categories: 1) Academician/Research Engineer, 2) Industry Professional, 3) Consultant and 4) Other. The responses for each group were analyzed separately with a tally for each question, allowing trends and patterns to be exposed among the different categories. Understanding that there might be some regional bias to the responses, there was actually a strong representation of respondents from each of the four major U.S. time-zones: 33% stationed in EST, 16% in CST, 12% in MST, and 35% from PST. A majority of the respondents (approximately 86%) also reported having a Master of Science/Master of Engineering degree or higher, and 78% stated that they had obtained their Professional Engineer (PE) licensure. Respondents that listed themselves as Academicians/Research Engineers reported the highest number of individuals, 25/27 (~93%), who held a master’s degree or higher. Of the Industry Professionals, 20/23 (~87%) reported

16 March 2014

having degrees beyond the baccalaureate with at least master’s degree. Of those who identified themselves as Consultants, 14/21 (67%) have at least a master’s degree, 3 have PhDs and 2 have MBAs. Familiarity with Nonlinearity Analysis Types For the Academicians/Research Engineers, it wasn’t a surprise when the results concluded that most of the respondents with higher level degrees were familiar with and use nonlinear analysis methods. Academicians/Research Engineers reported capturing both geometric and material nonlinearities; however, all were familiar with or had applied material nonlinearity. All except two (~7%) of the Academicians/ Research Engineers responded as having applied, or having familiarity with, geometric nonlinearity. More than 50% of the Academicians/Research Engineers had applied or were familiar with contact nonlinearity. 81% of the Consultants reported having expertise in material nonlinearities, 57% stated having expertise with geometric nonlinearities, but only 19% stated having expertise with contact nonlinearities. Of the Industry Professionals, 90% reported having familiarity with geometric nonlinearities, 86% with material nonlinearities, and 38% with contact nonlinearities. Types of Structures Analyzed Respondents had to indicate what type of structures upon which they actually performed nonlinear analyses (multiple selections could be made). The majority of the respondents focused on buildings, followed by bridges and then non-building structures. In fact, 48% of the Academicians/Research Engineers analyzed both bridges and buildings but only 14% of Consultants and 13% of Industry Professionals focused on both. Areas of Expertise The next set of questions focused on the specific area of seismic expertise/experience for which the nonlinear analyses was being conducted. All of the respondents indicated that they specialize in seismic design and seismic retrofit. There was an equal distribution of Industry Professionals and Academicians/Research Engineers who specialized in performance-based design. Of all the respondents, the Consultants reported having the least amount of experience in seismic isolation and earthquake engineering research. However, all Consultants who participated in the survey responded as having expertise in seismic design, and more than half also responded as having expertise in seismic retrofit and performance based design. Three of these respondents also stated that they had expertise in seismic isolation and earthquake engineering research.

120

Risk Analysis

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Inclusion/Modeling of Soil‐Structure Interaction Nonlinear Time History Analysis

Number of Responses

Risk Analysis 80

Site Specific Seismic Hazard

Application of Damping 40

Selection/Scaling of Ground Motions

Inclusion/Modeling of Soil‐Structure Interaction

Nonlinear Time History Analysis

Site Specific Seismic Hazard

Application of Damping

Selection Scaling of Ground Motions

Pushover Analysis

Pushover Analysis 10

Structural Modeling of NL Elements

0 N/A

Clear/Easy Ambiguous

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Figure 1: "Academicians/Research Engineers" responses regarding guidelines to support various Figure 1: “Academicians/Research Engineers” responses regarding applications

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Inclusion/Modeling of Soil‐Structure Interaction

Nonlinear Time History Analysis

Site Specific Seismic Hazard

Inclusion/Modeling of Soil‐Structure Interaction Nonlinear Time History Analysis

Selection Scaling of Ground Motions

Site Specific Seismic Hazard

10 8

Application of Damping

Selection Scaling of Ground Motions

Pushover Analysis

2 Structural Modeling of NL Elements

Ambiguous

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Figure 2: “Industry Professionals” responses regarding guidelines Figure 2: "Industry Professionals" responses regarding guidelines to support various applications

March 2014

that consider the guidance for performing these analyses to be ambiguous! Do you agree?! These results indicate that more direction and guidance for these topics would be useful. continued on next page

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Pushover analysis Selection/scaling of ground motions Application of damping to a structure Site specific seismic hazard analysis Nonlinear time history analysis Inclusion/modeling of soil-structure interaction • Risk analysis (e.g., fragility curves, damage assessment, etc.) The majority (more than 75%) of all respondents said that the guidance for most of these topics was ambiguous, where the patterns plotted by respondents (Academicians/Research Engineers, Industry Professionals, Consultants, and Other), are also noted in Figures 1 through 4, respectively. While the distribution is important to note, it is overwhelmingly apparent that there are more respondents

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Structural Modeling of NL Elements

in this category).

Participants were asked to identify “Which resources do you currently use or have used in the past?” In all categories of respondents, more than 75% referred to ASCE and FEMA/NEHRP documents as resources used or used in the past to conduct nonlinear analysis. Other resources to choose from were ATC, Tech Briefs, AASHTO/State DOT, and Other. As a follow-up question to this, respondents were prompted to rate whether or not available guidelines for the following topics were “clear and easy to apply” or “ambiguous and needs improvement”: • Structural modeling of nonlinear elements (e.g., fiber, elements, material modeling, etc.)

Ambiguous

Figure 4:”Other” Professionals’ responses regarding guidelines to Figure 4:"Other" Professionals' responses regarding guidelines to support various applications (Note: there were only 2 respondents in this category). support various applications (Note: there were only 2 respondents

to support various applications.

Use and Clarity of Nonlinear Analysis Resources

Clear/Easy

ETY OF NAV A CI O

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• THE ERS S NE

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Risk Analysis

Application of Damping

Structural Modeling of NL Elements

various applications.

Ambiguous

Figure 3: “Consultant” responses regarding guidelines to support Figure 3: "Consultants" responses regarding guidelines to support various applications

Number of Responses

guidelines to support various applications.

Clear/Easy

Which software packages/analysis programs do you rely upon for nonlinear analysis? Select all that may apply.

Figure 5: Responses to software packages/analysis programs used for nonlinear analysis.

Software Used for Nonlinear Analysis

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Trends in the specific software packages/ programs used to carry out any nonlinear analysis were solicited next. Given the wide range of nonlinear analysis packages available, respondents were asked to select from various academic and commercial software packages. The options included ANSYS, PERFORM-3D, SAP/ETABS, STAADPro, OpenSEES, other programs like State DOT Bridge Software, Oasys/LS DYNA, GTStrudl, NASTRAN, and in-house software. Zero participants selected ADAPT and NASTRAN. NASTRAN, written in FORTRAN, can be used for analyzing linear elastic static and dynamic case studies, and is targeted more for the aerospace and mechanical fields). The breakdown per respondent category was as follows, and is also shown in Figure 5:

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In your opinion, are any of the following barriers of entry for applying/using nonlinear analysis? Select all that may apply.

Figure 6: Breakdown of responses to barriers for implementing nonlinear analysis.

• Academicians/Research Engineers used OpenSEES more frequently followed by SAP/ETABS and PERFORM-3D. OpenSees (Open System for Earthquake Engineering Simulation) is a software framework used to simulate structural and geotechnical systems for earthquake analysis. The software can be used for parallel computing of scalable simulations. • Industry Professionals responded that they primarily used SAP/ ETABS, followed by PERFORM-3D and STAADPro which are 3-D structural analysis programs commonly used by structural engineers, especially for nonlinear analysis and performance assessment. • Consultants did not provide as many responses compared to the other respondents but it was concluded that SAP/ETABS was the most popular. • In the Other category, both participants responded as having experience using SAP/ETABS with one also using PERFORM-3D.

Barriers for Implementing Nonlinear Analysis Identification of any real or perceived barriers to applying/using nonlinear analysis was queried. The respondents were asked to provide their personal opinion, where multiple answers could have been selected. Of all the respondents, “Not Practical/Time-Consuming (61%),” “Lack of Adequate Guidelines

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March 2014

Please identify the #1 topic for which having additional guidance would be helpful. Select only one.

Figure 7: Responses to a “desire for additional guidance” question.

(43%),” “Too Complicated (29%),” “Lack of Research (22%),” and “Inadequate Software Capabilities (21%)” were all reasons for barriers to entry. A breakdown of the responses is shown in Figure 6.

Desired Additional Guidance Key to furthering the use of nonlinear analysis tools is to identify the #1 topic (one selection only allowed) for which having additional guidance would be most helpful. Academicians/Research Engineers responded that the #1 need was for there to be benchmark problems with solutions followed by modeling nonlinear elements/understanding parameters. Consultants and Industry Professionals both responded that the #1 need was for guidance on modeling nonlinear elements/understanding parameters. Overall, as shown in Figure 7, modeling nonlinear elements/understanding parameters was the #1 topic followed by the need for benchmark problems with solutions. The other options for selection were the need for software-specific input files and guidance, selection of ground motion input or load parameters, and NLA procedures (NLTH, etc.).

Additional Comments The survey also had a “free field” comment box that also brought to light some important concerns, where respondents were asked to report any barriers to entry into nonlinear analysis techniques and tools. Simply stated, many reported that a barrier to entry is the high complexity

and time required for inputting the information and interpreting the data, which is not always straightforward. Moreover, many noted that there is a need to be able to communicate the importance of doing a nonlinear analysis to the owners, as the apparent gain to pay for a more extensive analysis is not always clear. It is known that advanced nonlinear analysis can be used to assess the need for a retrofit process, structural upgrade, or simply a need to verify adequate capacity given certain demand loads on a structure. But when is it appropriate to even conduct or justify the cost of conducting a nonlinear versus linear analysis? As such, barriers to entry are not just limited to the need for more guidance and examples, but for communicating the benefit of such an analysis (where appropriate) to the owners. All of these highlights were important to note and can be game-changers when entering (or not!) the field of nonlinear analysis.

Conclusions and Recommendations The use and applicability of nonlinear analysis tools and design guidelines to solve various problems facing the earthquake engineering community are not always straightforward and by no means fully transparent. Despite the advancing capabilities of powerful software packages, there is quite a level of detail and background knowledge needed in order to develop refined models and perform various analyses, where the amount of input information required can be daunting given the multitude of input options that may not always be known at the time of analysis. While there is a gap between the end-users, design guidelines, and software packages, it is known that many benefits can be gained when a nonlinear analysis is conducted for an appropriate case. However, when to apply nonlinear

Take the opportunity to chime in on the discussion. Please visit the link and complete the survey, www.surveymonkey.com/s/nonlinearanalysis. Be sure to leave your opinion as to the next big effort that should be made to improve this aspect of Earthquake Engineering.

analysis is not always straightforward or transparent, especially when various nonlinearities need to be captured. With the advent of faster processors, implementing nonlinear techniques in design practice will be more feasible in the coming decade provided adequate guidelines exist. While not alarming, the results from this survey helped to: • reveal to users within the earthquake engineering community who (Academicians/Research Engineers, Industry Professionals, Consultants, and Other) is using what type of software, for what application of nonlinear analysis, and the various tools/guidelines being used to support the analysis; • identify some barriers for entry into nonlinear analysis such as the high complexity, time consumption, lack of clear guidance, and communicating the benefit of advanced analyses to the owners; and • provide users within the earthquake engineering community, especially the ASCE Subcommittee on Emerging Analysis Methods in Earthquake Engineering, with evidence to possibly develop more

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guidance materials that offer benchmark problems with solutions followed by modeling nonlinear elements/understanding parameters when conducting nonlinear analysis. To assist with bridging the gap of seemingly inadequate guidance for conducting nonlinear analysis, users are recommended to reference available resources, such as tech briefs published by NEHRP, which appear to currently be underutilized and could assist in reducing the overall ambiguity that is perceived around this topic.

Nonlinear analysis can be used when owners request ways to reduce costs (for new construction) by optimizing material use, more likely though as a way of demonstrating a building retrofit is perhaps not even necessary (or if it is, that only minor changes/systems are needed rather than what the code would require), or even as a way of quantifying performance for owners, insurance and risk managers that may look at inventories of structures. There is, however, a cost associated with performing this analysis. So the major

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Acknowledgment The findings and opinions presented herein are those of the authors and are not necessarily those of the American Society of Civil Engineers (ASCE) Subcommittee on Emerging Analysis Methods in Earthquake Engineering. The support provided by the Subcommittee is greatly appreciated. The Subcommittee cordially invites all those interested in learning more about its activity and future pursuits to attend Subcommittee meetings during the 2014 ASCE Structures Congress in Boston, MA.▪ Monique Head, Ph.D., is an Assistant Professor in the Department of Civil Engineering at Morgan State University, Baltimore, MD. She is the vice chair of the ASCE/SEI Subcommittee on Emerging Analysis Methods in Earthquake Engineering. She may be reached at monique.head@morgan.edu. Sheri Dennis is a Master of Science degree candidate in the Department of Transportation and Urban Infrastructure Studies at Morgan State University. She may be reached at sheribdennis@aol.com.

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Structural

SuStainability sustainability and preservation as they pertain to structural engineering

he Timber Tower Research Project by Skidmore, Owings & Merrill, LLP (SOM) was publically released in June of 2013, and is available for download at SOM’s website. The goal of the research project was to develop a structural system for tall buildings that uses mass timber as the main structural material and minimizes the embodied carbon footprint of the building. The structural system research was applied to a prototypical building based on an existing concrete benchmark for comparison. The concrete benchmark building is the Dewitt-Chestnut Apartments, a 395-foot tall, 42-story building in Chicago designed by SOM and built in 1966. SOM’s proposed system is the “Concrete Jointed Timber Frame”. This system relies primarily on mass timber for the main structural elements, with supplementary reinforced concrete at the highly stressed locations of the structure: the connecting joints. This system plays to the strengths of both materials and allows the structural engineer to apply sound tall building engineering fundamentals. The result is believed to be an efficient structure that could compete with reinforced concrete and structural steel systems, while reducing the embodied carbon footprint of the structure by 60 to 75%.

Timber Tower Research Project By Benton Johnson, P.E., S.E., David Horos, P.E., S.E., LEED AP and William Baker, P.E., S.E., F. ASCE, FIStructE

Project Basis Benton Johnson, P.E., S.E., is an Associate at Skidmore Owings & Merrill LLP, Chicago, IL. He is the Project Engineer on the Timber Tower Research Project and can be reached at benton.johnson@som.com. David Horos, P.E., S.E., LEED AP, is a Director at Skidmore Owings & Merrill LLP, Chicago, IL. Timber Tower Research Project and can be reached at david.horos@som.com. William F. Baker, P.E., S.E., F. ASCE, FIStructE is the Structural Engineering Partner for Skidmore, Owings & Merrill LLP. Bill has dedicated himself to structural innovation, most notably developing the “buttressed core” structural system for the Burj Khalifa, the world’s tallest manmade structure. He is a Fellow of both the ASCE and the IStructE and a member of the National Academy of Engineering.

The basis of the research project was rooted in sustainable urban development. Recent population projections have estimated the current world population of 7.0 billion people to increase to 11.0 billion people by the year 2050. More importantly, the number of people that will be living in cities has been estimated to double from 3.5 billion people to 7.0 billion people in the same time frame. Tall buildings will likely be needed in order to house that many additional people in growing cities. Tall buildings constructed to meet population demands need to be developed in sustainable ways to limit environmental impacts. Tall buildings built using current technology and materials pose a challenge to sustainable city development because they offer both positive and negative environmental impacts. Positive impacts include reducing urban sprawl, promoting alternative transportation, and efficient energy use. These benefits come at the cost of emitting more carbon dioxide to produce the materials and to construct the building. These carbon emissions are referred to as the embodied carbon footprint of a building. A tall building’s embodied carbon footprint is significantly higher relative to lowrise buildings on a per square foot basis. This is

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because the structure is usually responsible for the majority of the building’s embodied carbon footprint, and tall buildings require far more structure to support their height. The structural system chosen for a tall building can have a significant impact on the overall embodied carbon footprint of the building.

Design and Sustainability Issues Structural engineers currently have four primary materials in which to design buildings: steel, concrete, masonry, and wood. Tall buildings currently use steel or concrete almost exclusively, for two reasons. First, with some limited exceptions, non-combustible materials are required by most building codes for buildings greater than four stories tall. Second, steel and concrete have higher material strengths than masonry and wood, making them a natural choice for tall buildings which require support of very large loads. These factors have generally limited wood use to low-rise buildings. Recently, developments in mass timber technology are overcoming these challenges. Mass timber products such as cross-laminated timber (CLT) can be built up using small pieces of dimensional lumber and structural adhesives to achieve panels as large as 1foot thick and 40 feet long. These panels can be used as floors and shear walls with structural sizes necessary to support a tall wooden building. Wood members of this size have an equally important characteristic; they behave like heavy timbers in a fire and form an insulating char layer which protects underlying material. The charring behavior is predictable and preserves a portion of the member’s structural strength, making performance based fire design of mass timber structures possible. Mass timber has made wood a viable choice for multi-story buildings as evidenced by completed projects in Europe and Australia, and many other proposed projects around the globe. The structural and fire engineering advancements of mass timber have made recent multi-story wood buildings possible. However, the sustainability of wood seems to be an equally important consideration in the resurgence of multi-story timber buildings. Wood has been shown to be more sustainable than other materials because it generally requires less energy to produce compared to structural steel and reinforced concrete. More importantly, wood is approximately 50% carbon by weight, a carbon sink that is the natural result of photosynthesis. These sustainable aspects of wood make mass timber an attractive material from which to construct the sustainable cities of the future. The intersection of increasing urban populations, need for tall buildings, and the sustainability of wood has led to the increasingly popular concept of tall wood buildings. SOM has committed decades of tall building

Architectural detail of the wood structure proposed in the Timber Tower Research Report.

design expertise to furthering this concept, through the Timber Tower Research Project, by identifying key design and construction issues related to tall wood buildings and proposing the “Concrete Jointed Timber Frame” structural system. This system is optimized for tall buildings and could be competitive with existing tall building structural systems. The proposed system balances the requirements of building marketability, economy, and sustainability.

Material Optimization The primary goal of any structural system is to provide a marketable and valuable building to the owner and occupants. A marketable building must have adequate and flexible floor area to layout useful space for the occupants. The most marketable building layout is an open floor plan which allows a variety of room layouts and maximum flexibility for future changes. An open floor layout requires that the floor structure span the entire distance of the leasable area. This distance in the Benchmark Dewitt-Chestnut building was 28 feet 6 inches, with a clear span of 26 feet 3 inches. The most advantageous system to span this distance is a flat mass timber panel which minimizes floor-to-floor height of the building. The required panel thickness to span

the required distance was determined to be 13½ inches. This thickness was thought to be too great compared to the material required for the Reinforced Concrete Benchmark to be economically viable. Therefore, alternative methods to span the required distance were investigated in order to reduce the amount of structural materials used. The controlling design consideration for the mass timber floors was determined to be vibration due to occupant activity. The floors were analyzed according to American Institute of Steel Construction Design Guide 11, utilizing the velocity-based methodology, which was found to be more useful for flat slab-type floors. Evaluation of the criteria shows that increasing floor stiffness is the most effective way to control vibrations. The floor stiffening effect of end rotation restraint (fixed end condition) was quickly realized as an efficient way to reduce vibrations. It was determined that an 8-inch-thick mass timber floor panel could be used if end restraint was provided. This requires moment connections at the intersection of mass timber floor panels with vertical elements such as mass timber shear walls and structural glued laminated timber perimeter columns. Several connection schemes were investigated to provide the required moment connections. Steel reinforcing epoxy connected to the mass timber and cast-in reinforced concrete joints were determined to be the most reasonable solutions due to the ability of reinforced concrete to resist complex load paths. These reinforced concrete joints are able to resist floor-to-floor compression, shear, bending moments, and torsion, thus creating an efficient compositetimber system. The reinforced concrete joints also proved to be useful in other tall building aspects. The concrete jointing between timber floors and timber shear walls provides a link beam between individual wall panels. This creates a stiff lateral load resisting system which is required for a tall building. It was also determined that the demands on the link beams were beyond the capacity of a structural glued laminated wooden link beam, requiring the use of a material other than wood. The concrete joints and link beams were also useful in the design of the lateral system to resist net uplift due to lateral loads. The Prototypical Building has approximately 40% of the dead load of the Benchmark Building. This led to net uplift forces at the extremities of the lateral load resisting system. This net uplift would have been exacerbated without the concrete joints which account for over 50% of the entire structure dead load, yet only 20% of the structural material volume for a typical floor.

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SOM’s Timber Tower Research Project proposes a “Concrete Jointed Timber Frame” system that relies primarily on mass timber for the main structural elements, with supplementary reinforced concrete at the connecting points.

A comparison of the structural materials required to construct the Benchmark and Prototypical building shows that the proposed system is very efficient in material consumption and could be competitive with reinforced concrete. The goal of minimizing the structural materials used, namely mass timber, will help reduce costs and minimize new demands on forest resources which may become strained due to increasing populations and demands. The non-structural effects of the proposed system were evaluated and the most notable effect was the acoustic treatment required on top of the mass timber floors in order to achieve a marketable acoustic rating. The most effective treatment was determined to be a 2-inch-thick gypsum concrete topping. This treatment thickness, in addition to potential ceiling finishes, required 3 inches of additional floor-to-floor height in order to maintain the same floor-to-ceiling height as the Benchmark building. This has impacts on wind loads on the building, and non-structural costs such as the exterior wall system.

Conclusion SOM believes that the proposed system is technically feasible from the standpoint of structural engineering, architecture, interior layouts, and building services. Additional research and physical testing is necessary to verify the actual performance of the structural system relative to the theoretical behavior. SOM has also developed the system with consideration for constructability, cost, and fire protection. Reviews from experts in these fields, and physical testing related to fire, is also required before this system can be fully implemented in the market. Lastly, the design community must continue to work creatively with forward thinking municipalities and code officials using the latest in fire engineering and performance based design to make timber buildings a viable alternative for more sustainable tall buildings.▪

Building Blocks updates and information on structural materials

(b)

(a) Figure 1. CFS-NEES archetype building utilized to organize research and for full-scale testing: (a) rendering from BIM model, only shear walls sheathed (b) detail at shear wall chord stud.

eismic design of buildings using repetitively framed CFS (CFS) members has largely been enabled through a series of dedicated tests conducted on shear walls and compiled for convenient use in the AISIS213 standard, supported through the seismic response modification coefficients and procedures in ASCE 7. This approach has served engineers and industry well, but has not provided a clear path towards the development of new and novel seismic force resisting systems utilizing CFS, nor does it provide the necessary knowledge for modeling CFS buildings as systems. At its core, the seismic performance-based design (PBD) paradigm presumes an ability to efficiently model the key nonlinearities and redistributions inherent in a building under seismic demands. For CFS structures, important knowledge that is required to create such models for PBD is missing. To varying degrees, fundamental gaps exist with respect to the hysteretic performance of CFS connections, members, assemblages, and full buildings. Characterization and implementation into models are also lacking. The CFS-NEES effort has as its aim the development of experimental benchmarks, fundamental characterization, and the demonstration of efficient means to model CFS structures – even with their inherent complexity.

The CFS-NEES Effort Advancing CFS Earthquake Engineering By Benjamin Schafer, Ph.D., P.E.

Benjamin Schafer, Ph.D., P.E., is Professor and Chair of the Department of Civil Engineering at Johns Hopkins University. He is Director of the Cold-Formed Steel Research Consortium, Chair of the Structural Stability Research Council, and an active member of the AISI Committee on Specifications and Committee on Framing Standards. Benjamin can be reached at Schafer@jhu.edu.

Building Archetype Central to the CFS-NEES effort was the professional design of a two-story CFS commercial building sited in Orange County, CA (site class D) that is 49.75 feet x 23 feet in plan and 19.25 feet tall, with a total seismic weight of 78 kips. The design was completed by Devco Engineering, with input from the project team and the Industrial Advisory Board. A design narrative, complete calculations, and full drawings are available. A key feature was the selection of ledger framing, a choice that was strongly advocated by the Industrial Advisory Board based on current

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practice. In ledger framing, the building is constructed one floor at a time, but the floor joists are hung from the top of the studs. The joists and studs are not aligned, so a ledger, or carrier track, is attached to the interior face of the studs running along its length to provide a connection point for the joists (Figure 1). A key detail in such a system is the joining of the shear wall chord studs across stories: a flat plate attached to the stud web penetrates through the floor (Figure 1b).

Member Characterization Fundamental to the behavior of thin-walled CFS members are the stiffness reductions that may occur due to local, distortional, and global buckling under load. These reductions must be captured within designs and models if the full system created by CFS members is to be assessed. Existing test data facilitated the development of a new method for determining the stiffness reduction and backbone moment-rotation and/ or moment-curvature response under local and distortional buckling. The method is general and, in the spirit of the Direct Strength Method of CFS design, uses the cross-section slenderness to predict the reduced stiffness and full backbone response. Given a lack of available data on member cyclic response, the American Iron and Steel Institute (AISI), in collaboration with the CFSNEES effort, funded a companion project to provide explicit data on cyclic response of CFS members. Working at Virginia Tech, researchers carefully selected members and boundary conditions to study local, distortional, and global cyclic response of CFS members in axial and bending (Figure 2). These results form the basis for development of seismic force resisting systems that incorporate complete CFS member response, as opposed to current systems that largely seek to use alternative mechanisms, independent from the members – bearing in wood or steel connections, yielding of straps, etc. – to resist seismic demands.

OSB Sheathed Shear Wall Characterization The CFS-NEES archetype building employs CFS-framed, OSB-sheathed shear walls. This is a common type, available in AISI-S213 for prediction of strength and stiffness. However, actual construction differs from the tests used to develop the AISI-S213 tables: shear wall sizes are not equal to 4-foot x 8-foot OSB panel, so numerous additional horizontal and vertical seams exist in actual shear walls; a large 0.097-inch thick x 12-inch deep carrier or ledger track blocks out the last 12 inches at the top of a shear wall; gypsum board is sheathed on the interior face of the wall; and, in some cases, the field studs are not the same thickness as the chord studs that frame out the shear wall. In addition, complete hysteretic response of the shear walls is not available, requiring the initiation of a test program and characterization effort. Thanks to collaboration with the University of North Texas, the CFS-NEES project tested sixteen OSB-sheathed shear walls following the CUREE protocol (Figure 3); complete results are available in the test report and related papers. Strength degradation initiated at levels between 2% and 4% drift. Developed strength was in excess

Figure 2: Cyclic axial load-deformation response in local buckling for 600S162-33.

of AISI-S213 predictions, except in the case where shear wall field studs are thinner than the chord studs, a common practice for lightly loaded upper stories that should be accounted for in design. The addition of panel seams, ledger, and interior gypsum caused some divergence in stiffness predictions from AISI-S213 and may lead to greater than expected overstrength. Characterization of the test results was completed by determining parameters for one-dimensional (V-∆) phenomenological models employing the equivalent energy elastic-plastic (EEEP) model and the Pinching04 model. EEEP models are not appropriate for time-history analysis of these systems, only for

Figure 3. Hysteretic response of 1.22 m x 2.74 m OSB sheathed shear walls: (a) with ledger (b) gypsum board (c) baseline (d) extra vertical seam (e) front of Test 2. 10

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15.2cm 30.5cm

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Figure 4. Fastener testing assembly: (a) front (b) side detail (c) typical response. 8 7

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pushover analysis. The Pinching04 models are utilized in the CFS-NEES building models as discussed under Full Scale Building Modeling.

“Fastener” Characterization For CFS-framed, OSB-sheathed shear walls, the key energy dissipating mechanism occurs at the stud-fastener-sheathing connection. As the studs rack laterally, the fasteners tilt (and bend) as they bear into and damage the sheathing. Stiffness of the shear walls also relies on this same mechanism. In shear walls framed and sheathed with wood, it has been found that a similar mechanism dominates the response, and reasonable estimates of shear wall parameters can be derived directly from this local “fastener” response. Characterizing this “fastener” response required a series of cyclic tests on stud-fastener-sheathing assemblies (Figure 4). The tests varied stud thickness, fastener spacing, and sheathing type. The direct shear response of the fastener assemblies is similar to the full walls, but even more pinched. Each test was characterized using the Pinching04 model, and complete results are provided in a CFSNEES research report and a related paper. Work connecting the fastener response to the overall shear wall response is underway, and initial results indicate that with a little care – particularly with respect to hold-down flexibility – small-scale fastener tests have excellent predictive power for full-scale shear wall tests. A lack of knowledge on the stiffness and cyclic response of typical connections in CFS goes beyond the details common in shear walls. As a result, as a companion to the CFS-NEES effort, a project was undertaken at Virginia Tech to understand the cyclic response of CFS connections more fully. The results provide a key building block for models of CFS assemblages and full (b) buildings. continued on next page

(a)

Full-Scale Building Modeling

(b)

Figure 5. Three-dimensional OpenSees models of the CFS-NEES archetype building T=0.64 then 0.38 sec, where T=0.32 in the test: (a) shear walls only (b) shear walls and all gravity framing.

Figure 6. Shift in long- and short-direction first-mode period through construction phases: (a) LFRS and gravity steel only (b) exterior sheathed (c) inside face of exterior sheathed with gypsum (d) interior nonstructural walls & stairs (e) exterior DensGlass sheathed.

The CFS-NEES full-scale building modeling effort has two major goals: (1) to provide a model that can meaningfully predict the CFS-NEES building response in order to understand the behavior of the building better and examine its response against a full suite of seismic excitations; and (2) to evaluate what level of model fidelity is necessary for engineers and researchers modeling buildings framed from CFS. Modeling the response of CFS buildings, even a particular CFS building, introduces an enormous number of potential assumptions. The project explored a complete model tree spanning from two-dimensional models with strength and stiffness based on specifications available to engineers, to threedimensional models with shear walls based on direct experimental characterization and all steel framing explicitly modeled. Research is still underway, but preliminary results indicate that a high degree of model complexity is required for developing observed system response. Consider a threedimensional model with only shear walls included as rigid diaphragms (Figure 5a). The first-mode period, even using the experimentally calibrated shear wall stiffness, is 0.64 sec. In white noise testing, the same building

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Figure 7. CFS-NEES Full-scale building testing and measured drift during seismic excitation: (a) Phase 1 completed building (b) Phase 1, story drift, Canoga Park (~DBE) (c) Phase 2e completed building (d) Phase 2e, story drift, Rinaldi (~MCE).

with only shear walls in place has a first-mode period of 0.32 sec. An alternative model, with all wall framing explicitly included (Figure 5b), resulted in a much more accurate firstmode period of 0.38 sec. A key feature of this model is the inclusion of the full length ledger or carrier track, as well as the larger header members above openings. Work continues on several fronts with respect to the modeling: direct comparison with the full scale building testing, improving the complex model with a semi-rigid diaphragm, investigating how best to use models in engineering practice, and developing more robust reduced-order models for nonlinear time history analysis.

Full Scale Building Testing In the summer of 2013, the project conducted full-scale tests on the CFS-NEES building at the NEES facility at the University of Buffalo. Two buildings were constructed; the first (Phase 1) had the complete lateral force resisting system sheathed, but otherwise all other gravity framing as bare steel (Figure 7a). After full-scale testing using the Canoga Park motion from the 1994 Northridge

earthquake, this structure was dis-assembled and a second specimen (Phase 2) constructed. Testing examined the change in building response as a function of construction elements – gypsum, interior non-structural, etc. – as summarized in terms of shift in first-mode period (Figure 6 ). For Phase 2e (Figure 7c), the building was subjected to the Rinaldi ground motion, which is consistent with MCE-level spectral accelerations. Under seismic testing, both the Phase 1 and Phase 2e buildings experienced minimal drift and returned to straight after excitation (Figure 7b,d ). For the Phase 2e building, the story drift under Rinaldi was less than 1% and damage only occurred in the interior nonstructural walls, largely confined to corners near openings. This full-scale testing provides the first look at the full system effect for buildings framed from CFS and is significant across the board: the building is stiffer and stronger than engineering designs suggest; the building responds as a system, not as a set of uncoupled shear walls; and the gravity system contributes to the lateral response. Overall performance for the tested building was far better than code minimums, and

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far better than advanced engineering models (e.g., Figure 5a), but not necessarily for wellunderstood reasons. Significant work remains to decipher the collected data fully.

Conclusions CFS-NEES is providing a multi-prong effort to advance our understanding of seismic behavior and perform improved designs for structures framed from CFS. Significant progress has been made in hysteretic benchmarking and characterization at a variety of levels, from fastener, to member, to assemblages such as shear walls, as well as whole buildings. In addition, progress has also been made in predictive models, again across scales, that has potential for improved design. Full-scale testing of the CFS-NEES building provides a first look at the full system effect for buildings framed from CFS, which is significant across the board, requiring new approaches in prediction and design. Work remains to address details not fully explored (e.g., semi-rigid diaphragm behavior) and fully enable engineers working in this domain. For detailed reports and papers, visit www.ce.jhu.edu/cfsnees.▪

Structural ForenSicS investigating structures and their components

afety of exterior elevated decks, balconies and porches continues to be an important national issue. Section 1604.8.3 of the 2009 International Building Code (IBC) states that decks shall be anchored to the primary structure and designed for both vertical and lateral loads. The focus of this article is to address knowledge gaps on the performance of decks subjected to lateral loading. The information presented here builds on earlier works dealing with lateral loading on decks from wind, seismic and occupants (Lyman and Bender 2013; Lyman et al. 2013; Parsons et al. 2013). Armed with a better understanding of lateral loads, this article targets load transfer from decks to the main structure floor framing and diaphragm. Two 12-foot x 12-foot decks were connected to a portion of a light-frame wood diaphragm to simulate realistic support conditions. Lateral loads were applied to these decks with and without tension hold-down connectors, in accordance with a prescriptive detail that was introduced in the 2009 International Residential Code (IRC) (Figure 502.2.2.3). Load paths were characterized by instrumenting each lag screw and tension hold-down connector. The study yielded counterintuitive results that will help guide new design solutions and products to resist lateral loads.

Lateral Load Path and Capacity of Exterior Decks By Brian J. Parsons, Donald A. Bender, P.E., J. Daniel Dolan, P.E., Robert J. Tichy and Frank E. Woeste, P.E.

Materials and Deck/Diaphragm Construction

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

Two 12-foot x 12-foot decks were constructed using similar materials. One deck was constructed with a tension hold-down at two corners and the other had no hold-downs. The decks were built in accordance with Design for Code Acceptance 6 (DCA 6), which follows the 2009 IRC. The deck ledger was a 12-foot-long 2x10 with 2x10 joists spaced at 16 inches on center. The deck boards were wood-plastic composite (nominal 1x6) Trex Accents. All lumber used for the deck joists and ledger was incised and pressure preservative treated (PPT), No. 2 and Better Hem-Fir. Details on the deck construction and testing were reported in Parsons (2012). The simulated house diaphragm assembly was constructed in accordance with the 2009 IRC and the Wood Frame Construction Manual (AF&PA 2001). The sheathing (23/32-inch nominal thickness) was glued and nailed to the joists using construction adhesive designed for subfloor and deck applications. The nails were 2.5 inches by 0.131 inches and fastened the sheathing to the joists per IRC Table R602.3. When hold-downs were used, nails were spaced 6 inches on the

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Figure 1: Elevation and plan views of test set-up construction.

diaphragm joist with the hold-down attached. Figure 1 shows the elevation and plan views of the test set-up. The deck ledger was attached with lag screws that had 0.5-inch diameter full body, a length of 7 inches (to accommodate the load cell) and a root diameter of approximately 0.370 inches. Lag screws were installed 15 inches on center in a staggered pattern as specified in IRC Table R502.2.2.1. Deck joist hangers proved to be a weak link in the lateral load path, and deserve special attention (See the sidebar at the end of this article.) The hold-down connectors used on the second deck configuration were Simpson Strong-Tie (SST) DTT2Z with a “ZMAX” protective coating. The hold-down was 14-gauge steel and a 0.5-inch diameter threaded rod was used to connect the hold-downs from the deck to the house. The 0.25-inch x 1.5-inch screws used with the hold-down were SST Strong-Drive screws (Model No. SDS25112) that had a double-barrier coating with equivalent corrosion resistance to hot-dip galvanized.

Test Methods Lateral loading was applied as a resultant line load acting through the centroid of the deck surface. Since large displacements were anticipated, force was applied with a come-along as shown in Figure 2. A 10 kip load cell was installed

Figure 2: Load application set-up.

in-line with the come-along to record the force applied to the deck. A conservative assumption was made that the deck substructure would provide minimal lateral resistance; therefore, the deck was supported on rollers along the outer beam. The simulated house diaphragm was securely anchored to the laboratory reaction floor. Load cells were used to record forces in all lag screws connecting the deck ledger to the diaphragm rimboard and hold-downs. Seven string potentiometers were used to measure deck displacements.

Observed Damage

and fracture of deck board screws was also observed in this same region. In both tests, no damage was observed in the deck ledger-to-house rimboard connection or in the simulated house diaphragm. In the test that used hold-down tension connectors, deck joists fractured in weak axis bending. This was due to the hold-down installed on the compression chord producing larger rotational joist stiffness at the ledger connection than the joist hangers provided on the other joists. This caused load from the other deck joists to be attracted to the end joist, resulting in fracture. Once the end joist fractured, the remaining joists fractured due to progressive failure.

Moment couples formed by two screws in each deck board-to-deck joist connection provided the lateral stiffness of the deck diaphragm. In both tests, splitting of the top edges of the deck joists was the main source of damage, caused by the moment couple from the deck screws that induced stresses perpendicular to the grain. Each deck joist was completely split to the depth of screw penetration from the load drag strut to the ledger board. Significant yielding

For the test with no hold-down, the load displacement curve at the load drag strut, shown in Figure 3, can be divided into three segments. The first segment was a softening curve that is seen in tests of many mechanically connected structural assemblies as slip occurs and damage initiates. At a displacement of approximately 3.5 inches, significant joist splitting had occurred and most of the diaphragm stiffness from the deck board

Results and Discussion

Load-Displacement Curves

Figure 3: Load-displacement curves for deck with and without hold-downs.

STRUCTURE magazine

attachment was lost. The second segment of the load-displacement curve from 3.5 to 17 inches is approximately linear, with stiffness nearly equal to that of the bare frame (also shown in Figure 3). After 17 inches, the third segment shows an unexpected large increase in stiffness. For the test with hold-downs, slightly higher stiffness and load at 4-inch displacement were observed due to the hold-downs resisting rotation of the deck joists. Similar to the first test, the second segment from 4 to 15 inches reflects the frame stiffness, with deck boards contributing little stiffness. At a displacement of approximately 16 inches, the outer deck joists ruptured in weak-axis bending, followed by a sharp drop-off in load. In the third segment, a large increase in stiffness was once again seen at approximately a displacement of 17 inches, even after deck joists had severely fractured. Determining the exact reason for the large increase in stiffness does not have practical significance because it occurred at extreme levels of displacement that would likely cause column instability under gravity loads. At this point of increased stiffness, significant damage was also present in the joists which would compromise the safety of

Figure 4: Lag screw forces on deck without hold-downs. (Overturning tension force calculated assuming end joist resists full overturning moment.)

March 2014

Figure 5: Recorded hold-down force versus SDPWS calculations.

the deck. From a practical standpoint, deck failure could be defined as the point when the diaphragm stiffness was lost by joist splitting at a displacement of approximately 4 inches. Lag Screw Forces The two outermost lag screws were in tension on one side of the ledger board centerline and in compression on the other side, as expected. The sum of the forces in all the lag screws located in the tension region of the deck agree well with the calculated overturning tension force (Figure 4 , page 29 ), but did not show any visible signs of withdrawal yielding. Hold-Down Behavior and Geometric Effects If the deck behaved as a rigid body, the tension chord forces could be calculated using simple statics as shown in Figure 5. However, due to the flexibility of the deck, the measured forces in the hold-down connectors were dramatically different than anticipated. The hold-down expected to resist overturning tension forces actually diminished to zero as the deck deformed. The hold-down installed on the compression chord actually had significant tension force due to a geometric prying action caused by large joist rotations (Figure 6 ). This geometric effect would be diminished if the deck stiffness were increased, such as by installing the decking diagonally. According to the 2008 Special Design Provisions for Wind and Seismic (AWC, 2008), shear walls and diaphragms sheathed with diagonally-oriented boards results in a four-fold increase in stiffness when compared to horizontally-oriented sheathing. Also, if the joist connections to the ledger had low withdrawal capacity, such as when nails are used in the hangers, or toe-nails, then the

Figure 6: Plan view of deck joist rotation and resulting “prying” effect on hold-down.

tension hold-down connection would be expected to function as intended. Design Implications Joist hangers: Joist hangers are typically rated for gravity (vertical) loads. When a deck is loaded laterally, the outermost joists are loaded in tension. Joist hangers are not load-rated in tension (i.e. joist withdrawal from the hanger). Preliminary experiments revealed that joist hangers utilizing a toenailed fastener orientation did not perform well when the toe-nailed connection was subject to tension loads. As such, hangers used in this project had fasteners installed perpendicular to the joist faces (Figure 7 ). Joist hanger manufacturers generally permit joist hangers to be installed with either nails or screws, as specified in appropriate technical literature. In this project, screws were used with the joist hangers to meet the provisions of IRC-2009 Section R507.1 and IBC-2009 1604.8.3, which both state that the deck attachment to an exterior wall shall not be accomplished by nails loaded in withdrawal. These provisions have been widely interpreted as applying to the deck ledger attachment; however, they should equally apply to deck joist hanger attachment to the deck ledger needed to complete the lateral load path from the deck to house. Calculating the allowable withdrawal and lateral capacity of fastener groups (10d common nails versus #9 SST SD screws) that attach the hangers (10 fasteners into the ledger, six fasteners into the joist) determined the withdrawal design capacity for screws was 750 lb; whereas, the capacity for nails was 150 lb – a five-fold difference. One reason for the large difference is the 75% reduction in withdrawal capacity for smooth-shank nails subject to wet/dry cycling specified in Table

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10.3.3 of the National Design Specification® (NDS®) for Wood Construction (AWC 2012). Relying on any withdrawal capacity of joist hanger connections having nails subjected to tension is a potentially unsafe practice, in violation of model code provisions, and does not provide an element of structural redundancy. Some level of structural redundancy is recommended, even though in ideal laboratory conditions it was shown that sufficient withdrawal capacity could be provided by joist hanger connections when screws are used. It is important to note that both deck tests were conducted in a laboratory setting where materials were not exposed to environmental factors such as wet/dry cycles, and there was no wood decay or fastener corrosion present. Ledger attachment: Deck ledgers were attached with 0.5-inch diameter lag screws in a staggered pattern as specified in IRC Table R502.2.2.1. The deck ledger-to-house attachment was for the conditions tested. When no tension hold-down connectors were used, the outer two lag screws carried most of the withdrawal load with no visible signs of failure (Figure 4). Tension hold-down: The deck with tension hold-downs behaved in a counterintuitive way. The flexibility of the deck allowed significant rotation of the deck joists within the joist hangers. This resulted in a geometric “prying” effect that caused zero tension in the “tension hold-down” and significant tension in the “compression hold-down,” as shown in Figures 5 and 6. The hold-down connectors would behave in a more predictive manner if the deck lateral stiffness were increased. While hold-down devices did not appear to significantly improve deck performance in the two decks tested that utilized screws in the hangers, hold-down devices do provide some

Joist Hanger Issues Preliminary tests using deck joist hangers that employed a double-shear (toe-nail) type of fastening to the joist did not resist joist pullout well. This is not surprising since joist hangers are optimized for resisting vertical loads rather than lateral. To overcome this weak link in the lateral resisting system, a hanger was chosen that incorporated perpendicular fasteners into the deck joist, and hanger-manufacturer approved screws were substituted for nails (Figure 7 ). The hanger selected was Simpson Strong-Tie (SST) Model No. LU210 (and concealed flange No. LUC210Z on the two edge joists). The LUC210Z had a “ZMAX” coating, which is classified as a medium level of corrosion resistance. At the time of the testing, the LU210 hanger was not available in the ZMAX coating. Based on the environment and materials used, the design professional should take care to specify appropriate level of structural redundancy for decks in service that naturally experience different levels of deterioration or may contain installation errors such as improperly installed lag screws.

Conclusions • The deck joist-to-ledger connection proved early on to be a weak link in the lateral load path. The problem is twofold: 1) smooth nails when loaded in withdrawal easily pull out of the deck ledger, and 2) some joist hangers use a toe-nail type of attachment of the joist to the hanger that do not engage enough of the joist, resulting in tear-out. This weak link was reinforced by using joist hanger manufacturer-approved screws and a joist hanger that had three screws on each side of the joist that were perpendicular to the joist securing the joist to the hanger. • For two specific laboratory deck configurations that utilized screws in the deck joist hangers, no significant impact on short-term deck strength and stiffness was observed when two tension hold-downs were installed. A similar result would not be expected had nails been used in the joist hangers, since wet/dry cycling causes nails to lose 75% of withdrawal capacity. • While the hold-down devices did not appear to significantly improve lateralload deck performance in the two

corrosion protection for all hardware used in a deck (guidelines can be found in DCA 6, downloadable at www.awc.org /publications/dca/dca6/dca6-09.pdf). The joist hanger manufacturer permits their hangers to be installed with either nails or screws. IRC-2009 Section R507.1 and IBC-2009 1604.8.3 both state that the deck attachment to an exterior wall shall not be accomplished by nails subject to withdrawal. These provisions have been widely interpreted as applying to the deck ledger attachment; however, they should also apply to deck joist hanger attachment to the deck ledger needed to complete the lateral load path from the deck to house. Joist hanger screws were #9 (0.131 inch diameter, 1.5 inches long) SST StructuralConnector Screws (Model No. SD9112) and #10 (0.161 inch diameter, 1.5 inches long) SST Structural-Connector Screws decks tested, these devices do provide a level of structural redundancy that may be needed in the event that lag screws used for the ledger connection were not properly detailed per the IRC Table R502.2.2.1 and correctly installed per the lag screw requirements of the NDS. • Hold-downs used in lateral load deck tests exhibited significant counterintuitive behavior. This outcome was due to geometric effects caused by large lateral deck displacements and rotation of the deck joists inside the hangers. • Testing was terminated before an ultimate strength was achieved at a load of approximately 7,000 lb for both decks. The two lag screws nearest the deck tension chord experienced the largest forces, yet did not fail in withdrawal. These results point to the effectiveness of 0.5-inch diameter lag screws when selected and installed per the IRC deck ledger connection provisions and NDS. • The results obtained in this study generally apply to decks with an aspect ratio of 1:1 and less, where aspect ratio is defined as the deck dimension perpendicular to the house divided by the dimension parallel to the house. The study results should not be applied to decks having an aspect ratio greater than 1:1, as the failure modes and deck behavior may substantially change.

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March 2014

Figure 7: Deck joists had perpendicular fasteners into the joist and used joist hanger manufacturer approved screws.

(Model No. SD10112). These screws have a Class 55 2006 IRC-compliant mechanical galvanized coating that is required to resist corrosion. Additional research is planned to study other deck constructions and aspect ratios and to investigate other methods to achieve lateral stiffness and load capacity, and structural redundancy for new and existing decks.▪

Acknowledgments Donations of materials from Simpson Strong-Tie and Trex Company are gratefully acknowledged.

Brian J. Parsons, former graduate student, Civil and Environmental Engineering. Donald A. Bender, P.E., Director, Composite Materials & Engineering Center, and Weyerhaeuser Professor, Civil and Environmental Engineering, Washington State University, Pullman, WA. Donald may be reached at bender@wsu.edu. J. Daniel Dolan, P.E., Professor, Civil and Environmental Engineering, Washington State University. Daniel may be reached at jddolan@wsu.edu. Robert J. Tichy, Research Engineer, Composite Materials & Engineering Center, Washington State University, Robert may be reached at tichy@wsu.edu. Frank E. Woeste, P.E., Professor Emeritus, Virginia Tech University. Frank may be reached at fwoeste@vt.edu.

Production

InSIghtS

55% Modeling (includes stick modeling and connecting)

new trends, new techniques and current industry issues

Modeling Max 10% Stick Modeling (Main Members only)

45% Model Connecting

S 3D In-Model Shop Drawing Review By Mark Hershberg, P.E., S.E., LEED AP, John Reitmeier Jr., Wayne Morrison and Emil von Roth, P.E., S.E.

Mark Hershberg, P.E., S.E., LEED AP, is a Principal at KPFF Consulting Engineers in Los Angeles. He may be contacted at mhershberg@kpff-la.com. John Reitmeier Jr. is a Contracts Manager with The Herrick Corporation. He may be contacted at jreitmeier@herricksteel.com. Wayne Morrison is a PreConstruction Manager with The Herrick Corporation. He may be contacted at waynem@herricksteel.com. Emil von Roth, P.E., S.E., is an Associate at KPFF Consulting Engineers in Los Angeles. He may be contacted at evroth@kpff-la.com.

teel fabricators have been using three dimensional models to detail structural steel for many years now. In many modern shops, all information required for the fabrication of a project is included in the detailed model, including dimensional information, material grades, piece marks and specification of welds. Much of the CNC equipment used in the fabrication is programmed directly, using digital data derived directly from the models using a combination of commercial and custom software that is tailored to the particular equipment and practices used by the fabricator. Modern beam line equipment even etches the locations of connecting pieces, part marks, weld symbols, faying surfaces and shear tab connections to guide the manual assembly processes that follow. With automation at such a level today, the traditional two dimensional shop drawing is used for fewer purposes than before. In the state of the art shop, the two dimensional drawings are primarily used as a reference document for piece assembly by shop labor and inspection. Yet, in most cases, the two dimensional drawings remain the primary means for the engineer of record to review the steel fabrication information for conformance with the design intent. This requires the inclusion of far more information than is required for their use in the shop. Further, incorporating comments made in the structural engineer’s review requires modification to both the three dimensional model and two dimensional drawings. Even with modern detailing software automating the generation of two dimensional piece drawings from the fabrication model, approximately onethird of all detailing hours are spent creating and editing the two dimensional drawings (Figure 1). The Herrick Corporation, with the help of one of its subcontracted detailers, has recently developed a process to enable the structural engineer of record to review, comment on and approve the fabrication model directly without the need

32 March 2014

35% Drawing Production Drawing production includes: (a) Drawing extraction (b) the req’d editing

Figure 1: Typical timeline for development of detailing model and shop drawings.

for any two dimensional drawings. This process, dubbed “3D In-Model Review”, is enabled by a proprietary application created using the Tekla Open Application Programing Interface (API) and operates in conjunction with the Tekla Structures software platform. For Herrick, having the structural engineer of record review within the model allows the production of shop drawings to occur all at once, following the approval of the model by the structural engineer. This economizes detailing hours for drawing production and streamlines the incorporation of review comments into the final detailing model. The time required to prepare packages for the structural engineer’s review is also minimized, since shop drawing production does not occur prior to the submittal. This can benefit the project by allowing additional time to complete design for a fast track construction schedule, or by allowing an earlier construction start for an established design. The 3D In-Model Review application was designed to maximize the use of the three dimensional model and mimic the processes involved in a traditional paper shop drawing review that engineering firms have grown accustomed to. Models are submitted as Tekla Structures files through a file sharing site or FTP server, and remain with the design team as a record document. The models are broken down in manageable segments comparable to traditional shop drawing packages, typically broken into erection tiers and sections. The extent of a typical submittal is shown in Figure 2. Upon opening the model, the structural engineer is presented a default filtered three dimensional view of the specific steel members that are to be reviewed in the submittal. All members in the submittal package are color coded, representating current submitted status prior to any design team review status. Members colored cyan have no open issues, and members colored orange have a comment from the fabricator to the SEOR, for example. The reviewing engineer can systematically interrogate the pieces of interest for the review using the 3D In-Model

Figure 2: Typical size of a 3D In-Model Review submittal for SEOR review.

Figure 3: Display of connection detail in 3D In-Model Review application.

Figure 4: Three dimensional view of model with 3D In-Model Review main approval dialog.

Review application, which provides a summary information dialogue for data relevant to the review, such as member type, profile and material grade. After selecting an individual main member and viewing the basic information through the approval dialogue window, Tekla Structures automatically generates additional views with detailed information on the connections such as the quantity, size, spacing and type of bolts, size and type of welds, and connection plate grades and dimensions (Figure 3). The main approval dialogue window (Figures 4 and 5) provides the engineer numerous fields to comment in, including the ability to attach commented screenshots, design drawings, sketches, etc. After the engineer has reviewed the member and determined the approval status, the comments are permanently recorded in the model for transmittal back to the detailers. The visual representation of the member changes in the Tekla interface to one of the following: green for “Approved”, yellow for “Approved as Noted” or red for “Revise and Resubmit”. This visual cue is used by the reviewing engineer to track progress and to rapidly convey the approval status of the members within the model. KPFF Consulting Engineers is currently partnered with Herrick as part of an integrated project team on a new 560,000 GSF hospital facility in Southern California, which utilized

Figure 5: Enlarged view of the 3D In-Model Review main approval dialog with image attachment illustrating the review comments.

the 3D In-Model Review process. A review team of four engineers collaborated on reviewing the fabrication models for conformance with the structural design drawings. Work was split between the review team with multiple reviewers working in parallel on each package. Each reviewer was assigned a family of elements (e.g. beams, columns, etc.) and worked geographically through the individual members in the model using the color key to track progress. All information needed for review was quickly available to the reviewer through dialog boxes and views generated automatically through Tekla Structures. All miscellaneous pieces, such as connection plates and stiffeners, are associated with the members they attach to, so all pieces are reviewed through viewing the main members. After the initial training and learning curve on the use of the 3D In-Model Review Application, the review team found productivity to increase on each subsequent submittal. The reviewers found that the first review of an individual member was typically slower than in a two dimensional drawing review due to the multiple dialogs and views required to encompass all of the pertinent information. After reviewing one instance of a connection detail or common member, however, all other instances could be immediately identified and approved or commented

STRUCTURE magazine

March 2014

on in one action. This led to an increase in productivity as the review progressed. Though the review of the models is ongoing at the time of publication, KPFF projects that review hours per ton of steel will be lower using the 3D In-Model Review process when compared to the review of two dimensional drawings or three dimensional models with embedded two dimensional piece drawings. A final, immeasurable benefit for the review team is the dynamic nature of the three dimensional visualization of the project. There is an immediate familiarity with the three dimensional image of the work under review, which allows the review team to see each element in the context of the surrounding members and in complete detail before the design is sent off to fabrication. The interface also removes some of the drudgery associated with paging through huge stacks of paper drawings. The three dimensional visualization of the model was also useful for communication between team members, as screen captures could easily be annotated with questions or comments and emailed or included in the comments as file attachments. Though still in its infancy, 3D In-Model Review is proving to be a promising tool with productivity benefits for both fabricator and designer. As the tools continue to improve, they are likely to become more widely used to deliver steel building projects ever faster.▪

Historic structures significant structures of the past

Trenton Bridge looking easterly from Pennsylvania.

T Trenton Bridge First Bridge across the Delaware River By Frank Griggs, Jr., Dist. M. ASCE, D. Eng., P.E., P.L.S.

Dr. Griggs specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He was formerly Director of Historic Bridge Programs for Clough, Harbour & Associates LLP in Albany, NY, and is now an independent Consulting Engineer. Dr. Griggs can be reached at fgriggs@nycap.rr.com.

heodore Burr had just finished his Union Bridge in Waterford, NY (STRUCTURE, February 2014) when he was called to Trenton, New Jersey to build a toll bridge across the Delaware River just above the falls. The New Jersey Legislature passed an act for a bridge on March 3, 1798 and the Commonwealth of Pennsylvania passed a similar law on April 4, 1798, entitled “An Act to authorize the Governor of this Commonwealth to incorporate a company for erecting a Bridge over the River Delaware at or near Trenton.” There was a high degree of risk involved in building a bridge over a major river that was noted for flooding and ice jams. This made it difficult to sell shares in the bridge company and sufficient funds were not raised to start construction until May 21, 1804 when the cornerstone was placed. Burr was given charge of designing and building a five span bridge over the Delaware River between then Trent’s Town (now Trenton) New Jersey, and Morrisville, Pennsylvania, thus making it an interstate bridge. He designed a variation of a tied arch with iron verticals and a large upper wooden chord made of laminated planking. His iron verticals were suspender chains similar to those used by James Finley in his suspension bridges of the time, in that they were a series of loops. The wrought iron loops were forged together into a chain to arrive at the proper length, inserted through the wooden upper chord and held with a pin. Where did Burr get the idea for a tied arch in wood? Did he see his arch as a Finley Suspension Bridge reversed, or did he follow the statement of Thomas Young that “as the flexible chord hangs, so does inverted the rigid arch stand?” Did he know of a British patent #2,109 issued to James Jordan on May 24, 1796? Jordan claimed he could suspend “canal troughs or roadways from laminated wood or iron arches, the end connections to be so arranged that there was no thrust on the abutments.” Whatever the source, Burr

34 March 2014

designed and built a bridge that was unique to the United States, and one that was noted around the western world. During construction, which began on May 21, 1804, the company built the piers and abutments to a height they thought was well beyond any height the Delaware River could reach. After completing the stonework, but before the wooden superstructure was begun, the Delaware rose to a height that covered the newly placed piers and abutments. The contractor immediately raised the stonework above the highest water just experienced. James Mease’s Geological Account of the United States published in 1807 noted “The bridge at Trenton, over the Delaware River, thirty miles above Philadelphia, justly claims distinguished notice in the present work. It is to be regretted that the ingenious architect, Mr. Burr, had not given to the public a detailed account of a work of such great and general utility, the execution of which does him so much honour.” He wrote, “The superstructure of the bridge consisted of five wooden arches, respectively 203, 198, 161, 186 and 203 feet in the clear, each composed of five great arched ribs rising from the chord in the proportion of 13 feet to 100. These ribs were made of four-inch pine planks, a foot wide and from 35 to 50 feet long, built up into a thick, laminated rib, three feet wide [deep]. The relative placing of these ribs left two openings of 11 feet each in the center of the bridge for carriage ways, and two more, each 4 feet 6 inches wide on the sides for footwalks. The ribs were spaced and bound together on the top circumference of the arches by ties and diagonal braces, fastened to the ribs by bolts and screws at intervals of 8 feet. The floor was suspended from the ties by perpendicular iron rods, securely fastened in the wood. Wing arches and diagonal braces were effectively used throughout to eliminate all motion between the parts of the bridge, thus making it a rigid and solid structure.

Triple Protection Against Corrosion Early illustration of Trenton Bridge arches/roof and Pennsylvania Portal.

The bridge rested upon the abutments and four piers, all of stone. The piers were made about one-fourth higher than they had originally been planned… The ends of the piers upstream were semicircular and, after rising five feet, gradually receded to the top, where they were finished off in a half-dome. These piers were 62 feet long and 20 feet deep.” Herman Haupt, in his 1851 book on Bridge Construction, wrote that the diagonals were inserted and anchored into cast iron shoes at the lower end, and tied to the upper arches with iron straps acting as counter braces [ties]. They were twin 6 x 10-inch timbers, spiked to the lower chord and secured to the arches by long 2-inch by ½-inch iron straps… and indicated the links of the suspenders were of 1⅛-inch square bars, about 4 feet long and 5 inches wide, passing flatways through the arches and between the chords and counter braces; a key passing through the link on the top of the arch. The lower chord was made of a pair of 6½ x 13½-inch timbers and, again according to Haupt, “are connected by means of long straps of iron passing around the end of the arch at the skew-back, and bolted through the chords.” Haupt also described what he called spur arches, noting: “On the sides are large spur arches of the same dimensions of the main arch of the truss, extending from a point 8 feet outside of the truss on the abutments and piers, and terminating within 44 feet of the centre-spiked at the point of intersection

of the arches of the main truss.” In summary, the arches were supported by a combination of the diagonal ties, tension chords and bearing on the piers and abutments. Mease continued his write up of the bridge by stating, “the three great objects, convenience of travelling, strength, and durability, are all happily united in the model adopted...nor has ornament been wholly thrown aside. The access to the Bridge, on either side, and throughout the whole extent of the platform, presents to the traveller a plane, without any sensible rising.” Mease finished his report on the bridge with: Our bridge, we are assured by Mr. Burr, combines double the strength of either [Waterford or Fort Miller] of them; but what constitutes the greatest excellence of the Delaware bridge, is the prospect of its durability… the permanency of the stone work is not to be questioned, and by the proposed covering, the stamina, or main parts of the wooden superstructure, will be effectually protected from decay by the wet, while those parts we exposed to injury from the weather, are all susceptible of the most complete repair. The entire structure was roofed shortly after construction. Instead of the standard gabled roof running along the axis of the bridge, Burr used a conventional gable over his piers with a gabled roof running transversely over his arches. The only walls that were covered were over the piers, leaving the arches exposed on the side. The primary reason he did not cover his walls and have the normal gable roof over his arches, besides cost, was the large wind loads that would have made if difficult, if not impossible, to brace the arches. The bridge was opened on January 30, 1806 to a large celebration that was described in the United States Gazette on February 5, 1806: On Thursday the 30 th ult. the bridge over the Delaware River near this place was

STRUCTURE magazine

March 2014

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The entire bridge was covered by a roof of cedar shingles, and was enclosed at each end. Originally there were high and elaborate fronts, both on the New Jersey and the Pennsylvania ends of the bridge, with great arched doorways over the carriageways and footwalks. Balustrades four feet high ran along the whole length of the bridge, outside of the footwalks, to protect the pedestrians.

Stevenson drawing of Trenton Bridge.

Picture looking at southerly side of bridge with reinforced arches for railroad, beginning of iron truss bridge on left.

opened to the public. The collection of people assembled on this pleasing occasion was great. A procession was formed, commencing at the house of Amos Howell, from whence they set out at 12 o’clock. Burr and his workmen led the parade across the bridge, while 17 cannons [one for each state in the Union] at each end of the bridge fired throughout the procession. Once across the bridge, the entire procession switched from foot to carriages and came back across the bridge in the same order. When they returned, a ceremony was held primarily to honor Burr and his men. The President of the bridge company made the following address: To you sir, as their principal architect, great commendation is due, for the fidelity, zeal, and unremitted industry, you have manifested in the superintendance and judicious management of this important and arduous undertaking; being your self equally expert to work in stone and in iron as in wood, you have given a direction and energy to each of those branches which tended greatly to facilitate their progress. The skill and ingenuity discovered in the design and plan of our bridge, has already been the subject of high encomium by a discerning public, as well as by the Board; and it now only remains for them to testify their entire approbation of the masterly stile in which the workmanship in every part appears to be executed, but as it regards neatness and strength, to those who are judges of architecture the simplicity and justness of the plan of our Bridge, and the order and symmetry of its structure, cannot fail to be highly appreciated, while to those artists in the science of

Bridge building, it must prove a valuable model and worthy of future imitation… At a reception Abraham Hunt made another toast to Burr as follows: Theodore Burr – may the Trenton Delaware Bridge prove as useful to the publick, as the simplicity and strength of the plan, and the skill and ingenuity displayed in the workmanship thereof, will reflect credit on its Chief Architect. David Stevenson, a noted British Engineer, visited the bridge about 30 years after its opening and obtained a copy of the drawings. He reproduced them in his 1838 Sketch of the Civil Engineering of North America, Comprising remarks on the harbours, river and lake navigation…and other works in that country. He gave a detailed description of the bridge and brought the design to an audience in England, but there is no evidence that anyone in England picked up on this design. In 1835, rails were laid on the north wagon road, and trains, generally hauled by horses, crossed the river. This required the southerly carriageway to handle two-way traffic on a width that was inadequate. In the summer of 1848, a decision was made to widen the bridge to accommodate two-way carriage traffic, one sidewalk and one railroad track. To handle this additional traffic, they simply removed the roof, all overhead bracing and the deck structure between the two southerly most arches. Then they moved the southerly arch five feet south on the existing piers that were wide enough to accommodate this shift. They strengthened the two southerly arches by placing additional arches of planking over them and bound the new arches to the old with short ties. After the arches were strengthened, they placed new overhead bracing to stabilize them. In the great Delaware River flood of 1841, only Timothy Palmer’s Easton Bridge, opened

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in 1806, and Burr’s Trenton Bridge survived. A steel bridge was built on the enlarged piers for the Pennsylvania Railroad in 1892. In 1898, another parallel steel railroad span was built on expanded piers, making a total of three lines of trusses on the same piers. The two railroad spans were removed in 1903 when an adjacent concrete arch bridge for the railroad was built. The current vehicular bridge, called the Lower Trenton Bridge, was built in 1928 on Burr’s piers and is illuminated with a sign stating, Trenton Makes the World Takes. Burr went on to build many bridges across the Mohawk, Susquehanna and other rivers. All of these, however, were similar to his Waterford Union Bridge. He never used the Trenton Bridge design in the future, even though it was a success.▪

Burr’s Arches modified to carry railroad track.

the TRanSfoRM aTion

HiSToRiC fiRST naTional STaTE Bank BUilding of the

Newark New Jersey – Part 2 By D. Matthew Stuart, P.E., S.E., F.ASCE, F.SEI, SECB, MgtEng and Ed D. Cahan, P.E., S.E.

ennoni’s involvement with the adaptive reuse of an existing historic building included the development of several innovative structural solutions. Part 1 of this article (February 2014) provided a history of the building and the initial major renovation challenges (Figure 1). This article will discuss life safety improvements and enhancements to utilization of the 12th floor and roof spaces. Improvements in life safety codes since the original construction of this mid-rise building meant that the existing egress paths through the structure were no longer adequate to support the use of the structure as a hotel. In order to rectify this, the project added 800 square feet of floor space to each floor, containing a new stair and elevator. Accessing this new 9-story tower required demolishing a large portion of the existing east exterior wall. Supporting it presented significant structural challenges, because the footprint of the addition was located directly above the existing low-rise building but did not match its column grid. Pennoni presented two alternatives. The first consisted of vertical diagonal braces supporting the floor beams of the addition from the existing east exterior columns, such that each level of the new tower was essentially “corbeled” from the mid-rise tower. The second option consisted of new columns at the east face of the addition that would be supported by significantly reinforcing the existing beams of the existing low-rise roof to act as transfer girders. Ultimately, the design team chose the latter due to architectural restrictions in the geometry and location of the diagonal braces proposed by the first option. After establishing the support scheme for the stair and elevator addition, Pennoni concentrated on evaluating the impact of the new addition on the existing structure. First, the double spandrel beams were analyzed for the loads imposed by the reaction of the new adjacent framing.

Figure 2: New tie-in conditions at existing spandrels.

STRUCTURE magazine

Figure 1: Partially erected stair and addition framing.

Because the outer beam was primarily designed to support only the east exterior brick cladding that was being removed to allow for the addition, it was able to support the new beams of the stair tower without the need for reinforcement (Figure 2). Furthermore, the new net reaction at the columns supporting the outer beam of the double spandrel was approximately the same as the original condition. Transferring the lateral load of the stair tower addition into the lateral-load-resisting system of the existing mid-rise structure involved drilling horizontal epoxy dowels into the existing concrete slab at each level. Additionally, placing continuous reinforcing steel at the north and south ends of the addition, and doweling it into the adjacent existing concrete slab, created diaphragm chord members in the new slab. It was also necessary to investigate the distribution of new lateral forces created by expanding the sail area and square footage of the structure as a result of the additions. The lateral load resisting system of the existing structure had sufficient capacity to support these loads based on the exception allowed by the International Existing Building Code (IEBC) when the increase is no more than 10% over the original. The lateral resisting system used in the First National State Bank Building involved angle bracket kicker braces, encased in concrete, that attached the girders and columns together to form rigid moment frame connections at a number of locations throughout the

March 2014

Figure 3: New transfer girder connection at existing column.

Figure 4: Existing beam and slab strengthening.

building. This type of system was first used in the Reliance Building in Chicago in the mid-1890s. The final structural engineering task for the stair and elevator addition involved reinforcing the existing third floor beams of the low-rise that were to support the new columns above by adding transverse stiffeners beneath the new column flanges. These stiffeners also transferred the new concentrated column base reactions to new transfer girders built tight to the bottom flange of the existing beams. The transfer girders are designed to support the entire column reaction, which allows the existing beam to transfer the load through the stiffeners to the new beams below and still function as a part of the existing third floor framing. The top flange of the transfer girder was welded to the bottom flange of the existing beam to enable the transfer of any horizontal shear flow between the two members. The increased reactions where the transfer girders connect to the existing columns required stiffened seat connections to strengthen the existing conditions (Figure 3). Welding HSS sections to the column flanges increased the weak-axis bending capacity of the column section to handle the resulting eccentricity. Transforming the existing roof of the mid-rise building into an occupied terrace with outdoor dining required reinforcing the structure to support the new loads. While the existing drawings showed the steel framing, they provided no concrete slab information such as compressive strength or thickness. They also showed no roof sections, and it was suspected that the slope that existed for proper drainage involved a concrete topping slab in addition to the main roof slab. Roof cores at suspected high and low points enabled determination of the slab composition and physical properties, confirming Pennoni’s suspicions. This meant that it would be possible to use the existing slab in conjunction with new steel channel shear connectors on top of the existing beams to create composite action with the primary roof girders, thereby increasing their structural capacity. The existing roof beams spanning between girders would be overstressed as a result of the heavier loading associated with the new terrace. In order to prevent this, new steel roof beams were added between the existing ones in order to decrease the tributary width (Figure 4). The higher yield strength of the existing steel was taken into consideration, determined via coupon tests, as opposed to the strength provided by historical databases. An alternate method to reinforce the roof structure was to weld steel split T or WT sections to the underside of the existing beams and girders. However, this method was not used because of the excessive demolition that would have been required in order to remove a portion of the concrete encasement that provided fire protection for the existing beams.

Completing the renovation design required solving several other challenges, which included green initiatives to conserve energy via the introduction of a solar panel array along the top of the existing roof ’s copper cornice. The outrigger support structure did not have sufficient capacity; therefore, a new structural frame was designed to cantilever over and above the existing cornice roof, by connecting to the exterior column extensions located within the existing parapet. In addition, two new energy recovery units (ERUs) were required in order to provide conditioned air efficiently throughout the structure. However, the large size and weight of the ERUs made it difficult to find acceptable locations for their placement. The higher existing floor-to-roof height at the 12th floor made this location an attractive option. In order to limit the resulting loss of floor space, the units were stacked by providing a new steel dunnage frame supported by the existing floor girders at the lower ERU, and a new steel frame that spanned between the existing steel columns at the upper ERU (Figure 5). Taking advantage of the increase in steel yield strength found via coupon tests eliminated the need for reinforcing. The height at the 12th floor also allowed for the construction of luxury suites with lofts constructed of metal floor decks and cold-formed steel walls bearing on the existing floor beams, which were reinforced with new steel split T sections welded to the underside in certain locations where the existing structure was overstressed. Upgrading the vertical transportation systems with machine-roomless custom elevator cabs to fit the existing floor openings also posed

STRUCTURE magazine

Figure 5: New 12 th floor interior dunnage frame.

March 2014

challenges that included supporting the elevator sheave beams and providing the required overhead clearance. Ultimately, this resulted in the need to demolish a portion of the existing elevator penthouse roof structure and design a new one that met the needs of the new elevator manufacturer. The existing roof beams supporting the new penthouse also required strengthening. Upgrading the original fourstory stair and elevator tower located at the rear of the facility involved the complete demolition of the masonry walls and steel framing from the ground floor up, with a new six-story tower taking its place. A 2012 report entitled, The Greenest Building: Quantifying the Environmental Value of Building Reuse, produced by the Preservation Green Lab of the National Trust for Historic Preservation, found that when comparing buildings of similar size and function, environmental savings are almost always found through building reuse, rather than demolition and new construction. Additionally, the article found that reusing existing buildings provides benefits to the surrounding community and local economy. The adaptive reuse of the First National State Bank Building is an exemplary rehabilitation project highlighting these social, economic, and environmental benefits by restoring pride in a historic structure and creating revitalized business in an evolving neighborhood, all while decreasing the overall carbon footprint (Figure 6 ). This transformation would not have been possible without creative, yet simple and constructible, structural engineering solutions for the complex renovation changes.▪ D. Matthew Stuart, P.E., S.E., F.ASCE, F.SEI, SECB, MgtEng (MStuart@Pennoni.com), is the Structural Division Manager at Pennoni Associates Inc in Philadelphia, Pennsylvania. Ed D. Cahan, P.E., S.E. (ECahan@Pennoni.com), is a project engineer with Pennoni Associates Inc in Philadelphia, Pennsylvania.

Figure 6: Rendering. Courtesy of DeRosa Group Architects.

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STRUCTURE magazine

March 2014

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An Overall Strong Year for Steel Expected, Companies Say Some Government Projects Cut Back as Commercial Jobs Grows By Larry Kahaner

ompanies in the steel fabrication business are expecting a strong 2014 as firms continue to innovate, improve their existing products and focus on their particular sectors. “We are seeing some strong activity in large commercial construction projects in select metropolitan areas. These are primarily high-rise residential properties,” says Kevin Bates, Vice President Sales & Marketing at MMFX Steel Corporation, (www.mmfx.com) in Irvine, California. “We have seen some slowing in highway and bridge construction recently, but this has really varied state by state in the U.S. States having robust bridge construction markets have put through state level highway funding appropriations versus being highly dependent on federal money. Additionally, states that have some level of acceptance to allow Public-Private Partnerships are seeing good construction activity.” At the Vulcraft/Verco Group (www.nucor.com) in Norfolk, Nebraska, Michael Klug, Marketing Coordinator, New Products and Market Development, notes: “The re-shoring of American jobs will surely push our market in the right direction; this is, in large part, due to our domestic energy supply that continues to grow and create an environment that is conducive for manufacturing.” Others, like Bob Allen, U.S. Construction Hardware Product Manager of ITW Building Components Group, (www.itwbcg.com) in Pompano Beach Florida, suggest that companies that have survived the past downturn will do well simply because they have weathered a severe storm and have come out stronger. “They all believe that the worst is behind them. There’s a lot less of them to meet market demand, so from that standpoint, I feel the ones that survived are going to be in good shape.” Allen notes that his company will be introducing a series of face mount hangars within the next 30 days. “It’s a new series for us, one that will serve as a replacement for up to four diﬀerent series that our competitors currently oﬀer.” He adds: “That, by itself, will be a huge advantage. The load carrying capacity of these face STRUCTURE magazine

“We are seeing some strong activity in large commercial construction projects in select metropolitan areas.” mount hangars will be the highest in the industry… It fulfills a need that is one of the most sought after among our core customers.” (See ad on page 46.) Klug says that his company is always looking for ways to take care of its customers which range from steel fabricators, to GC’s, to erectors, to large corporations and private owners. “Being a part of Nucor Corporation gives us an unmatched supply chain, as well as a wide variety of other steel products that we can oﬀer.” He adds: “We work closely with Tekla, SDS/2 and Revit to ensure that we are leading the industry in the use of BIM.” MMFX’s Bates says that his company has just introduced its ChromX4100 Grade 100 reinforcing steel. “This high-strength steel, meeting all of the mechanical properties of ASTM A1035, is targeted at applications that can benefit from using less reinforcing steel.” He adds: “The ChromX4100 complements our other product, MMFX2, which provides uncoated corrosion resistance along with the Grade 100 high strength properties. The chemistry of the ChromX4100 has been modified slightly from the MMFX2 to lower the corrosion resistance and as a result lowering its production costs. This makes the ChromX4100 rebar ideal for applications needing the higher strength but not requiring the same level of corrosion resistance.” As far as trends are concerned, Bates sees reinforced concrete structures starting to increase the use of high strength reinforcing steels. “As designs are requiring more steel, buildings are getting taller. The natural evolution is towards taking advantage of higher strengths. Grade 75 steel is seeing increased use. MMFX Steel has been oﬀering Grade 100 steel for 12 years and are now seeing its

March 2014

“As designs are requiring more steel, buildings are getting taller. The natural evolution is towards taking advantage of higher strengths. ... We see a future that will move toward even higher strengths when combined with higher strength concrete mixes.”

use pick up since the recent changes to construction codes. We see a future that will move toward even higher strengths when combined with higher strength concrete mixes.” Jason Hoover, Industry Outreach Executive for SidePlate Systems, Inc., (www.sideplate.com) who works out of Strongville, Ohio, says that the company in 2013 debuted a new SidePlate Bolted moment connection option for R=3 projects. “For projects with higher seismic criteria, SidePlate connections are now listed as prequalified moment connections in Supplement #2 of the AISC 358-10 standard.” Hoover says that SE’s will appreciate its features. “The SidePlate Bolted connection delivers the same stiﬀness benefits as our fillet-welded connection, but the bolted option has no field welding at all so it’s even easier to erect. As of today, we only have a few data points for actual completed bolted projects, but responses from fabricators and erectors have been extremely positive support of the time and money saved at each stage.” He adds: “On the seismic side, the SidePlate connection’s prequalifications from other agencies predate the original AISC 358 standard, but we recently decided to go through the eﬀort of getting AISC’s thirdparty approval as well. We’re happy to say that their committee reviewed and approved of all of our testing data and results. The new Supplement #2 to AISC 358-10 also includes an extensive commentary on the history and evolution of SidePlate connections that will be helpful for engineers looking for more background.” Hoover says that SidePlate’s business is doing very well, and it’s growing. “Our biggest market is healthcare, which has been steady. The commercial market seems to be picking up some steam, and

while we’ve historically worked on a large number of government projects, those have dried up considerably. Our customers seem to be cautiously optimistic about the near-term general construction forecast, and we’ve heard rumblings that the steel mills are warning of longer lead times as a burst of construction activity is expected in mid-2014.” At ITW’s TrusSteel Division, (www.trussteel.com) oﬃcials report that business has been steadily increasing as the residential construction market improves. Dave Dunbar, National Sales Manager, says: “Cold-formed steel trusses are the ideal solution for pitched roofs requiring non-combustible materials. With improvement in the retail, oﬃce and hospitality sectors, we are projecting continued growth. Two core sectors for our type of construction, assisted living and education, will see mixed results based on regional demographics, and government construction is slowing due to reduced government spending.” TrusSteel recently introduced a new chord shape, the TSC300. “The TSC300 chord provides all the benefits of the TrusSteel continued on page 46

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STRUCTURE magazine

March 2014

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TSC400 chords, but is more cost effective because it maximizes the effective area. This 3-inch tall chord is utilized for mid-span ranges of 35 to 50 feet at both 24-inch on-center and 48-inch oncenter, depending on load conditions,” says Dunbar. The TSC300 was the result of collaboration with the company’s customer base. “This new chord was identified as an opportunity to help them offer more competitive systems.” (See ad on page 48.) Rich Madden, Marketing Manager at New Millennium Building Systems, (www.newmill.com) in Fort Wayne, Indiana, says that the company focuses on product development that can improve steel project performance and contribute to a range of total-project cost reductions. “Our research in this area has led to the release of our FreeSpan Beams, our expanded specifications for special profile steel joists, and our new Flex-Joist Gravity Overload Safety System.” He adds: “Our FreeSpan line of castellated and cellular beams features either hexagon or circular openings as a result of the castellation process. The beam is 50% deeper and up to 40% stronger than the original ‘parent’ beam, without adding any additional weight. So now you have a very efficient beam that can be exceptionally cost-saving as well as design-enhancing, because you can create wide-open, wide-span designs with fewer and narrower support beams. All the HVAC and electrical can run through the beams, light flows through the beams for added night safety, and the beams have a strong aesthetic appeal.” This spring, New Millenium will release its updated special profile joists catalog, which enables both the architect and engineer to design unique rooflines on their buildings for a reasonable price.

Courtesy of Bradken Inc.

Products include bowstring joists, gable joists, scissor joists, as well as other special shapes. The new catalog provides specification tables for over 40,000 possible design combinations. Madden says that The Flex-Joist Gravity Overload Safety System addresses a growing concern over unanticipated extreme roof snow and rain loads, which may in some cases be attributable to climate change. “The approach is to engineer a steel joist to flex much more gradually and deeply than a traditionally engineered steel joist in the event of an extreme overload, so as to establish the element of time delay. With sensors in place to detect this deflection, an early warning system can signal the need for evacuation, roof shoring or possible removal of the overload.” On the software side of the steel business, companies like Lincoln, Nebraska-based Design Data (www.sds2.com) are continuing to oﬀer new products and services. “SDS/2 2014 is the latest release continued on page 48

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STRUCTURE magazine

March 2014

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of our flagship product, SDS/2 Detailing,” says Michele Arnett, Marketing Manager. “The connection design capability has grown by leaps and bounds in this past version. The SDS/2 Erector is a new product that helps general contractors and erectors plan the job site, and the SDS/2 Approval helps project managers and reviewers to be much more productive.” She says that Design Data’s core customer base is steel fabricators and steel detailers involved in commercial and industrial fabrication, and that list has expanded. “With the recent growth of BIM, the SDS/2 customer base has grown to include designers, general contractors and other segments who benefit from the 3D model.” At RISA Technologies, LLC, (www.risatech.com) in Foothill Ranch, California, Director of Marketing Amber Freund says that RISAConnection v4.0 was recently released. “This new version includes the design of HSS connections, which are a hot topic in the engineering community today.” Freund says of the company’s products, “Unlike more basic connection software that is on the market, RISAConnection designs the connection for all of the applied forces, including axial forces due to beam tension/compression, and flexure on the face of tubes due to shear connection eccentricity.” She adds: “Ever since RISAConnection v1.0 was released, we have received an overwhelming demand to add HSS connection design to the software. The design criteria for HSS is relatively new (by the standards of the engineering community) and many engineers prefer to have reputable software help guide them in application of the latest design practices. By introducing

connection design through a RISA product, we hope to educate engineers on the limit states associated with HSS connection design and make them more comfortable with HSS on their everyday projects.” As for business conditions, RISA clients are getting more projects, says Freund. “We are hoping this is a steady trend that continues throughout this year. The industrial sector has remained pretty strong and we are seeing more commercial building projects as well… We are continuing to see BIM being used on projects. Where it used to only be used on larger projects, we are now seeing it used even on smaller projects like curtain walls. The integration between BIM and structural analysis software is important to ensure the accurate exchange of information during the design and construction processes.” (See ad on page 76.) Raoul Karp, Director, Product Management for Bentley Systems, Inc. (www.bentley.com) in Exton, Pennsylvania, sees three major drivers in the industry today: BIM adoption driving increasing structural complexity, tighter schedules requiring closer collaboration and competitive design environment pushing greater productivity. “In 2013 we had a release in each of our major product lines RAM, STAAD and ProStructures. The capabilities of which were squarely focused on addressing these key driving forces,” he says. For the RAM Structural System and RAM Concept, Bentley added several modeling, analysis and reporting productivity improvements including shearwall coupling beam design, code updates and 64bit capability to enable larger more continued on page 50

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STRUCTURE magazine

March 2014

introDucing the

HoW/2

Design ConneCtions with SDS/2

SerieS by SDS/2

true connection DeSign, not SimpLy connection veriFicAtion SDS/2 is the only system that provides true connection design — for individual members, as well as all interacting members in a structural joint.

compLete connection DeSign reportS

FuLL Joint AnALySiS Instead of choosing a connection from a library, SDS/2 designs the connection for you, based on parameters that you establish at the beginning of a project. All connections SDS/2 automatically designs will comply with the connection design code standards the user chooses.

learn more Want to see how simple it really is to design connections in SDS/2? Scan the QR code to watch SDS/2’s connection design in action.

SDS/2 provides long-hand calculations of all designed connections, which simplifies the verification process. Scan the QR code to view an example of SDS/2’s automatically generated calculation design reports.

cLASh prevention SDS/2 checks for interaction with other connections within a common joint. That means adjusting connections for shared bolts, checking driving clearances for bolts, sharing, adjusting and moving gusset and shear plates when required, and assuring erectablity of all members. All adjusted connections are automatically verified based on selected design criteria.

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complex structures and faster analysis. For STAAD.Pro and STAAD Foundation Advanced, they added half a dozen code updates including torsion design for AISC360, productivity enhancements with added advanced capabilities for machine and mat foundations, tension cables and for the first time the ability to run multiple design options of your structure on the cloud and compare and contrast results through your browser. For ProStructures, they expanded the steel modeling and detailing capabilities with advanced rapid stair, handrail and anchor bolt modeling. “We also have expanded our collaboration capabilities with IFC import and export, iModels for collaboration (scheduling, clash detection) with Bentley and non-Bentley products, and added new SolidWorks and updated Revit and Bentley product interoperability with Integrated Structural Modeling,” says Karp. (See ad on page 75.) We are seeing continued innovation in welding and cutting products as well, according to Mark Elender, Senior Vice President North American Sales, ESAB Welding & Cutting Products (www.esabna.com) in Hanover, Pennsylvania. He says: “ESAB has a long history of continuous product improvement and development in delivering high quality, leading-edge equipment and solutions to address the needs of steel fabricators. Our ICE process for Submerged Arc Welding (SAW) is a product of particular interest to structural steel engineers. The ICE process can increase a fabricator’s productivity while exceeding the requirements for critical welds. ICE is an elegantly simple technology; instead of adding energy, ICE exploits the excess heat from the Twin SAW

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process to melt an additional non-powered welding electrode. This results in double the productivity in deposition rates and in root welds when compared to single wire welding. ICE boosts output significantly without the investment in new welding systems, extra capacity, or additional skilled welding labor.” Adds Elender: “Also of interest to steel fabricators is our new Warrior – a multi-process welding machine capable of delivering up to 500 amps at 60% duty cycle. Warrior offers users a very good stable arc in multiple processes, including GMAW (MIG), FCAW (Flux-Cored), SMAW (Stick), and GTAW (TIG) welding, as well as HELIX PEDESTRIAN BRIDGE, SEATTLE, WA ACAG (Arc Gouging), and is easy to use and energy efficient thanks to stateof-the-art inverter technology.” Also new to the company’s Cutting Systems line is the Hydrocut LX waterjet shape cutting machine, a combined waterjet and plasma cutting system. “The machine uses a patented combination of thermal and non-thermal processes operating on the same gantry, allowing the machine to cut with the high accuracy of waterjet where needed, but employ the high speed and low cost of plasma whenever possible. Steel fabricators benefit from the use of both technologies on the same part. High precision contours can be cut with waterjet, while non-critical contours can be cut with plasma,” says Elender. ESAB’s new offerings are developed WINNER OF A from customer input. “We constantly solicit customers’ input to understand National Council their requirements and expectations so of Structural that we can design and deliver products Engineers that help solve their chalAssociations lenges. We are a full line integrated supplier, so we Outstanding engineer and produce the Project Award products we sell.”▪

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STRUCTURE magazine

March 2014

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Structural teSting issues and advances related to structural testing

Figure 1: Test layouts at SCL and Tyrell Gilb laboratories.

tarting in 2008, members of the project team, along with others, performed tests to show that the American Concrete Institute (ACI) 318-11 Appendix D concrete shear capacities for steel anchor bolts with small edge distances connected to wood sill plates were extremely low compared to actual tested values. As a result of the tests, a code change was made to the International Building Code (IBC) 2009 to allow the use of much higher shear values for anchor bolts connected to wood sill plates. That project comprised a twoyear effort driven by practicing engineers in association with an industry partner. The current project is again aimed at providing practical test results to design professionals and building code officials, but this time for the

Testing Tension-Only Steel Anchor Rods Embedded in Reinforced Concrete Slabs By W. Andrew Fennell, P.E., CPEng, SECB, Gary L. Mochizuki, P.E., S.E., LEED AP, Kevin S. Moore, P.E., S.E., SECB, Steven E. Pryor, P.E., S.E. and Geoffrey A. Laurin

Figure 2: Phase 1 edge and field testing.

52 March 2014

frequently specified (but not necessarily accepted) anchorage details in tension-only anchor rod systems. With the support of our industry partner (Simpson Strong-Tie® Company, Inc.), this project is moving closer to converting valuable test data into a useful design methodology and possible regulation modification. The current project is divided into two phases as follows: Phase 1 (completed in June 2012) proposed a testing protocol for the extensive Phase 2 testing. To validate and/or refine the proposed testing protocol, Phase 1 concluded with the construction and testing of three full-scale experimental specimens featuring both center-of-slab and edge-of-slab details. Phase 1 experiments were performed at Scientific Construction Laboratories, Inc. (SCL) in Lafayette, California (Figure 1). Phase 2 is currently in progress at Simpson Strong-Tie’s Tyrell Gilb Research Laboratory in Stockton, California (Figure 1). Phase 2 features full-scale podium slab sections with varying levels

of special detailing for both center-of-slab and edge-of-slab conditions. Interim results suggest that test data correlates well with calculated uncracked average ultimate capacities of anchors (without anchor reinforcing) using ACI 318-11 Appendix D with all design reductions removed.

Testing Protocol Development The research team selected the use of a monotonic loading protocol in an effort to understand controlling behaviors and failure mechanisms, and to establish data that is comparable to a majority of test data already available. Results of a literature search indicated that adding plates at the bottom of the embedded tension-only anchor rods would increase the size of the concrete failure cone, and hence the load capacity. The research team did not find any literature on reinforced concrete sections. Phase 1 experiments at SCL validated the testing protocol using a high-capacity testing bridge previously developed by Simpson Strong-Tie. Figure 1 depicts each 8-foot x 8-foot x 1-foot concrete specimen with anchors cast in the field and at the edge. The slabs were elevated off the ground to avoid any potential restraining effect on slab bending. Slabs were reinforced to prevent bending

Table 1: Initial Phase 1 testing results and typical seismic design demand.

Model Building

Likely Rod System

Nominal Tensile Strength

Break-out Capacities ACI-318-11 App. D

Tested

3 story over podium @ edge San Francisco

3/4" Rod ASTM F 1554 Fu=58 ksi

19.2 kips

±10 kips Need 1.2 x 19.2k

TBD

4 story over podium @ edge Seattle

1" Rod ASTM F 1554 Fu=75 ksi

44.2 kips

±10 kips Need 1.2 x 44.2k

TBD

5 story over podium @ edge Los Angeles

1-1/4" Rod ASTM A193 B7 HS Fu=125 ksi

115.0 kips

±10 kips Need 1.2 x 115k

TBD

Test 3-2. Edge Lafayette

1-1/4" Rod ASTM A193 B7 HS Fu=125 ksi

115.0 kips

Nuc,m = 34.0 kips Nc,5% = 20.4 kips Design = 10.7 kips

80.9 kips*

Test 2-1. Center Lafayette

1-1/4" Rod ASTM A193 B7 HS Fu=125 ksi

115.0 kips

Nuc,m = 70.3 kips Nc,5% = 42.2 kips Design = 22.2 kips

85.4 kips

*Note: Test 3-2 (embed at edge) had some reinforcement in the breakout zone.

failure per ACI 318-11, section D.3.3.2. To ensure that a bending failure would not occur, the amount of steel specified for the test was 75 percent more than the amount required by analysis. The slab was analyzed using RISA-3D as a compilation of plate elements, with the testing bridge as an assembly of beam elements.

Figure 2 and Table 1 show a summary of the Phase 1 experiments. A published testing report describes further details related to the setup and results. Table 1 compares some model building design demands with currently available design capacities. The limited design capacity for concrete anchorage is immediately apparent once building height exceeds three stories. continued on next page

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STRUCTURE magazine

March 2014

2/6/14 9:53 AM

In all respects, Phase 1 experiments verified basic assumptions valuable to the development of the Phase 2 test program. Adequate reinforcing steel prevented bending failures in the slabs. The tension breakout cone failures for the center (away from edge) tests occurred as anticipated; however, the capacity was higher than expected. It appears that the flexural steel crossing through the cone added to its capacity. The tension breakout failure at the edge occurred at a load substantially higher than the ACI 318-11 Appendix D prediction. This was likely due to the unintentional benefit of vertically oriented hooks at the ends of the top and bottom flexural bars that crossed the cone failure plane in several locations. To understand the connection behavior more thoroughly, future tests (Phase 2) would include additional steel specifically to reinforce the cone (anchor reinforcing) located near the anchor and crossing the anticipated failure planes, as well as control tests without any anchor reinforcing.

Figure 4: Cross-section of slab showing one-piece anchor reinforcement.

Lessons Learned Large, high-strength anchor rods, nuts, washers and couplers are all potentially special-order items (not available at local construction supply centers), with significant lead times exceeding four days. To avoid substitution requests, consider providing enhanced submittal requirements in specifications for anchorage in design. Engineers are constantly receiving feedback and/or criticism from contractors regarding constructability. The Phase 1 experiments were no exception. The “as-planned” geometry for the anchor rods was difficult to achieve, particularly at edge conditions. Appropriately scaled sketches are necessary to verify the constructability of the anchorage connection. Non-destructive testing (NDT) techniques were employed as part of the Phase 1 effort. This testing was used to confirm as-built ADVERTISEMENT–For Advertiser Information, visit www.STRUCTUREmag.org

Figure 3: Non-destructive testing (NDT).

Figure 5: Cross-section of slab showing two-piece anchor reinforcement.

• Three, Two, and One Story Comparison of Seismic and Wind Based on 2012 IBC / 2013 CBC.

clearances and/or geometry. NDT was limited to pachometers and ground-penetrating radar (GPR) methods (Figure 3). Based on NDT results, the research team confirmed that Structural Observation and/ or Special Inspection prior to concrete placement will provide good conformance with designed anchorage detailing. To ensure consistent construction quality, engineers might consider requiring and/or incorporating photo-documentation of installed anchorage and reinforcing around the anchor. While unique, this level of inspection and confirmation is important given the critical nature of most anchorages of this type.

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Phase 2 Testing

236

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The Phase 2 testing at Tyrell Gilb Research Laboratory involved 12 test specimens to research anchorage away from edges (i.e., field). These tests were designed to examine different conditions associated with various

STRUCTURE magazine

March 2014

flexural reinforcement ratios, as well as the effect of localized anchor reinforcement. Future testing will be performed with anchors located at the edge of slabs to determine the effects of anchorage reinforcement at the slab edges. All specimens had a concrete compressive strength of 5,000 psi and reinforcing steel per ASTM A615, Grade 60. The 8-foot x 8-foot x 1-foot slabs were tested while supported off the ground by use of a grid of 4x4 timber members aligned with the test frame located at the top of the slab above. A hollow-core hydraulic ram applied displacement-controlled loads to the anchor. A representative from Testing Engineers Inc. independently observed and documented all Phase 2 experiments. The initial Phase 2 experiments were run in pairs to test three main configurations: (1) anchors without additional anchor reinforcement; (2) anchors with one-piece inclined anchor reinforcement (Figure 4 ); and (3) anchors with two-piece inclined anchor reinforcement (Figure 5). For each of these continued on page 56

Table 2: Initial Phase 2 testing. See Simpson Upper Mat HS Anchor Anchor Peak Load Figure 6 Test ID Flexural Dia. (in) Reinforcing (kips)

A B C D E F

U666 A Inadequate

1.5

U666 B

Inadequate

1.5

U665 A

Adequate

1.5

110

U665 B

Adequate

1.5

U668 A Inadequate

1.75

Yes. 1-piece

135

U668 B

Inadequate

1.75

Yes. 1-piece

135

U942 A Inadequate

1.75

Yes. 2-piece

155

U942 B

Inadequate

1.75

Yes. 2-piece

148

U667 A

Adequate

1.75

Yes. 1-piece

197

U667 B

Adequate

1.75

Yes. 1-piece

194

U941 A

Adequate

1.75

Yes. 2-piece

223

U941 B

Adequate

1.75

Yes. 2-piece

237

configurations, two slab flexural reinforcement conditions were varied. In the first case, flexural reinforcement was insufficient to prevent plastic hinging (Inadequate). In the second case, flexural reinforcement was sufficient to prevent plastic hinging (Adequate). As shown in Table 2 and Figure 6, Phase 2 results indicate that under-reinforced concrete slabs will typically fail in flexure before a shear cone can develop. These results prove the adequacy of ACI 318-11 Appendix D considering a limitation on anchor capacity for locations where a plastic hinge can form as noted in D.3.3.2. Figure 7 shows failure surfaces and corresponding peak loads (average of two tests) for two different anchor reinforcement conditions. Test B has no supplemental anchor reinforcement; Test E has a one-piece inclined bar anchor reinforcement. Figure 8 also shows the failure cone and corresponding peak loads (average of two tests) for reinforced and unreinforced anchor

Figure 6: Initial Phase 2 test results. Pre-test photos courtesy of Testing Engineers, Inc. (TEI).

specimens. Test B has no supplemental anchor reinforcement; Test F has two-piece anchor reinforcement. With sufficient flexural reinforcement, a shear cone will develop as indicated in ACI 318-11 Appendix D. By adding anchorage reinforcement (bars dedicated to resist shear cone failure), the capacity of the anchor in tension can be significantly increased. ACI 318-11 section D.5.2.9 permits the strength of properly developed anchor reinforcing to be used in lieu of the tension cone breakout strength in determining design resistance. It is clear from the results of Tests “B” and “D” that neglecting the strength of the concrete breakout would be a highly conservative assumption for the tested condition. Test D was identical to Test B except for the addition of single-piece anchor reinforcing (Figure 4 ) that crossed the theoretical plane of the tension cone at a 45° angle. Cone failures controlled both test series, with the Test D cone projecting through the sloping sides of the anchor reinforcing.

Figure 7: Test “B” and Test “E”. Pre-test photos courtesy of TEI.

STRUCTURE magazine

It is interesting that, in these sloping sides, the vertical component of the nominal yield strength was 105 kips. Figure 7 shows that the ultimate capacity of Test D is nearly the sum of this 105 kips and the Test B results. Predicting the average uncracked ultimate tension cone breakout force of Test B using ACI 318-11 equation (D-6) with kc=40 (instead of 24; this removes reductions for cracks and reduction from average result to 5% fractile), and also properly considering the ANC /ANCO ratio in accordance with D.5.2.1 and D.5.2.8, yields a predicted strength of 83 kips. Combined with the 105 kips from the anchor reinforcing results in a predicted sum of 188 kips, which is also very close to the 195-kip average test result. As mentioned previously, it is thought that the horizontal flexural reinforcing that passes through the breakout cone increases cone capacity, but the effect remains undetermined. When attempting to correlate test data and calculations for tension cone breakout strength,

Figure 8: Test “B” and Test “F”. Pre-test photos courtesy of TEI.

March 2014

Acknowledgments The Structural Engineers Association of Northern California (SEAONC) provided a $10,000 grant through their 2012 Special Projects Initiative. In the Phase 1 experiments, the Simpson Strong-Tie Company (SSTC) generously loaned their loading bridge and helped guide the protocol development. In the Phase 2 testing, Simpson Strong-Tie did virtually everything, from procuring the materials and constructing the specimens to leading the complex analysis of the data. This article is adapted from the Proceedings of the 10 th U.S. National Conference on Earthquake Engineering, which will take place July 21-25, 2014 in Anchorage, Alaska, hosted by the Earthquake Engineering Research Institute (EERI). This article is derived from an Interim Report on Testing of Tension-Only Steel Anchor Rods Embedded in Reinforced Concrete Slabs, SEAOC Convention Proceedings, September 2013, San Diego, CA.

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it was necessary not only to adjust the calculation to reflect the average instead of a 5 percent fractile result, but also to disregard the crack reduction factor even though bi-directional cracks through the cone were clearly evident during the testing and prior to cone failure, which should result in a larger crack reduction than just one-directional cracking. The inability of cracked concrete to transfer tensile stress across the crack in the cone area changes the internal force distribution in the cone, resulting in a smaller breakout force; yet it is also known that the amount of reinforcing in cracked concrete testing can significantly influence the result. More research is needed to understand how the crack reduction factor should be applied in concrete with significant flexural reinforcing passing through the cone. Developing anchor reinforcement on each side of the failure plane can further increase tension capacity. Different arrangements of shear reinforcement can shift the failure plane of the shear cone farther away from the anchor, increasing the area of the cone and the tension capacity of the anchor. Interim results suggest that some change to existing regulations may be warranted when considering this specific condition (anchor reinforcing in “thin” reinforced elements). Other groups (e.g. NEHRP Provisions Update Committee, Issue Team 3, and Simpson Strong-Tie) are further exploring this effort analytically and experimentally.▪ W. Andrew Fennell, P.E., CPEng, SECB (andy.sclabs@earthlink.net), is a Principal Engineer at Scientific Construction Laboratories, Inc., in Lafayette, California. Gary L. Mochizuki, P.E., S.E., LEED AP (gary@structsol.com), is a Principal Engineer at Structural Solutions, Inc., in Walnut Creek, California. Kevin S. Moore, P.E., S.E., SECB (ksmoore@sgh.com), is a Principal at Simpson, Gumpertz and Heger in San Francisco, California. Steven E. Pryor, P.E., S.E. (spryor@strongtie.com), is the International Director of Building Systems at Simpson Strong-Tie® Company, Inc., in Pleasanton, California. Geoffrey A. Laurin (laurincons@gmail.com), is a General Contractor and Principal at Laurin Consulting, LLC, in El Dorado, California.

STRUCTURE magazine

March 2014

Just the FAQs questions we made up about ... Masonry

Tested shear wall.

Reinforcing Shear Walls in Seismic Zones By Max L. Porter, Ph.D., P.E., Dist. M. ASCE, F. SEI, F. TMS, F. ACI, F. ACFE, D. ASFE, CFC and Gregory P. Baenziger, P.E.

Max L. Porter, Ph.D., P.E., Dist. M. ASCE, F. SEI, F. TMS, F. ACI, F. ACFE, D. ASFE, CFC, is University Professor Emeritus of Civil Engineering at Iowa State University (ISU), Ames, Iowa. He has served as president of seven organizations, including The Masonry Society and the Structural Engineering Institute, and chaired the Masonry Standards Joint Committee (MSJC). He has also chaired several national code committees in the areas of masonry, reinforced concrete, and FRP. Greg Baenziger, P.E., is a Ph.D. student at ISU. He has been an instructor at the university and is the primary author of the research mentioned in the article. Greg is currently working at Bratney Companies in Des Moines, Iowa while he continues his degree studies.

Question

I have been told that horizontal joint reinforcement is not allowed by the 2011 Building Code Requirements for Masonry Structures (TMS 402-11/ACI 530-11/ASCE 5-11) for shear reinforcement in high seismic zones. Where is this stated and why? Are there plans to change this to allow joint reinforcement?

Response Use of joint reinforcement has been allowed for many years and will continue to be allowed as prescriptive reinforcement. Joint reinforcement provides the added benefits of improved crack restraint and satisfying prescriptive and horizontal reinforcement spacing requirements for all types of masonry shear walls. Regarding shear reinforcement in high seismic zones, joint reinforcement can be used as primary shear reinforcement according to the MSJC code (TMS 402-11/ACI 530-11/ASCE 5-11) provided you use the chapter on Allowable Stress Design. However there are limits if you use the chapter on Strength Design. Section 3.1.8.3 restricts the yield strength of the joint reinforcement wire used as primary shear reinforcement to 60 ksi (414 MPa) or less. Things change in the chapter on Strength Design with the advent of the 2013 edition of the code. These include: 1) Joint reinforcement consisting of 3/16inch (4.8-mm) diameter wires can have yield strengths up to 85 ksi (586 MPa) (Section 9.1.8.3.2) as primary shear reinforcement. Hot dipped galvanized

58 March 2014

3/16-inch (4.8-mm) wire commonly has a yield strength of approximately 85 ksi (586 MPa). 2) For joint reinforcement used as the primary shear reinforcement, there will be limits on the diameter of wire and on the quantity and spacing of wire based in part on the Seismic Design Category (SDC). These lower limits are that the joint reinforcement wire must be at least 3/16-inch (4.8-mm) diameter wire and that the quantity and spacing of wire be at least: a) two 3/16-inch (4.8-mm) wires per bed joint at 16-inch maximum spacing in partially grouted walls for SDC A and B; b) two 3/16-inch (4.8-mm) wires per bed joint at 8-inch maximum spacing in partially grouted walls for SDC C through F; and, c) four 3/16-inch (4.8-mm) wires per bed joint at 8-inch maximum spacing in fully grouted walls for SDC C through F. Research has shown that 1) joint reinforcement can function as shear reinforcement if a sufficient quantity of joint reinforcement is provided to satisfy strength requirements and 2) that two 3/16-inch (4.8-mm) diameter longitudinal wires in each bed joint have sufficient strength and strain elongation capacity to act as primary shear reinforcement and can perform equivalent to bond beams with deformed reinforcement at 48 inches (1.22 m) on center.

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STR_7-13

LegaL PersPectives

discussion of legal issues of interest to structural engineers

The Hyatt Regency Disaster Revisited By Matthew R. Rechtien, P.E., Esq.

he more than three decades that have passed since the collapse of hanging walkways at Kansas City’s Hyatt Regency Hotel have turned that catastrophe into the textbook cautionary tale for structural engineers. No other collapse (save maybe the World Trade Center) has led me to as many discussions within and out of the profession. I recall my own mentor, Javier Horvilleur, discussing it with laypeople and peers at least twice a year. It is the profession’s quintessential “teachable moment.” So, with Santayana’s warning in mind that “[t]hose who cannot remember the past are condemned to repeat it,” this article takes a fresh look, from a legal perspective, at some of the enduring lessons it holds.

Background In the late 1970s, Gillum-Colaco, Inc. (“GCE”) agreed to perform all structural engineering services for the design and construction of the hotel. GCE designated Jack Gillum (“Gillum”) as engineer of record, but put Daniel Duncan (“Duncan”) in day-to-day charge of the project. GCE’s design called for second and fourth floor walkways to hang from the roof of the atrium they spanned by six continuous or “single” steel rods that were to connect to each walkway through “box” connections. GCE designed – insufficiently, it turns out; by one estimate, the design expressed in the drawings carried only 60% of what the local building code required – the box connections so that the second floor walkway would not hang from the fourth floor walkway. The steel fabricator proposed a “double rod” system, in which the six rods would become twelve, and the second floor walkway would hang from the fourth floor walkway, amplifying the load on the box connections. Duncan, who seemingly failed to detect the load amplification, approved the fabricator’s proposal and vouched for its soundness without doing what (he later admitted) would have been necessary to confirm it. Neither Gillum nor Duncan reviewed the shop drawings reflecting the change, despite GCE’s policy that drawings of non-redundant connections receive more than the technician’s review of sizes and materials that they got.

Disciplinary Proceedings In 1981, after just a year of service, the box connections failed and the walkways collapsed, killing more than 100 revelers. In 1984, Missouri’s Board for Professional Engineers commenced disciplinary proceedings against Duncan, Gillum and GCE. After a 27-day hearing, the tribunal (the “Commission”) conducting the proceedings issued findings covering 442 pages. The Commission found all three grossly negligent and revoked their licenses.

Legal Challenges Duncan, Gillum and GCE (the “Appellants”) unsuccessfully appealed their discipline up to the Missouri Court of Appeals. That court rejected all of their “legalistic” challenges. It rejected their claim that the governing statute’s gross negligence standard was unconstitutionally vague, and that discipline was improper because no one shortcoming amounted to gross negligence. It also rejected the Appellants’ claim that their negligent design of the rods was not a basis for discipline because the rods did not fail, responding that in a disciplinary proceeding – unlike in a negligence lawsuit – causation is irrelevant. In addition, the Court rejected Gillum’s and GCE’s contention that neither could be disciplined vicariously for Duncan’s misconduct. The Court also rejected the Appellants’ more substantive attacks on the sufficiency of the evidence, and pointing to the following: 1) Duncan was responsible for designing and approving the building structure; 2) The walkways fell in that scope; 3) “… [T]he walkways offered a potential of great danger to human life if defectively designed;” 4) Duncan approved the fabricator’s change, recommended it to the architect, and approved shop drawings reflecting it without confirming its acceptability; 5) The change effectively doubled the box connection load; and, 6) Duncan never reviewed the shop drawings, even though such review is an “engineering function” that even GCE’s in-house policies required he do.

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March 2014

Courtesy of Dr. Lee Lowery, Jr., P.E.

On those facts, the Court affirmed that “[t]he conduct of Duncan from initial design through shop drawing review and through the subsequent requested connection review … supports the Commission’s finding of conscious indifference to [non-delegable] professional duty.” The Court also affirmed Gillum’s “gross negligence” and resulting discipline because he failed, as engineer of record, “to assure that the Hyatt engineering designs and drawings were structurally sound … prior to impressing thereupon his seal,” and failed “to assure adequate shop drawing review.”

Lessons The story of the Hyatt collapse is, for me, a little like the classic film The Natural. With each look, I get a bit something different out of it; it is a Rorschach test. The remainder of this article shares a few of my observations. “Heavy lies the head that wears the crown” – William Shakespeare The Hyatt case demonstrates that, as a legal matter, the “buck stops” at the engineer of record. The Court did not focus on Gillum’s errors of commission, or failings as an engineer, but on his failings as a manager. It found that as the engineer of record, “Gillum was by statute responsible for” not only the drawings that his firm drafted, but also the resulting shop drawings. Gillum “accepted such responsibility when he entered into the contract and utilized his seal.” The Court rejected Gillum’s every effort to downplay these legal implications of affixing his seal and disown the failed connection. Citing

what it dubbed the “plain and unambiguous language” of Missouri’s “controlling regulatory statutes,” the Court concluded that Gillum’s sealing of the plans made “him responsible for the entire engineering project and all documents connected therewith …” including the design of connections “whether he in fact designs them himself or not.” Bottom line: by stamping the structural drawings without qualification, Gillum was by law responsible for the structure and the collapse. “Custom reconciles us to everything” – Edmund Burke Customs are a pervasive and persistent source of law in our “common law” legal system. The Hyatt case is, however, an object lesson on the limitations of those customs as a source of law. It demonstrates how customs do not survive collisions with statutes. The Court rejected Gillum’s argument that “usual and customary engineering practices” entitled him to rely on the fabricator for the design of the box connections because it conflicted with the plain language of Missouri’s engineering practice statute. The lesson is that while custom may have shifted the practical responsibility for the box connection design, it could not have affected a transfer of the legal responsibility, a lesson Gillum learned the hard way. Keep “the main thing the main thing” – Steven Covey The Hyatt case demonstrates that not all beams are created equal. Perhaps recognizing its audience, the Court put this in mathematical terms: “the level of care required of a professional engineer is directly proportional to the potential for harm arising from his design …” It was the importance of the box connections that amplified the gap between what the Appellants did, and what they should have done. This importance elevated what the Court said “might [have] constitute[d] inadvertence” had “no danger exist[ed] … to conscious indifference,” i.e., gross negligence, because “the potential danger to human life [wa]s great.” Accordingly, it is not just a good idea for engineers to allocate their scarce design resources with a mind towards the potential for harm; it is the law. This potential for harm matters in at least one other way. It has been said that “nondoctrinal” – legally irrelevant – facts often drive judicial decision-making. The great jurist Oliver Wendell Holmes, Jr. expressed this as “hard cases make bad law.” Once the

Appellants’ gross negligence was established, their discipline (or liability) should not necessarily have been driven by the magnitude of the horror that happened to result. Reading the Court’s opinion, however, and its multiple allusions to the patent danger of the walkways, it is difficult to avoid the conclusion that Gillum, Duncan and GCE left the Court with a particularly “hard case,” which may have factored into the hard result. “Attitude is a little thing that makes a big difference” – Winston Churchill The Hyatt case demonstrates the price of admission to the structural engineering profession. Gillum faced discipline not just for gross negligence, but also for what the Court called “unprofessional conduct … in his refusal to accept his responsibility … and his denial that such responsibility existed.” It was not just what he did that mattered, but his attitude and his defense of the allegations against him. The Court noted that Gillum’s “refusal to accept a responsibility so clearly imposed by the [engineering practice] statute manifests both the gross negligence and unprofessional conduct found by the Commission.” The Court agreed with the Commission that Gillum’s “cavalier” attitude about his “responsibilities as an engineer” and his steadfast refusal to admit the responsibility that law imposed on him, as the engineer of record, was a separate and independent breach of his obligations as a professional engineer. The Hyatt case is a grim reminder that whatever may go in hard-nosed commerce outside of the profession will not necessarily pass muster within it, and that the esteem that comes with membership in a learned profession may be hefty. “… Hoist with his own petard… ” – William Shakespeare

All firms have internal policies; they serve worthy purposes, like managing risk. The Hyatt case demonstrates not the imprudence of these policies, but of adopting them casually. That is because one lesson of the case is that having a policy that you do not follow may be worse than not having it at all. Because, as the Court put it, the Appellants’ “own internal procedures” “called for a detailed check of all special connections” like the box connection; the Court treated their failure to make the check as proof of their negligence. A policy that was (surely) intended to reduce Appellants’ exposure ended up increasing it.

“Investors don’t like uncertainty” – Kenneth Lay

Finally, the Hyatt case demonstrates that ambiguity in design injects risk into the shop drawing review process. Though the case involved many missteps, the first was Duncan’s failure to address the box connection design in his drawings. Although Duncan testified that “he intended for the fabricator to design the [box] connections,” his drawings failed to include, for example, the loads that might have communicated his intention to the fabricator. As a result, the fabricator “prepared its shop drawings on the basis that the connections shown on the design drawings had been designed by the structural engineer.” Thus, although either engineer or fabricator could have designed the connection, the ambiguity of the drawings, their failure to express clearly the engineer’s intentions, meant that neither did. Of course, there was still time to rectify this oversight. Nevertheless, the clear lesson is that ambiguity in the design injected tremendous risk into the shop drawing process, a process that is often poorly suited to address significant design decisions.

Conclusion The Hyatt Regency disaster was a tragedy for the victims and for the engineers involved. In a larger sense though, it was also a tragedy for the structural engineering profession. The best we can do to honor the victims, the engineers involved, and the profession, is to listen to this alarm bell, heed the enduring lessons, and improve our practices. My hope is that this article contributes to that effort.▪ Matthew R. Rechtien, P.E., Esq. (MRechtien@BodmanLaw.com), is an attorney with Bodman PLC in Ann Arbor, Michigan, where he specializes in construction law, commercial litigation, and insurance law. Prior to becoming a lawyer, he practiced structural engineering in Texas for five years. The facts and quotations in this article are adapted from Duncan v. Missouri Board for Architects, Professional Engineers and Land Surveyors, 744 S.W.2d 524 (Mo. App. E.D. 1988). The full opinion by Missouri’s Court of Appeals is a good read for anyone with further interest in the subject.

Disclaimer: The information and statements contained in this article are for information purposes only and are not legal or other professional advice. Readers should not act or refrain from acting based on this article without seeking appropriate legal or other professional advice as to their particular circumstances. This article contains general information and may not reflect current legal developments, verdicts or settlements; it does not create an attorney-client relationship.

STRUCTURE magazine

March 2014

Software UpdateS ADAPT Corporation Phone: 650-306-2400 Email: info@adaptsoft.com Web: www.adaptsoft.com Product: ADAPT-Builder Suite Description: For the integrated and efficient design of concrete buildings. Streamline your process by using one model to run your gravity and lateral designs. Don’t waste time by maintaining separate data for your slab, foundation, column and wall designs. Designs post-tensioned and mild reinforced projects. Seamlessly integrates with Revit Structure.

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Software UpdateS Simpson Strong-Tie Phone: 925-560-9000 Email: web@strongtie.com Web: www.strongtie.com Product: Simpson Strong-Tie® Holdown Selector Web App Description: Holdowns are typically used to prevent shearwall segments from overturning. The Simpson Strong-Tie Holdown Selector is a quick and easy tool that selects the optimum solution based on the type of installation, factored load and the species of the post. Visit the website to use the app. Product: Simpson Strong-Tie® Anchor Designer™ Software Description: The latest anchorage design tool for structural engineers. It quickly and accurately analyzes an existing design or suggests anchorage solutions based upon user-defined design elements in cracked and uncracked concrete conditions. The real-time design features a fullyinteractive 3D graphic user interface. Product: Simpson Strong-Tie® Literature Library Mobile App Description: It’s easy to take Simpson Strong-Tie catalogs with you. With the new and improved version of the Literature Library app, you can access all Simpson Strong-Tie catalogs, fliers and technical bulletins on your mobile devices. Download the app and start customizing your “library.” Visit the website to download the app.

SPACE GASS Phone: 559-897-5697 Email: steve@spacegass.com Web: www.spacegass.com Product: SPACE GASS 12 Description: Contains new super fast analysis and graphics engines, making it one of the fastest structural analysis programs around. Large models of 10,000 nodes or more analyze in just a few seconds, and you can view your fully rendered results in smooth real-time with the new graphics engine.

Standards Design Group, Inc. Phone: 800-366-5585 Email: info@standardsdesign.com Web: www.standardsdesign.com Product: Wind Loads on Structures 4 Description: Performs computations in ASCE798, 02 or 05, Section 6 and ASCE7-10, Chapters 26-31. Computes wind loads by analytical method rather than simplified method, provides basic wind speeds from a built-in version of wind speed, allows user to enter wind speed, has numerous specialty calculators.

news and information from software vendors

Product: Window Glass Design 5 Description: Performs all required calculations to design window glass according to ASTM E 1300-09. This software also performs window glass design using ASTM E 1300 02/03/04, ASTM E 1300-98/00 and ASTM E 1300-94. GANA endorses WGD5 as best tool available in designing window glass to resist wind and longterm loadings.

Strand7 Pty Ltd Phone: 252-504-2282 Email: anne@beaufort-analysis.com Web: www.strand7.com Product: Strand7 Description: An advanced, general purpose, FEA system used worldwide by engineers, designers, and analysts for a wide range of structural analysis applications. It comprises preprocessing, solvers (linear and nonlinear static and dynamic capabilities) and postprocessing. Features include staged construction, quasi-static solver for shrinkage and creep/relaxation problems.

StrucSoft Solutions Phone: 514-731-0008 Email: info@strucsoftsolutions.com Web: www.strucsoftsolutions.com Product: MWF Advanced Floor Description: Extend the power and reach of Revit® & MWF with MWF Advanced Floor. This upgrade to the popular framing extension allows engineers and estimators to engineer and determine optimum joists for their projects. Take advantage of a finite element analysis engine and the almost limitless configurations you expect from MWF.

Structural Engineers Inc. Phone: 540-731-3330 Email: tmmurray@floorvibe.com Web: www.floorvibe.com Product: FloorVibe v2.10 Description: Proposed floor designs can be analyzed to determine if they meet the human tolerance limits in the AISC Design Guide 11 Floor Vibrations due to Human Activity using FloorVibe. Floor framing can be hot-rolled, build-up sections, or joists. Expert advice is provided for all required input and for results. All Resource Guides and Updates for the 2014 Editorial Calendar are now available on the website, www.STRUCTUREmag.org. Listings are provided as a courtesy. STRUCTURE® magazine is not responsible for errors.

STRUCTURE magazine

March 2014

StructurePoint Phone: 847-966-4357 Email: info@structurepoint.org Web: www.StructurePoint.org Product: Reinforced Concrete Analysis and Design Programs Description: PCA’s concrete design suite is now: spSlab, spColumn, spMats, spWall, spBeam & spFrame. Formerly pcaSlab, pcaColumn, pcaMats, pcaWall, pcaBeam & pcaFrame. These programs are widely used for analysis, design & investigation of reinforced concrete buildings, bridges and structures. The programs provide support for ACI 318 and CSA A23.3 standards.

Struware, LLC Phone: 904-302-6724 Email: email@struware.com Web: www.struware.com Product: Struware Code Search Description: Provides easy to use updated structural engineering software. So quick and easy the software will pay for itself the first time you use it. Programs include Code Search (wind, seismic, snow, dead & live loadings), CMU walls, tilt-up concrete walls, floor vibration and concrete beams. Demos at the webiste.

USP Structural Connectors Phone: 952-898-8772 Email: info@uspconnectors.com Web: www.uspconnectors.com Product: USP Specifier™ Description: Simplifies the access to information on over 3,000 structural connectors through an intuitive and graphical desktop interface. Looking up connector capacities, viewing code evaluation reports and even mapping from competitor products to USP products are a snap with just a couple of clicks. Download Specifier today from the website.

WoodWorks® Software Phone: 800-844-1275 Email: sales@woodworks-software.com Web: www.woodworks-software.com Product: Design Office 10 Suite Description: Wood Engineering software: Conforms to IBC 2012, ASCE 7-10, NDS 2012, SDPWS 2008; SHEARWALLS: designs perforated and segmented shearwalls; generates loads; rigid and flexible diaphragm distribution methods. SIZER: designs beams, columns, studs, joists up to 6 spans; automatic load patterning. CONNECTIONS: Wood to: wood, steel or concrete.

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Advanced Seismic Systems and Code Evolution By Jerod G. Johnson, Ph.D., S.E.

he past decades have seen major progress toward the broad utilization of advanced nonlinear analysis methods for seismic design. Many of us have witnessed continual development and evolution of the modern standard for nonlinear seismic design – ASCE 41, now titled Seismic Evaluation and Retrofit of Existing Buildings. Its precursors include FEMA 273 and FEMA 356, not to mention countless others (e.g., ATC 40) that emerged even earlier in an effort to address the vast inventory of infrastructure that cannot satisfactorily qualify under standards for new construction. It has been interesting to witness the evolution of modern methodologies as they have progressed from “White Papers” to “Guidelines” and then to “Pre-Standards” and finally full-fledged “Standards”. Countless individuals have contributed untold hours and shared vast amounts of experience and knowledge to bring us modern seismic rehabilitation requirements. These efforts have led to a clearly pragmatic approach for dealing with the threat of earthquakes. The use of basic nonlinear analysis methods can even demonstrate the frailties of some of the most prolific seismic-force-resisting systems found in modern infrastructure. As an example, consider a concentric braced frame. What are the consequences of braces buckling in compression? It is an interesting rhetorical question and has led some to believe that if the concentric braced frame were to be introduced as a new product today for regions of high seismicity, it would be an uphill battle to win approval. However, its successful use over many decades has yielded a grandfathering of sorts, accompanied by some adaptations; very large beams (to account for unbalanced brace forces) in chevron or “V” configurations come to mind. The widespread adaptation and use of nonlinear analysis methods have shed light on behaviors that, in worst cases, have been ignored, in better cases are only misunderstood, and in the best cases are handled head-on by the structural engineer. We often deal with irregular behaviors and geometries, and go to painstaking efforts to qualify our projects as “regular” so as to be acceptable following the prescriptive code standards.

However, nonlinear analysis methods (static pushover) can demonstrate that even a perfectly symmetrical, conventionally braced frame structure will develop an extreme torsional irregularity when considering the prescribed 5% accidental eccentricity. This is because the likelihood of complementary braces on opposing sides buckling simultaneously is extremely low. How do we address the consequences of this behavior? One valid, yet economically impractical approach is to design structures to remain elastic (R=1). Good luck explaining to your clients why your design is four times the cost of your competitor’s! Such an approach may seem extreme, but has actually appeared among published opinion statements regarding future seismic code development. The more pragmatic way forward is to embrace and design systems that are better prepared to handle the nonlinearity and mitigate its global effects on the structure. Such systems are already recognized in the code, but in an indirect manner. Simply observe the highest prescribed R factors to identify the systems with superior nonlinear performance. Among these, buckling restrained braced frames have emerged as a solution preferred by many. Other steel systems include special moment frames, eccentric braced frames and steel plate shear walls. Each of these systems has a demonstrated ability for well-balanced and primarily symmetric hysteretic behavior. In essence, frames using these systems can experience repeated cycles of elasto-plastic deformation while maintaining their ability to support gravity loads. In so doing, they dissipate energy in a stable, controlled and targeted manner. The emergence and utilization of nonlinear analysis methods afford engineers the tools to address seismic design in this manner. While conventional approaches are still valid, the increasing ease of use for nonlinear methods make the application of the R factor to the global structure seem increasingly less reliable, perhaps even less practical. Even so, reconnaissance efforts following major events suggest that satisfactory performance can be realized using traditional methods, which have served us well.

STRUCTURE magazine

March 2014

It seems somewhat ironic, though, to see virtually all modern structural analysis software, coupled with amazingly powerful desktop computers, automatically develop the prescribed pseudo-static seismic forces in virtually the same manner as hand calculations and slide rules from many years ago. The tools that we use have powers and capabilities for seismic analyses far beyond most of the analytical tasks to which we apply them. Seems a bit like swatting flies with a sledge hammer! What will the future hold? Nonlinear methods have a demonstrated ability to provide a more reliable outcome and the means for meeting specific seismic performance objectives. Will direct nonlinear analysis methods replace the current simplified and indirect methods? Time will tell.

Clarification The author’s article, Development Length: More Complexity or Saving Grace? (STRUCTURE, December 2013), concluded by stating, “Furthermore, ACI 318 section 12.2.5 allows for reduction of development length and lap splice in direct proportion to the amount of excess reinforcement provided. Owing to the discreteness of bar sizes, excess reinforcement can usually be quantified such that embedment and lap splicing requirements can be rationally adjusted accordingly.” While this is correct for development length and embedment, an alert reader pointed out that such a reduction is now explicitly prohibited for lap splice length by ACI 318 section 12.15.1. Consequently, option (b) presented earlier in the text is not actually a viable alternative, unless the requirements of ACI 318 section 12.15.2 can be satisfied, which would allow the lap splices to be Class A rather than Class B.▪ Jerod G. Johnson, Ph.D., S.E. (jjohnson@reaveley.com), is a principal with Reaveley Engineers + Associates in Salt Lake City, Utah.

award winners and outstanding projects

Spotlight

UC Berkeley California Memorial Stadium Protecting and Strengthening a Landmark on an Active Fault By David Friedman, S.E., René Vignos, S.E. and Chris Petteys, S.E. Forell/Elsesser Engineers, Inc. was an Outstanding Award Winner for the UC Berkeley California Memorial Stadium Seismic Upgrade project in the 2013 NCSEA Annual Excellence in Structural Engineering awards program (Category – Forensic/ Renovation/Retrofit/Rehabilitation Structures).

uilt as a memorial to fallen heroes of World War I from California, Memorial Stadium has endured as one of the most picturesque venues in college football from its opening in 1923 for the “Big Game” versus Stanford to the present day. Over the years, the original Stadium had shown its age and also shown signs of how the earth is slowly creeping on the Hayward Fault beneath it. This movement had shown up through leaning columns, building cracks, and column offsets and misalignments. The challenge for the design team was how to preserve this monument in place and ensure the safety of its occupants should a large earthquake occur. Due to the site location and geology, design for large ground motions dominated the engineering decisions that were made about the structure. The unique thing about Memorial Stadium is that it not only needed to be designed for the strong ground motions, but also needed to accommodate potential surface rupture of the Hayward Fault below certain portions of the building. Geologists on the project located and mapped the Hayward Fault around the Stadium. Due to the historic nature of the Stadium and its location adjacent to campus, it was decided to retrofit the structure in place and develop a solution that safely accommodated the potential surface rupture. Through an interdisciplinary effort between structural engineers, geologists, seismologists, and geotechnical engineers, a scheme was designed to accommodate the estimated 6 feet of lateral movement and simultaneous 2 feet of vertical movement that could occur on site. For Memorial Stadium, this meant breaking the seating bowl into discrete blocks where the fault crossed so that these portions of the building could move in response to possible surface rupture without affecting the rest of the structure. The challenge was to determine the best location and the optimum width for these joints while still accommodating the architectural

design of the building. Through study of this problem with advanced finite-element modeling and through physical study models, along with many meetings and debates between the engineers, geologists, and seismologists, a final scheme was developed that satisfied the requirements of all the stakeholders. While a large portion of the engineering effort was devoted to solving the issue of accommodating surface rupture, another large engineering challenge also dominated the design of the stadium. One of the signature architectural features of the Stadium is a two story, 375-foot long press box that hovers above the west side of the seating bowl. This press box not only houses print, radio, and TV media, but also has a club space with views and seating facing the field and a dramatic 25-foot cantilever balcony that faces campus with views of the San Francisco Bay and the Golden Gate beyond. While designing a press box on limited supports to give the appearance of “hovering” is a challenge, the real challenge came in safely bracing this structure for large potential ground motions at this site. The press box structure is designed as a 3D space truss with diagonals in all directions which allows the structure to cantilever toward the field, span 90 feet between vertical supports, and be braced seismically between the supporting cores. Even with the extensive bracing, the press box is still relatively flexible when compared with the concrete stadium seating bowl below. As the design progressed, the engineers realized that a large flexible steel structure on four slender concrete core walls attaching to a very stiff triangular seating bowl created a dynamic incompatibility within the combined structure. In addition to conflicting dynamics, there would also be a high concentration of damage to the core walls where they attached to the seating bowl, this was a big concern since the cores serve as the only path for exit stairs and elevators, as well as providing gravity support for the press box itself.

STRUCTURE magazine

March 2014

Through many iterations of design, a solution that simultaneously solved the biggest concerns was developed: provide a separation between the bowl and press box structures and allow them to move independently, and link them with fluid viscous dampers. These dampers buffer the movements between the two structures, tempering the accelerations in the press box. To protect the tall slender core walls supporting the press box, the engineers intended them to rock as a rigid body at their base instead of trying to fix them to the foundation system rigidly. Vertical post-tensioning was utilized in all the cores and allows them to rock by providing a restoring force and a constant compression force to help them remain elastic. The combination of the vertical post-tensioning and fluid viscous dampers allows the two structures to work together to resist strong ground motions. With this project complete, Memorial Stadium is safe for the effects of surface rupture and ground shaking from the Hayward fault while retaining it’s iconic historic façade and providing UC Berkeley with all the amenities of a modern facility. Memorial Stadium is now the envy of the PAC12 and provides Cal fans with a safe place to spend a fall afternoon cheering on their team.▪ David Friedman, S.E., is a Senior Principal with Forell/Elsesser, former President and CEO, and currently the Chair of the Board of Directors. David can be reached at daf@forell.com. René Vignos, S.E., is a Principal with Forell/Elsesser. René can be reached at r.vignos@forell.com. Chris Petteys, S.E., is an Associate and licensed structural engineer at Forell/Elsesser. Chris can be reached at c.petteys@forell.com.

GINEERS

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New NCSEA Webinar Subscription Plan Off to a Great Start

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NCSEA Webinars March 13, 2014 Floor Vibrations: Technical Background and AISC Design Guide 11, Part 1 Brad Davis, Ph.D., S.E., Department of Civil Engineering, University of Kentucky March 27, 2014 Floor Vibrations: Technical Background and AISC Design Guide 11, Part 2 Brad Davis, Ph.D., S.E., Department of Civil Engineering, University of Kentucky April 10, 2014 Load Generators – What Exactly is My Software Doing? Sam Rubenzer, P.E., S.E., FORSE Consulting

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April 24, 2014 Structural Fire Resistance – Overview, Codes and Standards Background Nestor R. Iwankiw, Ph.D., P.E., S.E., Senior Engineer, Hughes Associates, Inc. EN

NCSEA News

NCSEA’s new Webinar Subscription Plan is off to a roaring start. NCSEA members are taking advantage of this new and exciting program to plan their continuing education for the year. The Subscription Plan offers unlimited live NCSEA webinars for one year for only $750. The subscription plan is open only to NCSEA members – members of NCSEA Member Organizations and NCSEA Sustaining, Associate and Affiliate members, and membership status will be verified. It includes live NCSEA webinars only and has a guarantee that a minimum of ten webinars, and up to 24, will be offered. A subscription plan can be started at any time. The one-year plan begins on the first day of the month in which the subscription application is received. Each webinar includes speaker slides, notes, and one free PDH certificate. If others watch the webinar with the subscriber and wish to obtain a certificate, they may purchase a certificate for $30 within two weeks of completing the webinar. After two weeks, additional certificates will not be available. The individual subscription plan was developed in response to numerous requests from small structural engineering firms and sole proprietorships, for webinars they could afford, states NCSEA Executive Director Jeanne Vogelzang. “To date, 90

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Diamond Reviewed

These courses will award 1.5 hours of continuing education. Approved for CE credit in all 50 States through the NCSEA Diamond Review Program. Time: 10:00 AM Pacific, 11:00 AM Mountain, 12:00 PM Central, 1:00 PM Eastern. NCSEA offers three options for registrations to NCSEA webinars: ala carte, Flex-Plan, and Yearly Subscription. Visit www.ncsea.com for more information or call 312-649-4600.

STRUCTURE magazine

“There are several providers for webinars in the industry. NCSEA’s quality of presenters are excellent and the subject matter is very relevant to our office’s needs for both Junior and Senior Structural Engineers. To get this type of quality for the annual price is a good investment for our team and our firm.” Julian Lineham, PE, SECB Principal, Studio NYL Structural Engineers

“We have appreciated the variety of topics and quality of presentations given in the past NCSEA-sponsored webinars, and we are confident that our 2014 annual subscription purchase will provide a great continuing education value to our firm’s engineers.” John Riley, S.E., P.E. Principal, Quantum Consulting Engineers LLC

percent of subscribers are names that are new to us and the webinar program; and some of these new subscribers are engineers who have joined their local SEA’s, in order to benefit.”

Safety Assessment Program Webinar Set for May 16 NCSEA will once again offer a Post-Disaster Safety Assessment webinar on May 16. The California Office of Emergency Management (CalOES) Safety Assessment Program (SAP), presented by NCSEA, is one of only two post-disaster assessment programs that will be compliant with the requirements of the forthcoming Federal Resource Typing Standards for engineer emergency responders. The course is offered twice per year as an NCSEA webinar. Based on ATC-20/45 methodologies and documentation, the SAP training course provides engineers, architects and code-enforcement professionals with the basic skills required to perform safety assessments of structures following disasters. Licensed design professionals and certified building officials will be eligible for SAP Evaluator certification and credentials following completion of this program and submission of required documentation. The NCSEA CalOES SAP Program consists of three webinar segments over one day’s time, and is taught by Jim Barnes, who has worked for 20 years for the State of California’s disaster agency. He has served as the lead statewide coordinator for the Safety Assessment Program (SAP) for five years, and has given around 150 classes in the subject since 2004.

Mark Your Calendar! NCSEA Annual Conference September 17-20, 2014 Astor Crowne Plaza Hotel, New Orleans March 2014

March 20 – 21, 2014 • The Meritage Resort & Spa, Napa, California

Featuring the following topics: • Negotiating to Get the Value & Compensation You Deserve • Baby Boomers Retirement Logjam • Ownership Transition Case Studies: Real-Life Applications

Take Your Seat at the Table

NCSEA News

Winter Leadership Forum

Discuss and develop new strategies and best practices, and learn what other principals are doing and thinking. News from the National Council of Structural Engineers Associations

• Dealing with Conflict in the Workplace • Dispute Resolution Options: Mediation, Arbitration or Litigation?

• Structuring a Defense When You’ve Been Sued

Get a taste of Napa Valley and network with other leaders at the Thursday Wine-Tasting Reception in the Meritage Wine Cave!

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STRUCTURE magazine | March 2014

Articles inside

Structural Forum

Spotlight

Engineer’s Notebook

Legal Perspectives

Just the FAQs

Historic Structures

Structural Testing

Structural Forensics

Structural Sustainability

InSights

Building Blocks

Structural Design

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